121
7 Mohamad H. Hussein Jerry A. DiMaggio Vice President GRL Engineers, Inc., Orlando, Florida Partner, Pile Dynamics, Inc., Cleveland, Ohio Senior Geotechnical Engineer Federal Highway Administration Washington, DC GEOTECHNICAL ENGINEERING I n a broad sense, geotechnical engineering is that branch of civil engineering that employs scientific methods to determine, evaluate, and apply the interrelationship between the geologic environment and engineered works. In a practical context, geotechnical engineering encom- passes evaluation, design, and construction invol- ving earth materials. The broad nature of this branch of civil engi- neering is demonstrated by the large number of technical committees comprising the Geo-Institute of the American Society of Civil Engineers (ASCE). In addition, the International Society for Soil Mech- anics and Geotechnical Engineering (ISSMGE) includes the following 31 Technical Committees: Calcareous Sediments, Centrifuge and Physical Model Testing, Coastal Geotechnical Engineering, Deformation of Earth Materials, Earthquake Geo- technical Engineering, Education in Geotechnical Engineering, Environmental Geotechnics, Frost, Geophysical Site Characterization, Geosynthetics and Earth Reinforcement, Ground Improvement, Ground Property Characterization from In-situ Testing, Indurated Soils and Soft Rocks, Instru- mentation for Geotechnical Monitoring, Landslides, Limit State Design in Geotechnical Engineering, Micro-geomechanics, Offshore Geotechnical Engin- eering, Peat and Organic Soils, Pile Foundations, Preservation of Historic Sites, Professional Practice, Risk Assessment and Management, Scour of Foundations, Soil Sampling Evaluation and Inter- pretation, Stress-Strain Testing of Geomaterials in the Laboratory, Tailings Dams, Tropical and Residual Soils, Underground Construction in Soft Ground, Unsaturated Soils, and Validation of Computer Simulations. Unlike other civil engineering disciplines, which typically deal with materials whose proper- ties are well defined, geotechnical engineering is concerned with subsurface materials whose prop- erties, in general, cannot be specified. Pioneers of geotechnical engineering relied on the “observa- tional approach” to develop an understanding of soil and rock mechanics and behavior of earth materials under loads. This approach was enhan- ced by the advent of electronic field instrumen- tation, wide availability of powerful personal computers, and development of sophisticated numerical techniques. These now make it possible to determine with greater accuracy the nonhomo- geneous, nonlinear, anisotropic nature and beha- vior of earth materials for application to engineering works. Geotechnical engineers should be proficient in the determination of soil and rock properties, engineering mechanics, subsurface investingation methods and laboratory testing techniques. They should have a thorough knowledge of design methods, construction methods, monitoring/inspec- tion procedures, and specifications and contracting practices. Geotechnical engineers should have broad practical experience, in as much as the practice of geotechnical engineering involves art as much as science. This requirement was clearly expressed by Downloaded from Digital Engineering Library @ McGraw-Hill (www.digitalengineeringlibrary.com) Copyright © 2004 The McGraw-Hill Companies. All rights reserved. Any use is subject to the Terms of Use as given at the website. Source: Standard Handbook for Civil Engineers

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Page 1: Senior Geotechnical Engineer GEOTECHNICAL NGINEERING · eering, Peat and Organic Soils, Pile Foundations, PreservationofHistoricSites,ProfessionalPractice, Risk Assessment and Management,

7Mohamad H. Hussein Jerry A. DiMaggio

Vice PresidentGRL Engineers, Inc., Orlando, Florida

Partner, Pile Dynamics, Inc., Cleveland, Ohio

Senior Geotechnical EngineerFederal Highway Administration

Washington, DC

GEOTECHNICAL

ENGINEERING

In a broad sense, geotechnical engineering isthat branch of civil engineering that employsscientific methods to determine, evaluate,and apply the interrelationship between the

geologic environment and engineered works. In apractical context, geotechnical engineering encom-passes evaluation, design, and construction invol-ving earth materials.

The broad nature of this branch of civil engi-neering is demonstrated by the large number oftechnical committees comprising the Geo-Instituteof the American Society of Civil Engineers (ASCE).In addition, the International Society for Soil Mech-anics and Geotechnical Engineering (ISSMGE)includes the following 31 Technical Committees:Calcareous Sediments, Centrifuge and PhysicalModel Testing, Coastal Geotechnical Engineering,Deformation of Earth Materials, Earthquake Geo-technical Engineering, Education in GeotechnicalEngineering, Environmental Geotechnics, Frost,Geophysical Site Characterization, Geosyntheticsand Earth Reinforcement, Ground Improvement,Ground Property Characterization from In-situTesting, Indurated Soils and Soft Rocks, Instru-mentation forGeotechnicalMonitoring, Landslides,Limit State Design in Geotechnical Engineering,Micro-geomechanics,OffshoreGeotechnical Engin-eering, Peat and Organic Soils, Pile Foundations,Preservation of Historic Sites, Professional Practice,Risk Assessment and Management, Scour ofFoundations, Soil Sampling Evaluation and Inter-pretation, Stress-Strain Testing of Geomaterials in

the Laboratory, Tailings Dams, Tropical andResidual Soils, Underground Construction in SoftGround, Unsaturated Soils, and Validation ofComputer Simulations.

Unlike other civil engineering disciplines,which typically deal with materials whose proper-ties are well defined, geotechnical engineering isconcerned with subsurface materials whose prop-erties, in general, cannot be specified. Pioneers ofgeotechnical engineering relied on the “observa-tional approach” to develop an understanding ofsoil and rock mechanics and behavior of earthmaterials under loads. This approach was enhan-ced by the advent of electronic field instrumen-tation, wide availability of powerful personalcomputers, and development of sophisticatednumerical techniques. These now make it possibleto determine with greater accuracy the nonhomo-geneous, nonlinear, anisotropic nature and beha-vior of earthmaterials for application to engineeringworks.

Geotechnical engineers should be proficient inthe determination of soil and rock properties,engineering mechanics, subsurface investingationmethods and laboratory testing techniques. Theyshould have a thorough knowledge of designmethods, construction methods, monitoring/inspec-tion procedures, and specifications and contractingpractices. Geotechnical engineers should have broadpractical experience, in as much as the practice ofgeotechnical engineering involves art as much asscience. This requirement was clearly expressed by

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Source: Standard Handbook for Civil Engineers

Page 2: Senior Geotechnical Engineer GEOTECHNICAL NGINEERING · eering, Peat and Organic Soils, Pile Foundations, PreservationofHistoricSites,ProfessionalPractice, Risk Assessment and Management,

Karl Terzaghi, who made considerable contributionsto the development of soil mechanics: “The magni-tude of the difference between the performance ofreal soils under field conditions and the performancepredicted on the basis of theory can only beascertained by field experience.”

Geotechnical engineering is the engineeringscience of selecting, designing, and constructing fea-tures constructed of or upon soils and rock. Shallowfoundations, deep foundations, earth retainingstructures, soil and rock embankments and cuts areall specialty areas of geotechnical engineering.

Foundation engineering is the art of selecting,designing, and constructing for engineering worksstructural support systems based on scientific prin-ciples of soil and engineering mechanics and earth-structure interaction theories, and incorporatingaccumulated experience with such applications.

7.1 Lessons fromConstruction Claimsand Failures

Unanticipated subsurface conditions encounteredduring construction are by far the largest source ofconstruction-related claims for additional paymentby contractors and of cost overruns. Failures ofstructures as a result of foundation deficiencies canentail even greater costs, and moreover jeopardizepublic safety. A large body of experience hasidentified consistently recurring factors contribut-ing to these occurrences. It is important for theengineer to be aware of the causes of cost overruns,claims, and failures and to use these lessons to helpminimize similar future occurrences.

Unanticipated conditions (changed conditions)are the result of a variety of factors. The mostfrequent cause is the lack of definition of theconstituents of rock and soil deposits and theirvariation throughout the construction site. Relatedclaims are for unanticipated or excessive quantitiesof soil and rock excavation, misrepresentation ofthe quality and depth of bearing levels, unsuitableor insufficient on-site borrow materials, andunanticipated obstructions to pile driving or shaftdrilling. Misrepresentation of groundwater con-dition is another common contributor to workextras as well as to costly construction delays andemergency redesigns. Significant claims have alsobeen generated by the failure of geotechnical

investigations to identify natural hazards, such asswelling soils and rock minerals, unstable naturaland cut slopes, and old fill deposits.

Failures of structures during construction areusually related to undesirable subsurface con-ditions not detected before or during construction,faulty design, or poor quality ofwork. Examples arefoundations supported on expansive or collapsingsoils, on solutioned rock, or over undetected weakor compressible subsoils; foundation designs toodifficult to construct properly; foundations that donot perform as anticipated; and deficient construc-tion techniques or materials. Another importantdesign-related cause of failure is underestimationor lack of recognition of extreme loads associatedwith natural events, such as earthquakes, hurri-canes, floods, and prolonged precipitation. Relatedfailures include soil liquefaction during earth-quakes, hydrostatic uplift or water damage tostructures because of a rise in groundwater level,undermining of foundations by scour and over-topping, or wave erosion of earth dikes and dams.

It is unlikely that conditions leading to con-struction claims and failures can ever be completelyprecluded, inasmuchasdiscontinuities andextremevariation in subsurface conditions occur frequentlyin many types of soil deposits and rock formations.An equally important constraint that must beappreciated by both engineers and clients is thelimitations of the current state of geotechnicalengineering practice.

Mitigation of claims and failures, however, canbe achieved by fully integrated geotechnical inves-tigation, design, and construction quality assuranceconducted by especially qualified professionals. Inte-gration, rather than departmentalization of theseservices, ensures a continuity of purpose and philo-sophy that effectively reduces the risks associatedwith unanticipated subsurface conditions and designand construction deficiencies. It is also extremelyimportant thatownersandprimedesignprofessionalsrecognize that cost savings that reduce the quality ofgeotechnical services may purchase liabilities severalorders of magnitude greater than their initial“savings.”

7.2 Soil and RockClassifications

All soils are initially the products of chemicalalteration or mechanical disintegration of bedrock

7.2 n Section Seven

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GEOTECHNICAL ENGINEERING

Page 3: Senior Geotechnical Engineer GEOTECHNICAL NGINEERING · eering, Peat and Organic Soils, Pile Foundations, PreservationofHistoricSites,ProfessionalPractice, Risk Assessment and Management,

that has been exposed to weathering processes. Soilconstituents may have been subsequently modifiedby transportation processes such as water, wind,and ice and by inclusion and decomposition oforganic matter. Consequently, soil deposits maybe given a geologic as well as a constitutiveclassification.

Rock types are broadly classified by their modeof formation into igneous, metamorphic, and sedi-mentary deposits. The supporting ability (quality)assigned to rock for design or analysis shouldreflect the degree of alteration of the rock mineralsdue to weathering, the frequency of discontinuitieswithin the rock mass, and the susceptibility of therock to deterioration upon exposure.

7.2.1 Geologic Classification ofSoils

The classification of a soil deposit with respect to itsmode of deposition and geologic history is animportant step in understanding the variation insoil type and themaximum stresses imposed on thedeposit since deposition. (A geologic classificationthat identifies the mode of deposition of soildeposits is shown in Table 7.1.) The geologichistory of a soil deposit may also provide valuableinformation on the rate of deposition, the amountof erosion, and the tectonic forces that may haveacted on the deposit subsequent to deposition.

Geological and agronomic soil maps anddetailed reports are issued by the U.S. Departmentof Agriculture (www.usda.gov), U.S. GeologicalSurvey (www.usgs.gov), and corresponding stateoffices. Old surveys are useful for locating originalshore lines, stream courses, and surface-gradechanges.

7.2.2 Unified Soil ClassificationSystem

This is the most widely used of the variousconstitutive classification systems and correlatessoil type with generalized soil behavior. Allsoils are classified as coarse-grained (50% of theparticles .0.074 mm), fine-grained (50% of theparticles ,0.074 mm), or predominantly organic(see Table 7.2).

Coarse-grained soils are categorized by theirparticle size into boulders (particles larger than 8 in),

cobbles (3 to 8 in), gravel, and sand. For sands (S)and gravels (G), grain-size distribution is identifiedas either poorly graded (P) or well-graded (W), asindicated by the group symbol in Table 7.2. Thepresence of fine-grained soil fractions (under 50%),such as silt and clay, is indicated by the symbols Mand C, respectively. Sands may also be classifiedas coarse (larger than No. 10 sieve), medium(smaller than No. 10 but larger than No. 40), or fine

Table 7.1 Geologic Classification of Soil Deposits

Classification Mode of Formation

AeolianDune Wind deposition (coastal

and desert)Loess Deposition during glacial

periodsAlluvialAlluvium River and stream depositionLacustrine Lake waters, including glacial

lakesFloodplain Floodwaters

ColluvialColluvium Downslope soil movementTalus Downslope movement of rock

debrisGlacialGroundmoraine

Deposited and consolidated byglaciers

Terminalmoraine

Scour and transport at ice front

Outwash Glacier melt watersMarineBeach orbar

Wave deposition

Estuarine River estuary depositionLagoonal Deposition in lagoonsSalt marsh Deposition in sheltered tidal

zonesResidualResidualsoil

Complete alteration by in situweathering

Saprolite Incomplete but intensealteration and leaching

Laterite Complex alteration in tropicalenvironment

Decomposedrock

Advanced alteration withinparent rock

Geotechnical Engineering n 7.3

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Page 4: Senior Geotechnical Engineer GEOTECHNICAL NGINEERING · eering, Peat and Organic Soils, Pile Foundations, PreservationofHistoricSites,ProfessionalPractice, Risk Assessment and Management,

Table

7.2

Unified

SoilClassificationIncludingIden

tificationan

dDescriptiona

MajorDivision

Group

Symbol

Typical

Nam

eField

Iden

tification

Proceduresb

Lab

oratory

ClassificationCriteriac

A.Coarse-grained

soils(m

ore

than

halfofmateriallarger

than

No.200siev

e)d

1.Gravels(m

ore

than

halfofcoarse

fractionlarger

than

No.4siev

e)e

Clean

gravels(little

ornofines)

GW

Well-graded

gravels,

gravel-san

dmixtures,

little

ornofines

Widerangein

grain

sizes

andsu

bstan

tial

amounts

ofallinterm

ediate

particlesizes

D60/D

10.

41,

D30/D

10D

60,

3D

10,D

30,D

60¼

sizescorrespondingto

10,30,an

d60%

ongrain-sizecu

rve

GP

Poorlygraded

gravelsor

gravel-san

dmixtures,

little

ornofines

Predominan

tlyonesize,ora

rangeofsizeswithsome

interm

ediate

sizesmissing

Notmeetingallgradationrequirem

ents

forGW

Gravelswithfines

(appreciab

leam

ountoffines)

GM

Silty

gravels,gravel-

sand-siltmixtures

Nonplastic

fines

orfines

with

low

plasticity(see

MLsoils)

Atterberglimits

below

AlineorPI,

4

Soilsab

oveA

linewith

4,

PI,

7areborderline

cases,requireuse

of

dual

symbols

GC

Clayey

gravels,gravel-

sand-claymixtures

Plastic

fines

(see

CLsoils)

Atterberg

limitsab

ove

AlinewithPI.

7

2.San

ds(m

ore

than

halfofcoarse

fractionsm

allerthan

No.4siev

e)e

Clean

sands(little

ornofines)

SW

Well-graded

sands,

gravelly

sands,little

ornofines

Widerangein

grain

sizesan

dsu

bstan

tial

amounts

ofall

interm

ediate

particlesizes

D60/D

10.

61,

D30/D

10D

60,

3

SP

Poorlygraded

sandsor

gravel-san

ds,little

ornofines

Predominan

tlyonesize,ora

rangeofsizeswithsome

interm

ediate

sizesmissing

Notmeetingallgradationrequirem

ents

forSW

San

dswithfines

(appreciab

leam

ountoffines)

SM

Silty

sands,sand-silt

mixtures

Nonplastic

fines

orfines

with

low

plasticity(see

MLsoils)

Atterberg

limitsbelow

AlineorPI,

4

SoilswithAtterberglimits

aboveA

linewhile

4,

PI,

7areborderline

SC

Clayey

sands,sand-clay

mixtures

Plastic

fines

(see

CLsoils)

Atterberglimits

aboveA

line

withPI.

7

cases;requireuse

ofdual

symbols

7.4 n Section Seven

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Inform

ationrequired

fordescribingcoarse-grained

soils:

Forundisturbed

soils,ad

dinform

ationonstratification,deg

reeofcompactness,cemen

tation,moisture

conditions,an

ddrainag

ech

aracteristics.Givetypical

nam

e;indicateap

proxim

atepercentageofsandan

dgravel;m

axim

um

size,a

ngularity,surfacecondition,a

ndhardnessofthecoarse

grains;localorgeo

logical

nam

ean

dother

pertinen

tdescriptiveinform

ation;an

dsymbolin

paren

theses.Exam

ple:Salty

sand,

gravelly;ab

out20%

hard,an

gulargravel

particles,1 ⁄ 2

inmax

imum

size;rounded

andsu

ban

gularsandgrains,

coarse

tofine;

about15%

nonplastic

fines

withlow

dry

strength;wellcompactedan

dmoistin

place;

alluvialsand;(SM).

B.Fine-grained

soils(m

ore

than

halfofmateriallarger

than

No.200siev

e)d

Iden

tificationProceduresf

MajorDivision

Group

Symbol

Typical

Nam

eDry

Stren

gth

(Crush

ing

Characteristics)

Dilatan

cy(Reactionto

Shak

ing)

Toughness

(Consisten

cyNearPL)

Lab

oratory

Classification

Criteriac

Silts

andclay

swithliquid

limitless

than

50

ML

Inorgan

icsiltsan

dveryfinesands,rock

flour,siltyorclay

eyfinesands,orclay

eysiltswithslight

plasticity

Noneto

slight

Quickto

slow

None

CL

Inorgan

icclay

soflow

tomed

ium

plasticity,

gravelly

clay

s,sandy

clay

s,siltyclay

s,lean

clay

s

Med

ium

tohigh

Noneto

veryslow

Med

ium

OL

Organ

icsiltsan

dorgan

icsilty

clay

soflow

plasticity

Slightto

med

ium

Slow

Slight

Silts

andclay

swithliquid

limitmore

than

50

MH

Inorgan

icsilts,

micaceo

usor

diatomaceo

usfine

sandyorsiltysoils,

elasticsilts

Slightto

med

ium

Slow

tonone

Slightto

med

ium

Plasticitych

artlaboratory

classificationsoffine-grained

soilscompares

them

ateq

ual

liquid

limit.Toughnessan

ddry

strength

increase

withincreasing

plasticityindex

(PI)

CH

Inorgan

icclay

sofhigh

plasticity,fatclay

sNoneto

veryhigh

None

High

CH

Organ

icclay

sofmed

ium

tohighplasticity

Med

ium

tohigh

Noneto

veryslow

Slightto

med

ium

(Tablecontinued)

Geotechnical Engineering n 7.5

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Page 6: Senior Geotechnical Engineer GEOTECHNICAL NGINEERING · eering, Peat and Organic Soils, Pile Foundations, PreservationofHistoricSites,ProfessionalPractice, Risk Assessment and Management,

Table

7.2

(Continued)

C.Highly

organ

icsoils

Pt

Peatan

dother

highly

organ

icsoils

Readilyiden

tified

bycolor,odor,sp

ongyfeel,

andoften

byfibroustexture

Field

iden

tificationproceduresforfine-grained

soilsorfractions:g

Dilatan

cy(Reactionto

Shak

ing)

Dry

Stren

gth

(Crush

ingCharacteristics)

Toughness(C

onsisten

cyNearPL)

After

removingparticles

larger

than

No.40

siev

e,prepareapat

ofmoistsoilwitha

volumeofab

out1 ⁄ 2

in3.Adden

oughwater

ifnecessary

tomak

ethesoilsoftbutnotsticky.

Place

thepat

intheopen

palm

ofonehan

dan

dsh

akehorizo

ntally,strikingvigorously

against

theother

han

dseveral

times.A

positivereactionconsistsoftheap

pearance

of

water

onthesu

rfaceofthepat,w

hichch

anges

toaliveryconsisten

cyan

dbecomes

glossy.

When

thesample

issqueezedbetweenthe

fingers,thewater

andgloss

disap

pearfrom

thesu

rface,

thepat

stiffens,an

dfinally

itcracksorcrumbles.Therapidityofap

pear-

ance

ofwater

duringsh

akingan

dofits

disap

pearance

duringsqueezingassist

iniden

tifyingthech

aracterofthefines

inasoil.

Veryfinecleansandsgivethequickestan

dmost

distinct

reaction,whereasaplastic

clay

has

noreaction.Inorgan

icsilts,su

chas

atypical

rock

flour,sh

ow

amoderatelyquick

reaction.

After

removingparticles

larger

than

No.40

siev

e,mold

apat

ofsoilto

theconsisten

cyof

putty,ad

dingwater

ifnecessary.A

llow

thepat

todry

completely

byoven

,su

n,orairdrying,

then

test

itsstrength

bybreak

ingan

dcrumblingbetweenthefingers.This

strength

isameasu

reofch

aracteran

dquan

tity

ofthe

colloidalfractioncontained

inthesoil.T

hedry

strength

increaseswithincreasingplasticity.

Highdry

strength

isch

aracteristic

ofclay

softheCH

group.A

typical

inorgan

icsilt

possessesonly

veryslightdry

strength.Silty

finesandsan

dsiltshav

eab

outthesameslight

dry

strength

butcanbedistinguished

bythe

feelwhen

powderingthedried

specim

en.Fine

sandfeelsgritty,whereasatypical

silthas

the

smooth

feel

offlour.

After

particles

larger

than

theNo.40siev

eareremoved

,asp

ecim

enofsoilab

out1 ⁄ 2

in3in

size

ismolded

tothe

consisten

cyofputty.Ifitistoodry,water

must

bead

ded

.Ifitis

toosticky,thesp

ecim

ensh

ould

besp

read

outin

athin

layer

andallowed

tolose

somemoisture

by

evap

oration.T

hen

,thesp

ecim

enisrolled

outbyhan

dona

smooth

surfaceorbetweenthepalmsinto

athread

about

1 ⁄ 8in

indiameter.Thethread

isthen

folded

andrerolled

repeatedly.Duringthis

man

ipulation,themoisture

contentis

gradually

reducedan

dthesp

ecim

enstiffens,

finally

losesitsplasticity,an

dcrumbleswhen

theplastic

limit(PL)is

reached

.After

thethread

crumbles,thepiecessh

ould

belumped

together

andaslightkneadingactioncontinued

untilthe

lumpcrumbles.

Thetougher

thethread

nearthePLan

dthestifferthe

lumpwhen

itfinally

crumbles,themore

potentis

the

colloidal

clay

fractionin

thesoil.W

eaknessofthethread

atthePLan

dquickloss

ofcoheren

ceofthelumpbelow

the

PLindicateeither

organ

icclay

oflow

plasticityor

materialssu

chas

kao

lin-typeclay

san

dorgan

icclay

sthat

occurbelow

theA

line.

Highly

organ

icclay

shav

eaveryweakan

dsp

ongyfeel

atPL.

7.6 n Section Seven

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Inform

ationrequired

fordescribingfine-grained

soils:

Forundisturbed

soils,ad

dinform

ationonstructure,stratification,consisten

cyin

undisturbed

andremolded

states,moisture,an

ddrainag

econditions.Give

typical

nam

e;indicatedeg

reean

dch

aracterofplasticity;am

ountan

dmax

imum

size

ofcoarse

grains;colorin

wet

conditions,odor,ifan

y;localorgeo

logical

nam

ean

dother

pertinen

tdescriptiveinform

ation;an

dsymbolin

paren

theses.Exam

ple:Clayey

silt,brown;slightlyplastic;sm

allpercentageoffinesand;

numerousverticalrootholes;firm

anddry

inplace;loess;(M

L)

aAdap

tedfrom

recommen

dationsofU.S.Arm

yCorpsofEngineers

andU.S.Bureau

ofReclamation.Allsiev

esizesUnited

Statesstan

dard.

bExcludingparticles

larger

than

3in

andbasingfractionsonestimated

weights.

c Use

grain-sizecu

rvein

iden

tifyingthefractionsas

given

under

fieldiden

tification.

Forcoarse-grained

soils,determinepercentageofgravel

andsandfrom

grain-sizecu

rve.Dep

endingonpercentageoffines

(fractionssm

allerthan

No.200

siev

e),coarse-grained

soils

areclassified

asfollows:

Lessthan

5%fines:

GW,GP,

SW,SP

More

than

12%

fines:

GM,GC,SM,SC

5%to

12%

fines:

Bord

erlinecasesrequiringuse

ofdual

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aracteristicsoftw

ogroupsaredesignated

bycobinationsofgroupsymbols;forexam

ple,GW-G

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awell-graded

,gravel-san

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dTheNo.200siev

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Geotechnical Engineering n 7.7

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(smaller than No. 40). Because properties of thesesoils are usually significantly influenced by relativedensity Dr , rating of the in situ density and Dr is animportant consideration (see Art. 7.4).

Fine-grained soils are classified by their liquidlimit and plasticity index as organic clays OHor silts OL, inorganic clays CH or CL, or silts orsandy silts MH or ML, as shown in Table 7.2. Forthe silts and organic soils, the symbols H and Ldenote a high and low potential compressibilityrating; for clays, they denote a high and lowplasticity. Typically, the consistency of cohesivesoils is classified from pocket penetrometer orTorvane tests on soil samples. These index testsare convenient for relative comparisons but donot provide design strength values and shouldnot be used as property values for design oranalysis. The consistency ratings are expressed asfollows:

Soft—under 0.25 tons/ft2

Firm—0.25 to 0.50 tons/ft2

Stiff—0.50 to 1.0 tons/ft2

Very stiff—1.0 to 2.0 tons/ft2

Hard—more than 2.0 tons/ft2

7.2.3 Rock Classification

Rock, obtained from core samples, is commonlycharacterized by its type, degree of alteration(weathering), and continuity of the core. (Whereoutcrop observations are possible, rock structuremay be mapped.) Rock-quality classificationsare typically based on the results of compressivestrength tests or the condition of the core samples,or both. Rock types typical of igneous depositsinclude basalt, granite, diorite, rhyolite, and an-desite. Typical metamorphic rocks include schist,gneiss, quartzite, slate, andmarble. Rocks typical ofsedimentary deposits include shale, sandstone,conglomerate, and limestone.

Rock structure and degree of fracturing usuallycontrol the behavior of a rock mass that has beensignificantly altered by weathering processes. It isnecessary to characterize both regional and localstructural features that may influence design offoundations, excavations, and underground open-ings in rock. Information from geologic publi-cations and maps are useful for defining regional

trends relative to the orientation of bedding, majorjoint systems, faults, and so on.

Rock-quality indices determined from inspec-tion of rock cores include the fracture frequency(FF) and rock-quality designation (RQD). FF is thenumber of naturally occurring fractures per foot ofcore run, whereas RQD is the cumulative length ofnaturally separated core pieces, 4 in or more indimension, expressed as a percentage of the lengthof core run. The rock-quality rating also may bebased on the velocity index obtained fromlaboratory and in situ seismic-wave-propagationtests. The velocity index is given by (Vs/Vl)

2, whereVs and Vl represent seismic-wave velocities from insitu and laboratory core measurements, respect-ively. Proposed RQD and velocity index rock-quality classifications and in situ deformabilitycorrelations are in Table 7.3. A relative-strengthrating of the quality rock cores representative of theintact elements of the rock mass, proposed byDeere and Miller, is based on the uniaxialcompressive (UC) strength of the core and itstangent modulus at one-half of the UC.

(D. U. Deere and R. P. Miller, “Classification andIndex Properties for Intact Rock,” Technical ReportAFWL-TR-65-116, Airforce Special Weapons Cen-ter, Kirtland Airforce Base, New Mexico, 1966.)

Inasmuch as some rocks tend to disintegraterapidly (slake) upon exposure to the atmosphere,the potential for slaking should be rated fromlaboratory tests. These tests include emersion inwater, Los Angeles abrasion, repeated wettingand drying, and other special tests, such as a

Table 7.3 Rock-Quality Classification andDeformability Correlation

Classification RQDVelocityIndex

DeformabilityEd/Et*

Very poor 0–25 0–0.20 Under 0.20Poor 25–50 0.20–0.40 Under 0.20Fair 50–75 0.40–0.60 0.20–0.50Good 75–90 0.60–0.80 0.50–0.80Excellent 90–100 0.80–1.00 0.80–1.00

*Ed ¼ in situ deformation modulus of rock mass; Et ¼ tangentmodulus at 50% of UC strength of core specimens.

Source: Deere, Patton and Cording, “Breakage of Rock,”Proceedings, 8th Symposium on Rock Mechanics, AmericanInstitute of Mining and Metallurgical Engineers, Minneapolis,Minn.

7.8 n Section Seven

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slaking-durability test. Alteration of rock mineralsdue to weathering processes is often associatedwith reduction in rock hardness and increase inporosity and discoloration. In an advanced stage ofweathering, the rock may contain soil-like seams,be easily abraded (friable), readily broken, andmay(but will not necessarily) exhibit a reduced RQD orFF. Rating of the degree of rock alteration whenlogging core specimens is a valuable aid inassessing rock quality.

7.3 PhysicalPropertiesofSoils

Basic soil properties and parameters can be sub-divided into physical, index, and engineering cate-gories. Physical soil properties include density,particle size and distribution, specific gravity, andwater content.

The water content w of a soil sample representsthe weight of free water contained in the sampleexpressed as a percentage of its dry weight.

The degree of saturation S of the sample is theratio, expressed as percentage, of the volume of freewater contained in a sample to its total volume ofvoids Vv .

Porosity n, which is a measure of the relativeamount of voids, is the ratio of void volume to thetotal volume V of soil:

n ¼ Vv

V(7:1)

The ratio of Vv to the volume occupied by the soilparticles Vs defines the void ratio e. Given e, thedegree of saturation may be computed from

S ¼ wGs

e(7:2)

where Gs represents the specific gravity of the soilparticles. For most inorganic soils, Gs is usually inthe range of 2.67+0.05.

The dry unit weight gd of a soil specimen withany degree of saturation may be calculated from

gd ¼gwGsS

1þ wGs(7:3)

where gw is the unit weight of water and is usuallytaken as 62.4 lb/ft3 for fresh water and 64.0 lb/ft3

for seawater.

The particle-size distribution (gradation) of soilscan be determined by mechanical (sieve) analysisand combined with hydrometer analysis if thesample contains a significant amount of particlesfiner than 0.074 mm (No. 200 sieve). The soilparticle gradation in combination with the maxi-mum, minimum, and in situ density of cohesion-less soils can provide useful correlations withengineering properties (see Arts. 7.4 and 7.52).

7.4 Index Parameters for Soils

Index parameters of cohesive soils include liquidlimit, plastic limit, shrinkage limits, and activity.Such parameters are useful for classifying cohesivesoils and providing correlations with engineeringsoil properties.

The liquid limit of cohesive soils represents anear liquid state, that is, an undrained shearstrength about 0.01 lb/ft2. The water content atwhich the soil ceases to exhibit plastic behavior istermed the plastic limit. The shrinkage limitrepresents the water content at which no furthervolume change occurs with a reduction in watercontent. The most useful classification and corre-lation parameters are the plasticity index Ip, theliquidity index Il, the shrinkage index Is, and theactivity Ac. These parameters are defined inTable 7.4.

Relative density Dr of cohesionless soils maybe expressed in terms of void ratio e or unit dryweight gd:

Dr ¼ emax � eoemax � emin

(7:4a)

Dr ¼ 1=gmin � 1=gd1=gmin � 1=gmax

(7:4b)

Dr provides cohesionless soil property and param-eter correlations, including friction angle, permea-bility, compressibility, small-strain shear modulus,cyclic shear strength, and so on.

In situ field tests such as the Standard Pen-etration Test (SPT), static cone penetrometer,pressuremeter, and dilatometer can also be usedto determine the index properties of cohensionlessand cohensive soils.

(H. Y. Fang, “Foundation Engineering Hand-book,” 2nd ed., VanNostrandReinhold,NewYork.)

Geotechnical Engineering n 7.9

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7.5 Engineering Properties ofSoils

Engineering soil properties and parameters des-cribe the behavior of soil under induced stress andenvironmental changes. Of interest to mostgeotechnical applications are the strength, deform-ability, and permeability of in situ and compactedsoils. ASTM promulgates standard test proceduresfor soil properties and parameters.

7.5.1 Shear Strength of CohesiveSoils

The undrained shear strength cu, of cohesive soilsunder static loading can be determined by sev-eral types of laboratory tests, including uniaxialcompression, triaxial compression (TC) or exten-sion (TE), simple shear, direct shear, and torsionshear. The objective of soil laboratory testing is toreplicate the field stress, loading and drainageconditions with regard to magnitude, rate andorientation. All laboratory strength testing requireextreme care in securing, transporting and prepar-ing the test sample. The triaxial test is the mostversatile but yet the most complex strength test toperform. Triaxial tests involve application of acontrolled confining pressure s3 and axial stress s1

to a soil specimen. s3 may be held constant and s1

increased to failure (TC tests), or s1 may be heldconstant while s3 is decreased to failure (TE tests).Specimens may be sheared in a drained or un-drained condition.

The unconsolidated-undrained (UU) triaxialcompression test is appropriate and commonly usedfor determining the cu of relatively good-quality

samples. For soils that do not exhibit changes in soilstructure under elevated consolidation pressures,consolidated-undrained (CU) tests following theSHANSEP testing approach mitigate the effects ofsample disturbance.

(C. C. Ladd and R. Foott, “New Design Proce-dures for Stability of Soft Clays,” ASCE Journal ofGeotechnical Engineering Division, vol. 99, no. GT7,1974, www.asce.org.)

For cohesive soils exhibiting a normal claybehavior, a relationship between the normalizedundrained shear strength cu=s

0vo and the over-

consolidation ratio OCR can be defined indepen-dently of the water content of the test specimen by

cus 0vo

¼ K(OCR)n (7:5)

where cu is normalized by the preshear verticaleffective stress, the effective overburden pressures 0vo, or the consolidation pressure s 0

lc triaxial testconditions. OCR is the ratio of preconsolidationpressure to overburden pressure. The parameter Krepresents the cu=s

0vo of the soil in a normally

consolidated state, and n primarily depends on thetype of shear test. For CU triaxial compressiontests, K is approximately 0.32+0.02 and is lowestfor low plasticity soils and is a maximum for soilswith plasticity index Ip over 40%. The exponent n isusually within the range of 0.70+0.05 and tends tobe highest for OCR less than about 4.

In situ vane shear tests also are often used toprovide cu measurements in soft to firm clays. Testsare commonly made on both the undisturbed andremolded soil to investigate the sensitivity, theratio of the undisturbed to remolded soil strength.This test is not applicable in sand or silts or where

Table 7.4 Soil Indices

Index Definition* Correlation

Plasticity Ip ¼ Wl �Wp Strength, compressibility, compactibility, and so forth

Liquidity Il ¼Wn �Wp

IpCompressibility and stress rate

Shrinkage Is ¼ Wp �Ws Shrinkage potential

Activity Ac ¼Ip

mSwell potential, and so on

*Wl ¼ liquid limit; Wp ¼ plastic limit; Wn ¼ moisture content, %; Ws ¼ shrinkage limit; m ¼ percent of soil finer than 0.002mm(clay size).

7.10 n Section Seven

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hard inclusions (nodules, shell, gravel, and soforth) may be present. (See also Art. 7.6.3.)

The Standard Penetration Test, cone penetro-meter, pressuremeter, and dilatometer provideguidance on engineering properties of soils. Similarto laboratory tests ASTM procedures have beendeveloped for several of these tests. Each test,similar to the vane shear test, has advantages anddisadvantages, as well as being limited to certainsoil types.

Drained shear strength of cohesive soils isimportant in design and control of constructionembankments on soft ground as well as in otherevaluations involving effective-stress analyses.Conventionally, drained shear strength tf isexpressed by the Mohr-Coulomb failure criteria as:

tf ¼ c 0 þ s 0n tanf

0 (7:6)

The c0 and f0 parameters represent the effectivecohesion and effective friction angle of the soil,respectively. s 0

n is the effective stress normal to theplane of shear failure and can be expressed in termsof total stress sn as (sn 2 ue), where ue is the excesspore-water pressure at failure. ue is induced bychanges in the principal stresses (Ds1, Ds3). Forsaturated soils, it is expressed in terms of the pore-water-pressure parameter Af at failure as:

u ¼ Ds3 þ Af (Ds1 � Ds3)f (7:7)

The effective-stress parameters c0, f0, and Af arereadily determined by CU triaxial shear testsemploying pore-water-pressure measurements or,excepting Af, by consolidated-drained (CD) tests.

After large movements along preformed failureplanes, cohesive soils exhibit a significantly reduced(residual) shear strength. The corresponding effec-tive friction angle f0

r is dependent on Ip. For manycohesive soils, f0

r is also a function of s 0n. The f0

r

parameter is applied in analysis of the stabilityof soils where prior movements (slides) haveoccurred.

Cyclic loading with complete stress reversalsdecreases the shearing resistance of saturatedcohesive soils by inducing a progressive buildupin pore-water pressure. The amount of degradationdepends primarily on the intensity of the cyclicshear stress, the number of load cycles, the stresshistory of the soil, and the type of cyclic test used.The strength degradation potential can be deter-mined by postcyclic, UU tests.

7.5.2 Shear Strength ofCohesionless Soils

The shear strength of cohesionless soils under staticloading can be interpreted from results of drainedor undrained TC tests incorporating pore-pressuremeasurements. The effective angle of internalfriction f0 can also be expressed by Eq. (7.6), exceptthat c0 is usually interpreted as zero. For cohesion-less soils, f0 is dependent on density or void ratio,gradation, grain shape, and grain mineralogy.Markedly stress-dependent, f0 decreases with in-creasing s 0

n, the effective stress normal to the planeof shear failure.

In situ cone penetration tests in sands may beused to estimate f0 from cone resistance qc records.One approach relates the limiting qc values directlytof0.Where qc increases approximately linearlywithdepth,f0 can also be interpreted from the slope of thecurve for qc 2 svo vs. f

0vo, where svo ¼ total vertical

stress, s 0vo ¼ svo � u, and u ¼ pore-water pressure.

The third approach is to interpret the relative densityDr from qc and then relate f0 to Dr as a function ofthe gradation and grain shape of the sand.

Relative density provides good correlation withf0 for a given gradation, grain shape, and normalstress range. Awidely used correlation is shown inFig. 7.1. Dr can be interpreted from standardpenetration resistance tests (Fig. 7.12) and conepenetration resistance tests (see Arts. 7.6.2 and7.6.3) or calculated from the results of in situ ormaximum and minimum density tests. The mostdifficult property to determine in the relativedensity equation is e0, the in-situ void ratio.

Dense sands typically exhibit a reduction inshearing resistance at strains greater than thoserequired to develop the peak resistance. At relativelylarge strains, the stress-strain curves of loose anddense sands converge.Thevoid ratio atwhich there isno volume change during shear is called the criticalvoid ratio. A volume increase during shear (dila-tancy) of saturated, dense, cohesionless soils pro-ducesnegativepore-waterpressuresandatemporaryincrease in shearing resistance. Subsequent dissipa-tion of negative pore-water pressure accounts for the“relaxation effect” sometimes observed after pileshave been driven into dense, fine sands.

Saturated, cohesionless soils subject to cyclicloads exhibit a significant reduction in strength ifcyclic loading is applied at periods smaller than thetime required to achieve significant dissipation ofpore pressure. Should the number of load cyclesNc

Geotechnical Engineering n 7.11

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be sufficient to generate pore pressures that ap-proach the confining pressure within a soil zone,excessive deformations and eventually failure(liquefaction) is induced. For a given confiningpressure and cyclic stress level, the number ofcycles required to induce initial liquefaction Nc1

increases with an increase in relative density Dr .Cyclic shear strength is commonly investigated bycyclic triaxial tests and occasionally by cyclic,direct, simple-shear tests.

7.5.3 State of Stress of Soils

Assessment of the vertical s 0vo and horizontal s 0

ho

effective stresses within a soil deposit and themaximum effective stresses imposed on thedeposit since deposition s 0

vm is a general require-ment for characterization of soil behavior. The ratios 0vm=s

0vo is termed the overconsolidation ratio

(OCR). Another useful parameter is the ratio ofs 0ho=s

0vo, which is called the coefficient of earth

pressure at rest (Ko).For a simple gravitation piezometric profile, the

effective overburden stress s 0vo is directly related

to the depth of groundwater below the surfaceand the effective unit weight of the soil strata.

Groundwater conditions, however, may be charac-terized by irregular piezometric profiles thatcannot be modeled by a simple gravitationalsystem. For these conditions, sealed piezometermeasurements are required to assess s 0

vo.

Maximum Past Consolidation Stress n

The maximum past consolidation stress s 0vm of a

soil deposit may reflect stresses imposed prior togeologic erosion or during periods of significantlylower groundwater, as well as desiccation effectsand effects of human activity (excavations). Themaximum past consolidation stress is convention-ally interpreted from consolidation (oedometer)tests on undisturbed samples.

Normalized-shear-strength concepts provide analternate method for estimating OCR from good-quality UU compression tests. In the absence ofsite-specific data relating cu=s

0vo andOCR, a form of

Eq. (7.5) may be applied to estimate OCR. In thisinterpretation, s 0

vo represents the effective over-burden pressure at the depth of the UU-testsample. A very approximate estimate of s 0

vm canalso be obtained for cohesive soils from relation-ships proposed between liquidity index andeffective vertical stress (“Design Manual—Soil

Fig. 7.1 Chart for determining friction angles for sands. (After J. H. Schmertmann.)

7.12 n Section Seven

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Mechanics, Foundations, and Earth Structures,”NAVDOCKS DM-7, U.S. Navy). For coarse-grainedsoil deposits, it is difficult to characterize s 0

vm

reliably from either in situ or laboratory testsbecause of an extreme sensitivity to disturbance.

The coefficient of earth pressure at rest Ko can bedetermined in the laboratory from “no-lateralstrain” TC tests on undisturbed soil samples orfrom consolidation tests conducted in speciallyconstructed oedometers. Interpretation of Ko fromin situ CPT, PMT, and dilatometer tests has alsobeen proposed. In view of the significant impact ofsample disturbance on laboratory results and theempirical nature of in situ test interpretations, thefollowing correlations of Ko with friction angle f0

and OCR are useful. For both coarse- and fine-grained soils:

Ko ¼ (1� sinf0)OCRm (7:8)

A value for m of 0.5 has been proposed foroverconsolidated cohesionless soils, whereas forcohesive soils it is proposed that m be estimated interms of the plasticity index Ip as 0:581I

�0:12p .

7.5.4 Deformability of Fine-GrainedSoils

Deformations of fine-grained soils can be classifiedas those that result from volume change, (elastic)distortion without volume change, or a combi-nation of these causes. Volume change may be aone-dimensional or, in the presence of imposedshear stresses, a three-dimensional mechanismand may occur immediately or be time-dependent.Immediate deformations are realized without vol-ume change during undrained loading of saturatedsoils and as a reduction of air voids (volumechange) within unsaturated soils.

The rate of volume change of saturated, fine-grained soils during loading or unloading iscontrolled by the rate of pore-fluid drainage fromor into the stressed soil zone. The compressionphase of delayed volume change associated withpore-pressure dissipation under a constant load istermed primary consolidation. Upon completionof primary consolidation, some soils (particularlythose with a significant organic content) continueto decrease in volume at a decreasing rate. Thisresponse is usually approximated as a straight linefor a plot of log time vs. compression and is termedsecondary compression.

As the imposed shear stresses become asubstantial fraction of the undrained shear strengthof the soil, time-dependent deformations mayoccur under constant load and volume conditions.This phenomenon is termed creep deformation.Failure by creep may occur if safety factors areinsufficient to maintain imposed shear stressesbelow the creep threshold of the soil. (Also seeArt. 7.10.)

One-dimensional volume-change parametersare conveniently interpreted from consolidation(oedometer) tests. A typical curve for log conso-lidation pressure vs. volumetric strain 1v (Fig. 7.2)demonstrates interpretation of the strain-refer-enced compression index C0

c, recompression indexC0r, and swelling index C0

s. The secondary com-pression index C0

a represents the slope of the near-linear portion of the volumetric strain vs. log-timecurve following primary consolidation (Fig. 7.2b).The parameters C0

c, C0r, and C0

a be roughly estimatedfrom soil-index properties.

Deformation moduli representing three-dimen-sional deformation can be interpreted from thestress-strain curves of laboratory shear tests forapplication to either volume change or elasticdeformation problems.

(“Design Manual—Soil Mechanics, Founda-tions, and Earth Structures,” NAVDOCKS DM-7,U.S. Navy; T. W. Lambe and R. V. Whitman, “SoilMechanics,” John Wiley & Sons, Inc., New York,www.wiley.com.)

7.5.5 Deformability of Coarse-Grained Soils

Deformation of most coarse-grained soils occursalmost exclusively by volume change at a rateessentially equivalent to the rate of stress change.Deformation moduli are markedly nonlinear withrespect to stress change and dependent on theinitial state of soil stress. Some coarse-grained soilsexhibit a delayed volume-change phenomenonknown as friction lag. This response is analogousto the secondary compression of fine-grained soilsand can account for a significant amount of thecompression of coarse-grained soils composed ofweak or sharp-grained particles.

The laboratory approach previously describedfor derivation of drained deformation parametersfor fine-grained soils has a limited application forcoarse-grained soils because of the difficulty in

Geotechnical Engineering n 7.13

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obtaining reasonably undisturbed samples. Testsmay be carried out on reinstituted samples butshould be used with caution since the soil fabric,aging, and stress history cannot be adequatelysimulated in the laboratory. As a consequence, insitu testing techniques are often the preferredinvestigation and testing approach to the charac-terization of cohesionless soil properties (see Art.7.6.3).

The Static ConePenetration Test (CPT) n

The CPT is one of the most useful in situ tests forinvestigating the deformability of cohesionless soils.The secant modulus E0

s, tons/ft2, of sands has been

related to cone resistance qc by correlations of small-scale plate load tests and load tests on footings. Therelationship is given by Eq. (7.9a). The empiricalcorrelation coefficient a in Eq. (7.9a) is influenced bythe relative density, grain characteristics, and stresshistory of the soil (see Art. 7.6.3).

E0s ¼ aqc (7:9a)

The a parameter has been reported to range be-tween 1.5 and 3 for sands and can be expressed interms of relative density Dr as 2(1þD2

r ). amay alsobe derived from correlations between qc andstandard penetration resistance N by assuming thatqc/N for mechanical cones or qc/N þ 1 for electronictype cone tips is about 6 for sandy gravel, 5 forgravelly sand, 4 for clean sand, and 3 for sandy silt.However, it should be recognized that E0

s character-izations from qc or N are empirical and can provideerroneous characterizations. Therefore, the validityof these relationships should be confirmed by localcorrelations.Conepenetration test soundings shouldbe conducted in accordance with ASTM D-3441(Briaud, J. L. and Miran J. (1992). “The ConePenetrometer Test”, Federal Highway Adminis-tration, FHWA Report No. SA-91-043, WashingtonD.C; See also Art. 7.13)

Load-Bearing Test n One of the earliestmethods for evaluating the in situ deformability

Fig. 7.2 Typical curves plotted from data obtained in consolidation tests.

7.14 n Section Seven

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of coarse-grained soils is the small-scale load-bearing test. Data developed from these tests havebeen used to provide a scaling factor to express thesettlement r of a full-size footing from the settle-ment r1 of a 1-ft

2 plate. This factor r/r1 is given as afunction of the width B of the full-size bearing plateas:

r

r1¼ 2B

1þ B

� �2

(7:10)

From an elastic half-space solution, E0s can be

expressed from results of a plate load test in termsof the ratio of bearing pressure to plate settlementkv as:

E0s ¼

kv(1� m2)p=4

4B=(1þ B)2(7:9b)

m represents Poisson’s ratio, usually considered torange between 0.30 and 0.40. Equation (7.9b) assu-mes that r1 is derived from a rigid, 1-ft-diametercircular plate and that B is the equivalent diameterof the bearing area of a full-scale footing. Empiricalformulations such as Eq. (7.10) may be significantlyin error because of the limited footing-size rangeused and the large scatter of the data base. Further-more, consideration is not given to variations inthe characteristics and stress history of the bearingsoils.

Pressuremeter tests (PMTs) in soils and softrocks have been used to characterize E0

s fromradial pressure vs. volume-change data devel-oped by expanding a cylindrical probe in a drillhole (see Art. 7.6.3). Because cohesionless soils aresensitive to comparatively small degrees of soildisturbance, proper access-hole preparation iscritical.

(K. Terzaghi and R. B. Peck, “Soil Mechanics andEngineering Practice,” John Wiley & Sons, Inc.,New York; H. Y. Fang, “Foundation EngineeringHandbook,” 2nd ed., Van Nostrand Reinhold,New York; Briaud, J. L. (1989). “The PressuremeterTest For Highway Applications,” Federal HighwayAdministration Publication No. FHWA-IP-89-008,Washington D.C.)

7.5.6 California Bearing Ratio (CBR)

This ratio is often used as a measure of the qualityor strength of a soil that will underlie a pavement,

for determining the thickness of the pavement, itsbase, and other layers.

CBR ¼ F

F0(7:11)

where F ¼ force per unit area required to penetratea soil mass with a 3-in2 circular piston(about 2 in in diameter) at the rate of0.05 in/min

F0 ¼ force per unit area required for cor-responding penetration of a standardmaterial

Typically, the ratio is determined at 0.10 in pene-tration, although other penetrations sometimesare used. An excellent base course has a CBR of100%. A compacted soil may have a CBR of 50%,whereas a weaker soil may have a CBR of 10.

Tests to determine CBRmay be performed in thelaboratory or the field. ASTM standard tests areavailable for each case: “Standard Test Method forCBR (California Bearing Ratio) for LaboratoryCompacted Soils,” D1883, and “Standard TestMethod for CBR (California Bearing Ratio) of Soilsin Place,” D4429 (www.astm.org).

One criticism of the method is that it doesnot simulate the shearing forces that develop insupporting materials underlying a flexible pave-ment.

7.5.7 Soil Permeability

The coefficient of permeability k is a measure of therate of flow of water through saturated soil under agiven hydraulic gradient i, cm/cm, and is definedin accordance with Darcy’s law as:

V ¼ kiA (7:12)

where V ¼ rate of flow, cm3/s.

A ¼ cross-sectional area of soil conveyingflow, cm2

k is dependent on the grain-size distribution, voidratio, and soil fabric and typically may vary from asmuch as 10 cm/s for gravel to less than 1027 cm/sfor clays. For typical soil deposits, k for horizontalflow is greater than k for vertical flow, often by anorder of magnitude.

Soil-permeability measurements can be con-ducted in tests under falling or constant head,either in the laboratory or the field. Large-scale

Geotechnical Engineering n 7.15

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pumping (drawdown) tests also may be conductedin the field to provide a significantly larger scalemeasurement of formation permeability. Correla-tions of kwith soil gradation and relative density orvoid ratio have been developed for a variety ofcoarse-grained materials. General correlations of kwith soil index and physical properties are lessreliable for fine-grained soils because other factorsthan porosity may control.

(T. W. Lambe and R. V. Whitman, “SoilMechanics,” John Wiley & Sons, Inc., New York,www.wiley.com.)

7.6 Site Investigations

The objective of most geotechnical site investi-gations is to obtain information on the site andsubsurface conditions that is required for designand construction of engineered facilities and forevaluation and mitigation of geologic hazards,such as landslides, subsidence, and liquefaction.The site investigation is part of a fully integratedprocess that includes:

1. Synthesis of available data

2. Field and laboratory investigations

3. Characterization of site stratigraphy and soilproperties

4. Engineering analyses

5. Formulation of design and construction criteriaor engineering evaluations

7.6.1 Planning and Scope

In the planning stage of a site investigation, allpertinent topographical, geologic, and geotechnicalinformation available should be reviewed andassessed. In urban areas, the development historyof the site should be studied and evaluated. It isparticularly important to provide or require that aqualified engineer direct and witness all fieldoperations.

The scope of the geotechnical site investigationvaries with the type of project but typically in-cludes topographic and location surveys, explora-tory “drilling and sampling, in situ testing andgroundwater monitoring. Frequently the investi-gation is supplemented by test pits, geophysicaltests, air photos and remote sensing”.

7.6.2 Exploratory Borings

Typical boring methods employed for geotechnicalexploration consist of rotary drilling, auger drilling,percussion drilling, or any combination of these.Deep soil borings (greater than about 100 ft) areusually conducted by rotary-drilling techniquesrecirculating a weighted drilling fluid to maintainborehole stability. Auger drilling, with hollow-stemaugers to facilitate sampling, is a widely usedand economical method for conducting short- tointermediate-length borings. Most of the drill rigsare truck-mounted and have a rockcoring capability.Awide variety of drilling machines are available toprovide access to the most difficult projects.

With percussion drilling, a casing is usuallydriven to advance the boring. Water circulation ordriven, clean-out spoons are often used to removethe soil (cuttings) in the casing. This method isemployed for difficult-access locations whererelatively light and portable drilling equipment isrequired. A rotary drill designed for rock coring isoften included.

Soil Samples n These are usually obtainedby driving a split-barrel sampler or by hydrauli-cally or mechanically advancing a thin-wall(Shelby) tube sampler. Driven samplers, usually 2in outside diameter (OD), are advanced 18 in by a140-lb hammer dropped 30 in (ASTM D1586). Thenumber of blows required to drive the last 12 in ofpenetration constitutes the standard penetrationresistance (SPT) value. The Shelby tube sampler,used for undisturbed sampling, is typically a 12-to 16-gage seamless steel tube and is nominally3 in OD (ASTM D1587). In soils that are soft orotherwise difficult to sample, a stationary pistonsampler is used to advance a Shelby tube eitherhydraulically (pump pressure) or by the down-crowd system of the drill.

Rotary core drilling is typically used to obtaincore samples of rock and hard, cohesive soils thatcannot be penetrated by conventional samplingtechniques. Typically, rock cores are obtained withdiamond bits that yield core-sample diametersfrom 7⁄8 (AX) to 21⁄8 (NX). For hard clays and softrocks, a 3- to 6-in OD undisturbed sample can alsobe obtained by rotary drilling with a Dennison orPitcher sampler.

Test Boring Records (Logs) n These typi-cally identify the depths and material classification

7.16 n Section Seven

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of the various strata encountered, the samplelocation and penetration resistance, rock-coreinterval and recovery, groundwater levels encoun-tered during and after drilling. Special subsurfaceconditions should be noted on the log, for example,changes in drilling resistance, hole caving, voids,and obstructions. General information requiredincludes the location of the boring, surfaceelevation, drilling procedures, sampler and corebarrel types, and other information relevant tointerpretation of the boring log.

Groundwater Monitoring n Monitoringgroundwater levels is an integral part of boringand sampling operations. Groundwater measure-ments during and at least 12 h after drilling areusually required. Standpipes are often installed intest borings to provide longer-term observations;they are typically small-diameter pipes perforatedin the bottom few feet of casing.

If irregular piezometric profiles are suspected,piezometers may be set and sealed so as to measurehydrostatic heads within selected strata. Piezo-meters may consist of watertight 1⁄2 to 3⁄4 -in ODstandpipes or plastic tubing attached to porousceramic or plastic tips. Piezometers with electronicor pneumatic pressure sensors have the advant-ageof quick response and automated data acqui-sition. However, it is not possible to conduct in situpermeability tests with these closed-systempiezometers.

7.6.3 In Situ Testing Soils

In situ tests can be used under a variety ofcircumstances to enhance profile definition, toprovide data on soil properties, and to obtainparameters for empirical analysis and designapplications.

Quasi-static and dynamic cone penetrationtests (CPTs) quite effectively enhance profile defi-nition by providing a continuous record ofpenetration resistance. Quasi-static cone penetra-tion resistance is also correlated with the relativedensity, OCR, friction angle, and compressibility ofcoarse-grained soils and the undrained shearstrength of cohesive soils. Empirical foundationdesign parameters are also provided by the CPT.

The standard CPT in the United States consistsof advancing a 10-cm2, 608 cone at a rate between

1.5 and 2.5 cm/s and recording the resistance tocone penetration (ASTM D3441). A friction sleevemay also be incorporated to measure frictionalresistance during penetration. The cone may beincrementally (mechanical penetrometer) or con-tinuously (electronic penetrometer) advanced.

Dynamic cones are available in a variety ofsizes, but in the United States, they typically have a2-in upset diameter with a 608 apex. They aredriven by blows of a 140-lb hammer dropped 30 in.Automatically driven cone penetrometers arewidely used in western Europe and are portableand easy to operate.

Pressuremeter tests (PMTs) provide an insitu interpretation of soil compressibility and un-drained shear strength. Pressuremeters have alsobeen used to provide parameters for foundationdesign.

The PMT is conducted by inserting a probecontaining an expandable membrane into a drillhole and then applying a hydraulic pressure toradially expand the membrane against the soil, tomeasure its volume change under pressure. Theresulting curve for volume change vs. pressure isthe basis for interpretation of soil properties.

Vane shear tests provide in situ measurementsof the undrained shear strength of soft to firm clays,usually by rotating a four-bladed vane andmeasuring the torsional resistance T. Undrainedshear strength is then calculated by dividing T bythe cylindrical side and end areas inscribing thevane. Account must be taken of torque rod friction(if unsleeved), which can be determined bycalibration tests (ASTM D2573). Vane tests aretypically run in conjunction with borings, but insoft clays the vane may be advanced without apredrilled hole.

Other in situ tests occasionally used to providesoil-property data include plate load tests (PLTs),borehole shear (BHS) tests, and dilatometer tests.The PLT technique may be useful for providingdata on the compressibility of soils and rocks. TheBHS may be useful for characterizing effective-shear-strength parameters for relatively free-drain-ing soils as well as total-stress (undrained) shear-strength parameters for fine-grained soils. Dilat-ometer tests provide a technique for investigatingthe horizontal effective stress s 0

ho and soil compres-sibility. Some tests use small-diameter probes tomeasure pore-pressure response, acoustical emis-sions, bulk density, and moisture content duringpenetration.

Geotechnical Engineering n 7.17

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Prototype load testing as part of the geotechni-cal investigation represents a variation of in situtesting. It may include pile load tests, earth loadtests to investigate settlement and stability, andtests on small-scale or full-size shallow foundationelements. Feasibility of construction can also beevaluated at this time by test excavations, indicatorpile driving, drilled shaft excavation, rock ripp-ability trials, dewatering tests, and so on.

(H. Y. Fang, “Foundation Engineering Hand-book,” 2nd ed., VanNostrandReinhold,NewYork.)

7.6.4 Geophysical Investigations

Geophysical measurements are often valuable inevaluation of continuity of soil and rock stratabetween boring locations. Under some circum-stances, such data can reduce the number ofborings required. Certain of these measurementscan also provide data for interpreting soil and rockproperties. The techniques often used in engineer-ing applications are as follows:

Seismic-wave-propagation techniques includeseismic refraction, seismic rejection, and directwave-transmission measurements. Refraction tech-niques measure the travel time of seismic wavesgenerated from a single-pulse energy source todetectors (geophones) located at various distancesfrom the source. The principle of seismic refractionsurveying is based on refraction of the seismicwaves at boundaries of layers with differentacoustical impedances. This technique is illustratedin Fig. 7.3.

Compression P wave velocities are interpretedto define velocity profiles that may be correlatedwith stratigraphy and the depth to rock. The Pwave velocity may also help identify type of soil.However, in saturated soils the velocity measuredrepresents wave transmission through water-filledvoids. This velocity is about 4800 ft/s regardless ofthe soil type. Low-cost single- and dual-channelseismographs are available for routine engineeringapplications.

Seismic reflection involves measuring the timesrequired for a seismic wave induced at the surfaceto return to the surface after reflection from theinterfaces of strata that have different acousticalimpedances. Unlike refraction techniques, whichusually record only the first arrivals of the seismicwaves, wave trains are concurrently recorded byseveral detectors at different positions so as to

provide a pictorial representation of formationstructure. This type of survey can be conductedin both marine and terrestrial environmentsand usually incorporates comparatively expensivemultiple-channel recording systems.

Direct seismic-wave-transmission techniquesinclude measurements of the arrival times of Pwaves and shear S waves after they have traveledbetween a seismic source and geophones placed atsimilar elevations in adjacent drill holes. Bymeasuring the precise distances between sourceand detectors, both S and P wave velocities can bemeasured for a given soil or rock interval if the holespacing is chosen to ensure a direct wave-trans-mission path.

Alternatively, geophones can be placed atdifferent depths in a drill hole to measure seismicwaves propagated down from a surface source nearthe drill hole. The detectors and source locationscan also be reversed to provide up-hole insteadof down-hole wave propagation. Although thismethod does not provide as precise a measure ofinterval velocity as the cross-hole technique, it issubstantially less costly.

Directwave-transmission techniques are usuallyconducted so as to maximize S wave energygeneration and recognition by polarization of theenergy input. S wave interpretations allow calcu-lation of the small-strain shear modulus Gmax

required for dynamic response analysis. Poisson’sratio can also be determined if both P and S wavevelocities can be recorded.

Fig. 7.3 Illustration of seismic refraction concept.

7.18 n Section Seven

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Resistivity and conductance investigation tech-niques relate to the proposition that stratigraphicdetails can be derived from differences in theelectrical resistance or conductivity of individualstrata. Resistivity techniques for engineeringpurposes usually apply the Wenner method ofinvestigation, which involves four equally spacedsteel electrodes (pins). The current is introducedthrough the two end pins, and the associated poten-tial drop is measured across the two center pins.The apparent resistivity r is then calculated as afunction of current I, potential difference V, and pinspacing a as:

r ¼ 2paV

I(7:13)

To investigate stratigraphic changes, tests are run atsuccessively greater pin spacings. Interpretationsare made by analyzing accumulative or discrete-interval resistivity profiles or by theoretical curve-matching procedures.

A conductivity technique for identifying sub-surface anomalies and stratigraphy involvesmeasuring the transient decay of a magnetic fieldwith the source (dipole transmitter) in contact withthe surface. The depth of apparent conductivitymeasurement depends on the spacing and orien-tation of the transmitter and receiver loops.

Both resistivity and conductivity interpretationsare influenced by groundwater chemistry. Thischaracteristic has been utilized to map the extent ofsome groundwater pollutant plumes by conduc-tivity techniques.

Other geophysical methods with more limitedengineering applications include gravity andmagnetic field measurements. Surveys using thesetechniques can be airborne, shipborne, or ground-based. Microgravity surveys have been useful indetecting subsurface solution features in carbonaterocks.

Aerial surveys are appropriate where largeareas are to be explored. Analyses of conventionalaerial stereoscopic photographs; thermal and false-color, infrared imagery; multispectral satelliteimagery; or side-looking aerial radar can disclosethe surface topography and drainage, linearfeatures that reflect geologic structure, type ofsurface soil and often the type of underlying rock.These techniques are particularly useful in locatingfilled-in sinkholes in karst regions, which are oftencharacterized by closely spaced, slight surfacedepressions.

(M. B. Dobrin, “Introduction to GeophysicalProspecting,” McGraw-Hill Book Company, NewYork, books.mcgraw-hill.com.)

7.7 Hazardous Site andFoundation Conditions

There are a variety of natural hazards of potentialconcern in site development and foundationdesign. Frequently, these hazards are overlookedor not given proper attention, particularly in areaswhere associated failures have been infrequent.

7.7.1 Solution-Prone Formations

Significant areas in the eastern and midwesternUnited States are underlain by formations (carbon-ate and evaporate rocks) susceptible to dissolution.Subsurface voids created by dissolution range fromopen jointing to huge caverns. These features havecaused catastrophic failures and detrimental settle-ments of structures as a result of ground loss orsurface subsidence.

Special investigations designed to identify rock-solution hazards include geologic reconnaissance,air photo interpretation, and geophysical (resis-tivity, microgravity, and so on) surveys. To miti-gate these hazards, careful attention should begiven to:

1. Site drainage to minimize infiltration of surfacewaters near structures

2. Limitation of excavations to maximize thethickness of soil overburden

3. Continuous foundation systems designed toaccommodate a partial loss of support beneaththe foundation system

4. Deep foundations socketed into rock anddesigned solely for socket bond resistance

It is prudent to conduct special proof testing ofthe bearing materials during construction insolution-prone formations. Proof testing oftenconsists of soundings continuously recording thepenetration resistance through the overburden andthe rate of percussion drilling in the rock. Suspectzones are thus identified and can be improved byexcavation and replacement or by in situ grouting.

Geotechnical Engineering n 7.19

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7.7.2 Expansive Soils

Soils with a medium to high potential for causingstructural damage on expansion or shrinkage arefound primarily throughout the Great Plains andGulf Coastal Plain Physiographic Provinces. Heaveor settlement of active soils occurs because of achange in soil moisture in response to climaticchanges, construction conditions, changes in sur-face cover, and other conditions that influence thegroundwater and evapotransportation regimes.Differential foundation movements are broughtabout by differential moisture changes in thebearing soils. Figure 7.4 presents a method forclassifying the volume-change potential of clay as afunction of activity.

Investigations in areas containing potentiallyexpansive soils typically include laboratory swelltests. Infrequently, soil suction measurements aremade to provide quantitative evaluations of volume-change potential. Special attention during the fieldinvestigation should be given to evaluation of thegroundwater regime and to delineation of the depthof active moisture changes.

Common design procedures for preventingstructural damage include mitigation of moisturechanges, removal or modification of expansivematerial and deep foundation support. Horizontal

and vertical moisture barriers have been utilized tominimize moisture losses due to evaporation orinfiltration and to cut off subsurface groundwaterflow into the area of construction. Excavation ofpotentially active materials and replacement withinert material or with excavated soil modified bythe addition of lime have proved feasible whereexcavation quantities are not excessive.

Deep foundations (typically drilled shafts) havebeen used to bypass the active zone and to resist orminimize uplift forces that may develop on theshaft. Associated grade beams are usually con-structed to prevent development of uplift forces.

(“Engineering and Design of Foundations onExpansive Soils,” U.S. Department of the Army,1981. L. D. Johnson, “Predicting Potential Heaveand Heave with Time and Swelling FoundationSoils,” Technical Report S-78-7, U.S. Army Engin-eers Waterways Experiment Station, Vicksburg,Miss., 1978.)

7.7.3 Landslide Hazards

Landslides are usually associated with areas ofsignificant topographic relief that are characterizedby relatively weak sedimentary rocks (shales,siltstones, and so forth) or by relatively impervioussoil deposits containing interbedded water-bearingstrata. Under these circumstances, slides that haveoccurred in the geologic past, whether or notcurrently active, represent a significant risk forhillside site development. In general hillsidedevelopment in potential landslide areas is a mosthazardous undertaking. If there are alternatives,one of those should be adopted.

Detailed geologic studies are required to evalu-ate slide potential and should emphasize detectionof old slide areas. Procedures that tend to stabilizean active slide or to provide for the continuedstability of an old slide zone include:

1. Excavation at the head of the sliding mass toreduce the driving force

2. Subsurface drainage to depress piezometriclevels along potential sliding surface

3. Buttressing at the toe of the potential slidingmassto provide a force-resisting slide movement

Within the realm of economic feasibility, thereliability of these or any other procedures to

Fig. 7.4 Chart for ratingvolume-changepotentialof expansive soils.

7.20 n Section Seven

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stabilize active or old slides involving a significantsliding mass are generally of a relatively low order.

On hillsides where prior slides have not beenidentified, care should be taken to reduce thesliding potential of superimposed fills by removingweak or potentially unstable surficial materials,benching and keying the fill into competentmaterials, and (most importantly) installing effec-tive subsurface drainage systems. Excavations thatresult in steepening of existing slopes are poten-tially detrimental and should not be employed.Direction and collection of surface runoff so as toprevent slope erosion and infiltration are recom-mended. (“Landslides Investigation and Mitiga-tion,” Transportation Research Board SpecialReport 247 National Academy Press 1996)

7.7.4 Liquefaction of Soils

Relatively loose saturated cohesionless soils maybecome unstable under cyclic shear loading such asthat imposed by earthquake motions. A simplifiedmethod of analysis of the liquefaction potential ofcohesionless soils has been proposed for predictingthe ratio of the horizontal shear stress tav to theeffective overburden pressure s 0

vo imposed by anearthquake. (tav represents a uniform cyclic-stress

representation of the irregular time history of shearstress induced by the design earthquake.) This fieldstress ratio is a function of the maximum horizontalground-surface acceleration amax, the acceleration ofgravity g, a stress-reduction factor rd, and totalvertical stress svo and approximated as

tavs 0vo

¼ 0:65amax

g

svo

s 0vo

rd (7:14)

rd varies from 1.0 at the ground surface to 0.9 for adepth of 30 ft. (H. B. Seed and I. M. Idriss, “ASimplified Procedure for Evaluating Soil Liquefac-tion Potential,” Report EERC 70-9, EarthquakeEngineering Research Center, University of Cali-fornia, Berkeley, 1970.)

Stress ratios that produce liquefaction may becharacterized from correlations with field obser-vations (Fig. 7.5). The relevant soil properties arerepresented by their corrected penetration resistance

N1 ¼ (1� 1:25 logs 0vo)N (7:15)

where s 0vo is in units of tons/ft2. The stress ratio

causing liquefaction should be increased about 25%for earthquakes with Richter magnitude 6 or lower.(H. B. Seed, “Evaluation of Soil Liquefaction Effectson Level Ground during Earthquakes,” Symposiumon Liquefaction Problems and Geotechnical Engineering,

Fig. 7.5 Chart correlates cyclic-stress ratios that produce soil liquefaction with standard penetrationresistance. (After H. B. Seed.)

Geotechnical Engineering n 7.21

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ASCE National Convention, Philadelphia, Pa.,1976. (“Design Guidance: Geotechnical EarthquakeEngineering For Highways,” Federal HighwayAdministration, Publication No. FHWA-SA-97-076,May 1997)

Significantly, more elaborate, dynamic finite-element procedures have been proposed to evalu-ate soil liquefaction and degradation of undrainedshear strength as well as generation and dissipationof pore-water pressure in soils as a result of cyclicloading. Since stress increases accompany dissipa-tion of pore-water pressures, settlements due tocyclic loading can also be predicted. Such residualsettlements can be important even though liquefac-tion has not been induced.

(P. B. Schnabel, J. Lysmer, and H. B. Seed, “AComputer Program for Earthquake Response Anal-ysis ofHorizontally Layered Sites,” Report EERC 72-12, Earthquake Engineering Research Center,University of California, Berkeley, 1972; H. B. Seed,P. P. Martin, and J. Lysmer, “Pore-Water PressureChanges During Soil Liquefaction,” ASCE Journal ofGeotechnical Engineering Division, vol. 102, no. GT4,1975; K. L. Lee andA.Albaisa, “Earthquake-InducedSettlements in Saturated Sands,” ASCE Journal ofGeotechnical Engineering Division, vol. 100, no. GT4,1974, www.asce.org.)

7.8 Types of Footings

Spread (individual) footings (Fig. 7.6) are the mosteconomical shallow foundation types but are moresusceptible to differential settlement. They usually

support single concentrated loads, such as thoseimposed by columns.

Shallow Foundations

Shallow foundation systems can be classified asspread footings, wall and continuous (strip) foot-ings, and mat (raft) foundations. Variations arecombined footings, cantilevered (strapped) foot-ings, two-way strip (grid) footings, and discon-tinuous (punched) mat foundations.

Combined footings (Fig. 7.7) are used where thebearing areas of closely spaced columns overlap.Cantilever footings (Fig. 7.8) are designed toaccommodate eccentric loads.

Continuous wall and strip footings (Fig. 7.9) canbe designed to redistribute bearing-stress concen-trations and associated differential settlements inthe event of variable bearing conditions or local-ized ground loss beneath footings.

Mat foundations have the greatest facility forload distribution and for redistribution of subgradestress concentrations caused by localized anom-alous bearing conditions. Mats may be constantsection, ribbed, waffled, or arched. Buoyancy matsare used on compressible soil sites in combinationwith basements or subbasements, to create apermanent unloading effect, thereby reducing thenet stress change in the foundation soils.

(M. J. Tomlinson, “Foundation Design andConstruction,” John Wiley & Sons, Inc., New York(www.wiley.com); J. E. Bowles, “Foundation

Fig. 7.6 Spread footing. Fig. 7.7 Combined footing.

7.22 n Section Seven

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Analysis and Design,” McGraw-Hill Book Com-pany, New York. (books.mcgraw-hill.com) “SpreadFootings for Highway Bridges,” Federal HighwayAdministration, Publication No. FHWA-RD-86-185October, 1987)

7.9 Approach to FoundationAnalysis

Shallow-foundation analysis and formulation ofgeotechnical design provisions are generallyapproached in the following steps:

1. Establish project objectives and design orevaluation conditions.

2. Characterize site stratigraphy and soil rockproperties.

3. Evaluate load-bearing fill support or subsoil-improvement techniques, if applicable.

4. Identify bearing levels; select and proportioncandidate foundation systems.

5. Conduct performance, constructibility, andeconomic feasibility analyses.

6. Repeat steps 3 through 5 as required to satisfythe design objectives and conditions.

The scope and detail of the analyses vary accordingto the project objectives.

Fig. 7.8 Cantilever footing.

Fig. 7.9 Continuous footings for (a) a wall, (b) several columns.

Geotechnical Engineering n 7.23

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Project objectives to be quantified are essen-tially the intent of the project assignment and thespecific scope of associated work. The conditionsthat control geotechnical evaluation or design workinclude criteria for loads and grades, facility op-erating requirements and tolerances, constructionschedules, and economic and environmental con-straints. Failure to provide a clear definition ofrelevant objectives and design conditions can resultin significant delays, extra costs, and, under somecircumstances, unsafe designs.

During development of design conditions forstructural foundations, tolerances for total anddifferential settlements are commonly establishedas a function of the ability of a structure to toleratemovement. Suggested structure tolerances in termsof angular distortion are in Table 7.5. Angulardistortion represents the differential vertical move-ment between two points divided by the horizontaldistance between the points.

Development of design profiles for foundationanalysis ideally involves a synthesis of geologic andgeotechnical data relevant to site stratigraphy andsoil and rock properties. This usually requires siteinvestigations (see Arts. 7.6.1 to 7.6.4) and in situ orlaboratory testing, or both, of representative soiland rock samples (see Arts. 7.3 to 7.5.6).

To establish and proportion candidate foun-dation systems, consideration must first be given toidentification of feasible bearing levels. The depth ofthe foundation must also be sufficient to protectexposed elements against frost heave and to pro-vide sufficient confinement to produce a factor ofsafety not less than 2.5 (preferably 3.0) against shearfailure of the bearing soils. Frost penetration hasbeen correlatedwith a freezing index, which equalsthe number of days with temperature below 328 Fmultiplied by T 2 32, where T ¼ average dailytemperature. Such correlations can be applied inthe absence of local codes or experience. Generally,footing depths below final grade should be aminimum of 2.0 to 2.5 ft.

For marginal bearing conditions, considerationshould be given to improvement of the quality ofpotential bearing strata. Soil-improvement tech-niques include excavation and replacement oroverlaying of unsuitable subsoils by load-bearingfills, preloading of compressible subsoils, soildensification, soil reinforcement, and grouting tech-niques. Densification methods include high energysurface impact (dynamic compaction), on gradevibratory compaction, and subsurface vibratorycompaction by vibro-compaction techniques. Soilreinforcement methods include: stone columns,soil mixing, mechanically stabilized earth and soilnailing.

“Ground Improvement Technical Summaries,”Federal Highway Administration, Publication No.FHWA-SA-98-086, December 1999.

The choice of an appropriate soil improvementtechnique is highly dependent on performance re-quirements (settlement), site and subsurface condi-tions, time and space constraints and cost.

Assessment of the suitability of candidatefoundation systems requires evaluation of thesafety factor against both catastrophic failure andexcessive deformation under sustained and transi-ent design loads. Catastrophic-failure assessmentmust consider overstress and creep of the bearingsoils as well as lateral displacement of the foun-dation. Evaluation of the probable settlement be-havior requires analysis of the stresses imposedwithin the soil and, with the use of appropriate soilparameters, prediction of foundation settlements.Typically, settlement analyses provide estimates oftotal and differential settlement at strategiclocations within the foundation area and mayinclude time-rate predictions of settlement. Usually,the suitability of shallow foundation systems is

Table 7.5 Limiting Angular Distortions*

Structural ResponseAngularDistortion

Cracking of panel andbrick walls

1/100

Structural damage tocolumns and beams

1/150

Impaired overhead craneoperation

1/300

First cracking of panel walls 1/300Limit for reinforcedconcrete frame

1/400

Limit for wall cracking 1/500Limit for diagonallybraced frames

1/600

Limit for settlement-sensitivemachines

1/750

* Limits represent the maximum distortions that can be safelyaccommodated.

Source: After L. Bjerrum, European Conference on SoilMechanics and Foundation Engineering, Wiesbaden, Germany,vol. 2, 1963.

7.24 n Section Seven

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governed by the systems’ load-settlement responserather than bearing capacity.

7.10 Foundation-StabilityAnalysis

The maximum load that can be sustained byshallow foundation elements at incipient failure(bearing capacity) is a function of the cohesion andfriction angle of bearing soils as well as the widthB and shape of the foundation. The net bearingcapacity per unit area qu of a long footing is con-ventionally expressed as:

qu ¼ af cuNc þ s 0voNq þ bfgBNg (7:16)

where af ¼ 1.0 for strip footings and 1.3 forcircular and square footings

cu ¼ undrained shear strength of soil

s 0vo ¼ effective vertical shear stress in soil at

level of bottom of footing

bf ¼ 0.5 for strip footings, 0.4 for squarefootings, and 0.6 for circular footings

g ¼ unit weight of soil

B ¼ width of footing for square and rec-tangular footings and radius of foot-ing for circular footings

Nc, Nq, Ng ¼ bearing-capacity factors, functions ofangle of internal friction f (Fig. 7.10)

For undrained (rapid) loading of cohesive soils,f ¼ 0 and Eq. (7.16) reduces to

qu ¼ N0ccu (7:17)

where N0c ¼ af Nc. For drained (slow) loading of

cohesive soils, f and cu are defined in terms ofeffective friction angle f0 and effective stress c0u.

Modifications of Eq. (7.16) are also availableto predict the bearing capacity of layered soil andfor eccentric loading.

Rarely, however, does qu control foundationdesign when the safety factor is within the range of2.5 to 3. (Should creep or local yield be induced,excessive settlements may occur. This consider-ation is particularly important when selecting asafety factor for foundations on soft to firm clayswith medium to high plasticity.)

Equation (7.16) is based on an infinitely longstrip footing and should be corrected for othershapes. Correction factors by which the bearing-capacity factors should be multiplied are given inTable 7.6, in which L ¼ footing length.

The derivation of Eq. (7.16) presumes the soils tobe homogeneous throughout the stressed zone,which is seldom the case. Consequently, adjust-ments may be required for departures from homo-

Fig. 7.10 Bearing capacity factors for use in Eq. (7.16) as determined by Terzaghi and Meyerhof.

Geotechnical Engineering n 7.25

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geneity. In sands, if there is a moderate variation instrength, it is safe to use Eq. (7.16), but withbearing-capacity factors representing a weightedaverage strength.

For strongly varied soil profiles or interlayeredsands and clays, the bearing capacity of each layershould be determined. This should be done byassuming the foundation bears on each layersuccessively but at the contact pressure for thedepth below the bottom of the foundation of thetop of the layer.

Eccentric loading can have a significant impacton selection of the bearing value for foundationdesign. The conventional approach is to proportionthe foundation to maintain the resultant force

within its middle third. The footing is assumed tobe rigid and the bearing pressure is assumed tovary linearly as shown by Fig. 7.11b. If the resultantlies outside the middle third of the footing, it isassumed that there is bearing over only a portion ofthe footing, as shown in Fig. 7.11d. For theconventional case (Fig. 7.11a), the maximum andminimum bearing pressures are:

qm ¼ P

BL1+

6e

B

� �(7:18)

where B ¼width of rectangular footing

L ¼ length of rectangular footing

e ¼ eccentricity of loading

Fig. 7.11 Footings subjected to overturning.

Table 7.6 Shape Corrections for Bearing-Capacity Factors of Shallow Foundations*

Correction Factor

Shape of Foundation Nc Nq Ng

Rectangle† 1 þ (B/L) (Nq/Nc) 1 þ (B/L) tan f 1 2 0.4(B/L)

Circle and square 1 þ (Nq/Nc) 1 þ tan f 0.60

* After E. E. De Beer, as modified by A. S. Vesic. See H. Y. Fang, “Foundation Engineering Handbook,” Van Nostrand Reinhold,2d ed., New York.

† No correction factor is needed for long strip foundations.

7.26 n Section Seven

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For the other case (Fig. 7.11c), the soil pressureranges from 0 to a maximum of:

qm ¼ 2P

3L(B=2� e)(7:19)

For square or rectangular footings subject tooverturning about two principal axes and forunsymmetrical footings, the loading eccentricitiese1 and e2 are determined about the two principalaxes. For the case where the full bearing area of thefootings is engaged, qm is given in terms of thedistances from the principal axes c1 and c2, the radiusof gyration of the footing area about the principalaxes r1 and r2, and the area of the footing A as:

qm ¼ P

A1þ e1c1

r21þ e2c2

r22

� �(7:20)

For the case where only a portion of the footing isbearing, the maximum pressure may be approxi-mated by trial and error.

For all cases of sustained eccentric loading, themaximum (edge) pressures should not exceed theshear strength of the soil and also the factor ofsafety should be at least 1.5 (preferably 2.0) againstoverturning.

The foregoing analyses, except for completelyrigid foundation elements, are a very conservativeapproximation. Because most mat foundations andlarge footings are not completely rigid, their defor-mation during eccentric loading acts to produce amore uniform distribution of bearing pressuresthan would occur under a rigid footing and toreduce maximum contact stresses.

In the event of transient eccentric loading, experi-ence has shown that footings can sustain maximumedge pressures significantly greater than the shearstrength of the soil. Consequently, some buildingcodes conservatively allow increases in sustained-load bearing values of 30% for transient loads.Reduced safety factors have also been used inconjunction with transient loading. For cases wheresignificant cost savings can be realized, finite-element analyses that model soil-structure inter-action can provide a more realistic evaluation of aneccentrically loaded foundation.

Allowable Bearing Pressures n Approxi-mate allowable soil bearing pressures, withouttests, for various soil and rocks are given inTable 7.7 for normal conditions. These basicbearing pressures should be used for preliminary

Table 7.7 Allowable Bearing Pressures for Soils

Soil Material Pressure, tons/ft2 Notes

Unweathered sound rock 60 No adverse seam structureMedium rock 40Intermediate rock 20Weathered, seamy, or porous rock 2 to 8Hardpan 12 Well cementedHardpan 8 Poorly cementedGravel soils 10 Compact, well gradedGravel soils 8 Compact with more than 10% gravelGravel soils 6 Loose, poorly gradedGravel soils 4 Loose, mostly sandSand soils 3 to 6 DenseFine sand 2 to 4 DenseClay soils 5 HardClay soils 2 Medium stiffSilt soils 3 DenseSilt soils 11⁄2 Medium denseCompacted fills Compacted to 90% to 95% of maximum

density (ASTM D1557)Fills and soft soils 2 to 4 By field or laboratory test only

Geotechnical Engineering n 7.27

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design only. Final design values should be based onthe results of a thorough subsurface investigationand the results of engineering analysis of potentialfailure and deformation limit states.

Resistance to Horizontal Forces n Thehorizontal resistance of shallow foundations ismobilized by a combination of the passive soilresistance on the vertical projection of theembedded foundation and the friction betweenthe foundation base and the bearing soil and rock.The soil pressure mobilized at full passive resis-tance, however, requires lateral movements greaterthan can be sustained by some foundations. Con-sequently, a soil resistance between the at-rest andpassive-pressure cases should be determined onthe basis of the allowable lateral deformations ofthe foundation.

The frictional resistance f to horizontal trans-lation is conventionally estimated as a function ofthe sustained, real, load-bearing stresses qd from

f ¼ qd tan d (7:21)

where d is the friction angle between the foun-dation and bearing soils. d may be taken as equiv-alent to the internal-friction angle f0 of thesubgrade soils. In the case of cohesive soils, f ¼ cu.Again, some relative movement must be realized todevelop f, but this movement is less than thatrequired for passive-pressure development.

If a factor of safety against translation of at least1.5 is not realized with friction and passive soil/rock pressure, footings should be keyed to increasesliding resistance or tied to engage additionalresistance. Building basement and shear walls arealso commonly used to sustain horizontal loading.

7.11 StressDistributionUnderFootings

Stress changes imposed in bearing soils by earthand foundation loads or by excavations areconventionally predicted from elastic half-spacetheory as a function of the foundation shape andthe position of the desired stress profile. Elasticsolutions available may take into account foun-dation rigidity, depth of the compressible zone,superposition of stress from adjacent loads, layeredprofiles, and moduli that increase linearly withdepth.

For most applications, stresses may be com-puted by the pressure-bulb concept with themethods of either Boussinesq or Westergaard. Forthick deposits, use the Boussinesq distributionshown in Fig. 7.12a; for thinly stratified soils, usethe Westergaard approach shown in Fig. 7.12b.These charts indicate the stresses q beneath asingle foundation unit that applies a pressure at itsbase of qo.

Most facilities, however, involve not only mul-tiple foundation units of different sizes, but alsofloor slabs, perhaps fills, and other elements thatcontribute to the induced stresses. The stressesused for settlement calculation should include theoverlapping and contributory stresses that mayarise from these multiple loads.

7.12 Settlement Analyses ofCohesive Soils

Settlement of foundations supported on cohesivesoils is usually represented as the sum of theprimary one-dimensional consolidation rc, imme-diate ri, and secondary rs settlement components.Settlement due to primary consolidation is con-ventionally predicted for n soil layers by Eq. (7.22)and (7.23). For normally consolidated soils,

rc ¼Xni¼1

Hi C0c log

sv

s 0vo

� �(7:22)

where Hi ¼ thickness of ith soil layer

C0c ¼ strain referenced compression index

for ith soil layer (Art. 7.5.4)

sv ¼ sum of average s 0vo and average

imposed vertical-stress change Dsv inith soil layer

s 0vo ¼ initial effective overburden pressure at

middle of ith layer (Art 7.5.3)

For overconsolidated soils with sv . s 0vm,

rc ¼Xni¼1

Hi C 0r log

s 0vm

s 0vo

þ C0c log

sv

s 0vm

� �(7:23)

where C0r ¼ strain referenced recompression index

of ith soil layer (Art. 7.5.4)

s 0vm ¼ preconsolidation (maximum past con-

solidation) pressure at middle of ithlayer (Art. 7.5.3)

7.28 n Section Seven

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The maximum thickness of the compressible soilzone contributing significant settlement can betaken to be equivalent to the depth where Dsv ¼0:1s 0

vo.Equation (7.22) can also be applied to over-

consolidated soils if sv is less than s 0vm and C0

r issubstituted for C0

c.Inasmuch as Eqs. (7.22) and (7.23) represent one-

dimensional compression, they may provide ratherpoor predictions for cases of three-dimensionalloading. Consequently, corrections to rc have beenderived for cases of three-dimensional loading.These corrections are approximate but represent animproved approach when loading conditionsdeviate significantly from the one-dimensionalcase. (A. W. Skempton and L. Bjerrum, “AContribution to Settlement Analysis of Foun-dations on Clay,” Geotechnique, vol. 7, 1957.)

The stress-path method of settlement analysisattempts to simulate field loading conditions by

conducting triaxial tests so as to track thesequential stress changes of an average point orpoints beneath the foundation. The strains associ-ated with each drained and undrained loadincrement are summed and directly applied to thesettlement calculation. Deformation moduli canalso be derived from stress-path tests and used inthree-dimensional deformation analysis.

Three-dimensional settlement analyses usingelastic solutions have been applied to both drainedand undrained conditions. Immediate (elastic)foundation settlements ri, representing the un-drained deformation of saturated cohesive soils,can be calculated by discrete analysis [Eq. (7.25)].

ri ¼Xni¼1

His1 � s3

Ei(7:24)

where s1 2 s3 ¼ change in average deviatorstress within each layer influenced by applied

Fig. 7.12 Stress distribution under a square footing with side B and under a continuous footing withwidth B, as determined by equations of (a) Boussinesq and (b) Westergaard.

Geotechnical Engineering n 7.29

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load. Note that Eq. (7.24) is strictly applicableonly for axisymmetrical loading. Drained three-dimensional deformation can be estimated fromEq. (7.24) by substituting the secant modulus E0

s forE (see Art. 7.5.5).

The rate of one-dimensional consolidation canbe evaluated with Eq. (7.26) in terms of the degreeof consolidation U and the nondimensional timefactor Tv . U is defined by

U ¼ rtrc

¼ 1� utui

(7:25)

where rt ¼ settlement at time t after instantaneousloading

rc ¼ ultimate consolidation settlement

ut ¼ excess pore-water pressure at time t

ui ¼ initial pore-water pressure (t ¼ 0)

To correct approximately for the assumed in-stantaneous load application, rt at the end of theloading period can be taken as the settlementcalculated for one-half of the load application time.The time t required to achieve U is evaluated asa function of the shortest drainage path withinthe compressible zone h, the coefficient of con-

solidation Cv , and the dimensionless time factor Tv

from

t ¼ Tvh2

Cv(7:26)

Closed-form solutions Tv vs. U are available for avariety of initial pore-pressure distributions.(H. Y. Fang, “Foundation Engineering Handbook,”2nd ed., Van Nostrand Reinhold Company, NewYork.)

Solutions for constant and linearly increasing uiare shown in Fig. 7.13. Equation (7.27) presents anapproximate solution that can be applied to theconstant initial ui distribution case for Tv . 0.2.

U ¼ 1� 8

p2e�p2Tv=4 (7:27)

where e ¼ 2.71828. Numerical solutions for any uiconfiguration in a single compressible layer as wellas solutions for contiguous clay layers may bederived with finite-difference techniques.

(R. F. Scott, “Principles of Soil Mechanics,” Ad-dison-Wesley Publishing Company, Inc., Reading,Mass.)

The coefficient of consolidation Cv should beestablished based on experience and from site

Fig. 7.13 Curves relate degree of consolidation and time factor Tv .

7.30 n Section Seven

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specific conventional consolidation tests by fittingthe curve for time vs. deformation (for an appro-priate load increment) to the theoretical solution forconstant ui. For tests of samples drained at top andbottom, Cv may be interpreted from the curve forlog time or square root of time vs. strain (or dialreading) as

Cv ¼ TvH2

4t(7:28)

where H ¼ height of sample, in

t ¼ time for 90% consolidation (ffiffit

pcurve)

or 50% consolidation (log t curve), days

Tv ¼ 0.197 for 90% consolidation or 0.848 for50% consolidation

(See T. W. Lambe and R. V. Whitman, “SoilMechanics,” John Wiley & Sons, Inc., New York,www.wiley.com), for curve-fitting procedures.)Larger values of Cv are usually obtained with theffiffit

pmethod and appear to be more representative of

field conditions.Secondary compression settlement rs is as-

sumed, for simplicity, to begin on completion ofprimary consolidation, at time t100 correspondingto 100% primary consolidation. rs is then calculatedfrom Eq. (7.29) for a given period t after t100.

rs ¼Xni¼1

HiCa logt

t100(7:29)

Hi represents the thickness of compressible layersand Ca is the coefficient of secondary compressiongiven in terms of volumetric strain (Art. 7.5.4).

The ratio Ca to compression index Cc is nearlyconstant for a given soil type and is generallywithinthe range of 0.045+0.015. Ca, as determined fromconsolidation tests (Fig. 7.2), is extremely sensitiveto pressure-increment ratios of less than about 0.5(standard is 1.0). The effect of overconsolidation,either from natural or construction preload sources,is to significantly reduce Ca. This is an importantconsideration in the application of preloading forsoil improvement.

The rate of consolidation due to radial drainage isimportant for the design of vertical wick drains. As arule, drains are installed in compressible soils toreduce the time required for consolidation and toaccelerate the associated gain in soil strength.Vertical drains are typically used in conjunctionwith preloading as a means of improving thesupporting ability and stability of the subsoils.

(S. J. Johnson, “Precompression for ImprovingFoundation Soils,”ASCE Journal of SoilMechanics andFoundation Engineering Division, vol. 96, no. SM1,1970 (www.asce.org); R. D. Holtz andW. D. Kovacs,“An Introduction to Geotechnical Engineer-ing,” Prentice-Hall, Inc., Englewood Cliffs, N.J.(www.prenhall.com) “PrefabricatedVertical Drains”Federal Highway Administration, Publication No.FHWA-RD-86-168, 1986)

7.13 Settlement Analysis ofSands

The methods most frequently used to estimate thesettlement of foundations supported by relativelyfree-draining cohesionless soils generally employempirical correlations between field observationsand in situ tests. The primary correlative tests areplate bearing (PLT), cone penetration resistance(CPT), and standard penetration resistance (SPT)(see Art 7.6.3). These methods, however, aredeveloped from data bases that contain a numberof variables not considered in the correlations and,therefore, should be applied with caution.

Time rate of settlement in coarse grained soils isextremely rapid. This behavior may be used to theadvantage of the designer where both dead andlive loads are applied to the foundation. Oftenvertical deformations which occur during theconstruction process will have a minimal effect onthe completed facility.

Plate Bearing Tests n The most commonapproach is to scale the results of PLTs to full-sizefootings in accordance with Eq. (7.10). A lessconservative modification of this equation pro-posed by A. R. S. S. Barazaa is

r ¼ 2:5B

1:5þ B

� �2r1 (7:30)

where B ¼ footing width, ft

r ¼ settlement of full-size bearing plate

rl ¼ settlement of 1-ft square bearing plate

These equations are not sensitive to the relativedensity, gradation, and OCR of the soil or to theeffects of depth and shape of the footing.

Use of large-scale load tests or, ideally, full-scaleload tests mitigates many of the difficulties of thepreceding approach but is often precluded by costs

Geotechnical Engineering n 7.31

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and schedule considerations. Unless relatively uni-form soil deposits are encountered, this approachalso requires a number of tests, significantlyincreasing the cost and time requirements. (SeeJ. K. Mitchell and W. S. Gardner, “In-Situ Meas-urement of Volume-Change Characteristics,” ASCESpecialty Conference on In-Situ Measurement ofSoil Properties, Raleigh, N.C., 1975.)

Cone Penetrometer Methods n Corre-lations between quasi-static penetration resistanceqc and observation of the settlement of bearing platesand small footings form the basis of foundationsettlement estimates using CPT data. The Buisman-DeBeer method utilizes a one-dimensional com-pression formulation. A recommendedmodificationof this approach that considers the influence of therelative density of the soil Dr and increased secantmodulus E0

s is

r ¼Xni¼1

Hi1:15s 0

vo

(1þD2r )qc

logs 0vo þ Dsv

s 0vo

(7:31)

where r ¼ estimated footing settlement. The s 0vo and

Dsv parameters represent the average effectiveoverburden pressure and vertical stress change foreach layer considered below the base of thefoundation (see Art 7.12). Equation (7.31) haslimitations because no consideration is given to: (1)soil stress history, (2) soil gradation, and (3) three-dimensional compression. Also, Eq. (7.31) incorpor-ates an empirical representation of E0

s, given byEq. (7.32), and has all the limitations thereof (seeArt. 7.5.5).

E0s ¼ 2(1þD2

r )qc (7:32)

The foregoing procedures are not applicable tolarge footings and foundation mats. From fieldobservations relating foundation width B, meters,to r/B, the upper limit for r/B, for B . 13.5 m, isgiven, in percent, approximately by

r

B¼ 0:194� 0:115 log

B

10(7:33)

For the same data base, the best fit of the average r/Bmeasurements ranges from about 0.09% (B ¼ 20 m)to 0.06% (B ¼ 80 m).

Standard Penetration ResistanceMethods n A variety of methods have been pro-posed to relate foundation settlement to standardpenetration resistance N. An approach proposed

by I. Alpan and G. G. Meyerhof appears reasonableand has the advantage of simplicity. Settlement S,in, is computed for B , 4 ft from

S ¼ 8q

N0 (7:34a)

and for B � 4 ft from

S ¼ 12q

N02B

1þ B

� �2

(7:34b)

where q ¼ bearing capacity of soil, tons/ft2

B ¼ footing width, ft

N0 is given approximately by Eq. (7.34c) for s 0vo �

40 psi.

N0 ¼ 50N

s 0vo þ 10

(7:34c)

and represents N (blows per foot) normalized fors 0vo ¼ 40psi (see Fig. 7.14).(G. G Meyerhof, “Shallow Foundations,” ASCE

Journal of Soil Mechanics and Foundation EngineeringDivision, vol. 91, no. SM92, 1965; W. G. Holtz andH. J. Gibbs, “Shear Strength of Pervious GravellySoils,”ProceedingsASCE, paper 867, 1956 (www.asce.org); R. B. Peck, W. E. Hanson, and T. H. Thornburn,“Foundation Engineering,” John Wiley & Sons, Inc.,New York, www.wiley.com.)

Laboratory Test Methods n The limitationsin developing representative deformation param-eters from reconstituted samples were described inArt. 7.5.5. A possible exception may be for thesettlement analyses of foundations supported bycompacted fill. Under these circumstances, con-solidation tests and stress-path, triaxial shear testson the fill materials may be appropriate for pro-viding the parameters for application of settlementanalyses described for cohesive soils.

(D. J. D’Appolonia, E. D’Appolonia, and R. F.Brisette, “Settlement of Spread Footings on Sand,”ASCE Journal of Soil Mechanics and FoundationEngineering Division, vol. 94, no. SM3, 1968, www.asce.org.)

Deep Foundations

Subsurface conditions, structural requirements,site location and features, and economics generallydictate the type of foundation to be employed for agiven structure.

7.32 n Section Seven

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Deep foundations, such as piles, drilled shafts,and caissons, should be considered when:

Shallow foundations are inadequate and structuralloads need to be transmitted to deeper, more com-petent soil or rock

Loads exert uplift or lateral forces on the foun-dations

Structures are required to be supported over water

Functionality of the structure does not allow fordifferential settlements

Future adjacent excavations are expected

7.14 Application of Piles

Pile foundations are commonly installed forbridges, buildings, towers, tanks, and offshorestructures. Piles are of two major types: prefabri-cated and installed with a pile-driving hammer, orcast-in-place. In some cases, a pile may incorporateboth prefabricated and cast-in-place elements.Driven piles may be made of wood, concrete, steel,or a combination of these materials. Cast-in-placepiles are made of concrete that is placed into anauger-drilled hole in the ground. When thediameter of a drilled or augured cast-in-place pileexceeds about 24 in, it is then generally classified as

a drilled shaft, bored pile, or caisson (Arts. 7.15.2,7.2, and 7.23).

The load-carrying capacity and behavior of asingle pile is governed by the lesser of thestructural strength of the pile shaft and the strengthand deformation properties of the supporting soils.When the latter governs, piles derive their capacityfrom soil resistance along their shaft and undertheir toe. The contribution of each of these twocomponents is largely dependent on subsurfaceconditions and pile type, shape, and method ofinstallation. Piles in sand or clay deposits withshaft resistance predominant are commonly knownas friction piles. Piles with toe resistance primaryare known as end-bearing piles. In reality,however, most piles have both shaft and toeresistance, albeit to varying degrees. The sum of theultimate resistance values of both shaft and toe istermed the pile capacity, which when divided byan appropriate safety factor yields the allowableload at the pile head.

The capacity of a laterally loaded pile is usuallydefined in terms of a limiting lateral deflection ofthe pile head. The ratio of the ultimate lateral loaddefining structural or soil failure to the associatedlateral design load represents the safety factor ofthe pile under lateral load.

Piles are rarely utilized singly, but are typicallyinstalled in groups. The behavior of a pile in agroup differs from that of the single pile. Often, the

Fig. 7.14 Curves relate relative density to standard penetration resistance and effective vertical stress.

Geotechnical Engineering n 7.33

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group effect dictates the overall behavior of the pilefoundation system.

The following articles provide a general knowl-edge of pile design, analysis, construction, andtesting methods. For major projects, it is advisablethat the expertise of a geotechnical engineer withsubstantial experience with deep-foundationsdesign, construction, and verification methods beemployed.

7.15 Pile Types

Piles that cause a significant displacement of soilduring installation are termed displacement piles.For example, closed-end steel pipes and precastconcrete piles are displacement piles, whereasopen-end pipes and H piles are generally limited-displacement piles. They may plug during drivingand cause significant soil displacement. Auger-castpiles are generally considered nondisplacementpiles since the soil is removed and replaced withconcrete during pile installation.

Piles are usually classified according to theirmethod of installation and type of material.Preformed driven piles may be made of concrete,steel, timber, or a combination of these materials.

7.15.1 Precast Concrete Piles

Reinforced or prestressed to resist handling andpile-driving stresses, precast concrete piles areusually constructed in a casting yard and trans-ported to the jobsite. Pretensioned piles (commonlyknown as prestressed piles) are formed in verylong casting beds, with dividers inserted to produceindividual pile sections. Precast piles come in avariety of cross sections; for example, square,round, octagonal. They may be manufactured fulllength or in sections that are spliced duringinstallation. They are suitable for use as frictionpiles for driving in sand or clay or as end-bearingpiles for driving through soft soils to firm strata.

Prestressed concrete piles usually have solidsections between 10 and 30 in square. Frequently,piles larger than 24 in square and more than 100 ftlong are cast with a hollow core to reduce pileweight and facilitate handling.

Splicing of precast concrete piles should gene-rally be avoided.When it is necessary to extend pilelength, however, any of several splicing methodsmay be used. Splicing can be accomplished, for

instance, by installing dowel bars of sufficientlength and then injecting grout or epoxy to bondthem and the upper and lower pile sections.Oversize grouted sleeves may also be used.Alternatives to these bonding processes includewelding of steel plates or pipes cast at pile ends.Some specialized systems employ mechanicaljointing techniques using pins to make theconnection. These mechanical splices reduce fieldsplice time, but the connector must be incorporatedin the pile sections at the time of casting.

All of the preceding methods transfer sometension through the splice. There are, however,systems, usually involving external sleeves (orcans), that do not transfer tensile forces; this is apossible advantage for long piles in which tensionstresses would not be high, but these systems arenot applicable to piles subject to uplift loading. Forprestressed piles, since the tendons require bond-development length, the jointed ends of the pilesections should also be reinforced with steel bars totransfer the tensile forces across the spliced area.

Prestressed piles also may be posttensioned.Such piles are mostly cylindrical (typically up to66-in diameter and 6-in wall thickness) and arecentrifugally cast in sections and assembled toform the required length before driving. Stressingis achieved with the pile sections placed end to endby threading steel cables through precast ducts andthen applying tension to the cables with hydraulicdevices. Piles up to 200 fit long have been thusassembled and driven.

Advantages of precast concrete piles includetheir ability to carry high axial and inclined loadsand to resist large bending moments. Also,concrete piles can be used as structural columnswhen extended above ground level. Disadvantagesinclude the extra care required during handlingand installation, difficulties in extending andcutting off piles to required lengths, and possibletransportation difficulties. Special machines, how-ever, are available for pile cutting, such as saws andhydraulic crushing systems. Care, however, isnecessary during all stages of pile casting,handling, transportation, and installation to avoiddamaging the piles.

Precast concrete piles are generally installedwith pile-driving hammers. For this purpose, pileheads should always be protected with cushioningmaterial. Usually, sheets of plywood areused.Otherprecautions should also be taken to protect pilesduring and after driving. When driving is expected

7.34 n Section Seven

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to be through hard soil layers or into rock, pile toesshould generally be fitted with steel shoes forreinforcement and protection from damage. Whenpiles are driven into soils and groundwater contain-ing destructive chemicals, special cement additivesor coatings should be used to protect concrete piles.Seawater may also cause damage to concrete pilesby chemical reactions or mechanical forces.

(“Recommended Practice for Design, Manu-facture, and Installation of Prestressed ConcretePiling,” Prestressed Concrete Institute, 209 W.Jackson Blvd., Chicago, IL 60604, www.pci.org.)

7.15.2 Cast-in-Place Concrete Piles

These are produced by forming holes in the groundand then filling them with concrete. A steel cagemay be used for reinforcement. There are manymethods for forming the holes, such as driving of aclosed-end steel pipe, with or without a mandrel.Alternatively, holes may be formed with drills orcontinuous-flight augers. Two common methods ofconstruction are (1) a hole is excavated by drillingbefore placement of concrete to form a bored pile,and (2) a hole is formed with a continuous-flightauger (CFA) and grout is injected into the holeunder pressure through the toe of the hollow augerstem during auger withdrawal. A modification ofthe CFA method is used to create a mixed-in-placeconcrete pile in clean granular sand. There arenumerous other procedures used in constructingcast-in-place concrete piles, most of which are pro-prietary systems.

Advantages of cast-in-place concrete pilesinclude: relatively low cost, fast execution, ease ofadaptation to different lengths, capability for soilsampling during construction at each pile location,possibility of penetrating undesirable hard layers,high load-carrying capacity of large-size piles,and low vibration and noise levels during installa-tion. Construction time is less than that neededfor precast piles inasmuch as cast-in-place piles canbe formed in place to required lengths and with-out having to wait for curing time before installa-tion.

Pile foundations are normally employed wheresubsurface conditions are likely to be unfavorablefor spread footings or mats. If cast-in-place concretepiles are used, such conditions may create concernsabout the structural integrity, bearing capacity, andgeneral performance of the pile foundation. Thereason for this is that the constructed shape and

structural integrity of such piles depend onsubsurface conditions, concrete quality andmethod of placement, quality of work, and designand construction practices, all of which requiretight control. Structural deficiencies may resultfrom degraded or debonded concrete, necking, orinclusions or voids. Unlike pile driving, where theinstallation process itself constitutes a crudequalitative pile-capacity test and hammer-pile-soilbehavior may be evaluated from measurementsmade during driving, methods for evaluating cast-in-place piles during construction are generally notavailable. Good installation procedures and inspec-tion are critical to the success of uncased augured ordrilled piles.

(“Drilled Shafts: Construction Procedures andDesign Methods,” 1999, by Michael W. O’Neil andLymon C. Reese, Report No. FHWA-IF-99-025,Federal Highway Administration, 400 7th StreetSW, Washington, D.C. 20590 (www.fhwa.gov)various publications of The International Associ-ation of Foundation Drilling (ADSC), P.O. Box280379, Dallas, TX 75228.)

7.15.3 Steel Piles

Structural steel H and pipe sections are often usedas piles. Pipe piles may be driven open- or closed-end. After being driven, they may be filled withconcrete. Common sizes of pipe piles range from 8to 48 in in diameter. A special type of pipe pile isthe Monotube, which has a longitudinally flutedwall, may be of constant section or tapered, andmay be filled with concrete after being driven.Closed-end pipes have the advantage that they canbe visually inspected after driving. Open-end pipeshave the advantage that penetration of hard layerscan be assisted by drilling through the open end.

H-piles may be rolled or built-up steel sectionswith wide flanges. Pile toes may be reinforced withspecial shoes for driving through soils withobstructions, such as boulders, or for driving torock. If splicing is necessary, steel pile lengths maybe connected with complete-penetration welds orcommercially available special fittings. H piles,being low-displacement piles, are advantageous insituations where ground heave and lateral move-ment must be kept to a minimum.

Steel piles have the advantages of being rug-ged, strong, and easy to handle. They can be driventhrough hard layers. They can carry high compres-sive loads and withstand tensile loading. Because

Geotechnical Engineering n 7.35

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of the relative ease of splicing and cutting to length,steel piles are advantageous for use in sites wherethe depth of the bearing layer varies. Disadvan-tages of steel piles include small cross-sectionalarea and susceptibility to corrosion, which cancause a significant reduction in load-carryingcapacity. Measures that may be taken when pilecorrosion is anticipated include the use of largerpile sections than otherwise needed, use of surface-coating materials, or cathodic protection. In thesecases, the pipes are usually encased in or filled withconcrete.

Specifications pertaining to steel pipe piles aregiven in “Specification for Welded and SeamlessSteel Pipe Piles,” ASTM A252 (www.astm.org). Fordimensions and section properties of H piles, see“HP Shapes” in “Manual of Steel Construction,”American Institute of Steel Construction, One EastWacker Drive, Suite 3100, Chicago, IL 60601-2001.

7.15.4 Timber Piles

Any of a variety of wood species but usuallysouthern pine or douglas fir, and occasionally redor white oak, can be used as piles. Kept below thegroundwater table, timber piles can serve in apreserved state for a long time. Untreated piles thatextend above the water table, however, may beexposed to damaging marine organisms and decay.Such damage may be prevented or delayed andservice life prolonged by treating timber piles withpreservatives. Preservative treatment shouldmatchthe type of wood.

Timber piles are commonly available in lengthsof up to 75 ft. They should be as straight as possibleand should have a relatively uniform taper. Timberpiles are usually used to carry light to moderateloads or in marine construction as dolphins andfender systems.

Advantages of timber piles include their rela-tively low cost, high strength-to-weight ratio, andease of handling. They can be cut to length afterdriving relatively easily. Their naturally taperedshape (about 1 in in diameter per 10 ft of length) isadvantageous in situations where pile capacitiesderive mostly from shaft resistance. Disadvantagesinclude their susceptibility to damage during harddriving and difficulty in splicing.

Timber piles should be driven with care to avoiddamage. Hammers with high impact velocitiesshould not be used. Protective accessories should

be utilized, when hard driving is expected,especially at the head and toe of the pile.

Specifications relevant to timber piles arecontained in “Standard Specifications for RoundTimber Piles,” ASTM D25; “Establishing DesignStresses for Round Timber Piles,” ASTM D2899(www.astm.org); and “Preservative Treatment byPressure Processes,” AWPA C3, American WoodPreservers Association (www.awpa.com). Infor-mation on timber piles also may be obtained fromthe National Timber Piling Council, Inc., 446 ParkAve., Rye, NY 10580.

7.15.5 Composite Piles

This type of pile includes those made of more thanone major material or pile type, such as thick-walled, concrete-filled, steel pipe piles, precastconcrete piles with steel (pipe or H section) exten-sions, and timber piles with cast-in-place concreteextensions.

7.15.6 Selection of Pile Type

The choice of an appropriate pile type for a partic-ular application is essential for satisfactory foun-dation functioning. Factors that must be consideredin the selection process include subsurface con-ditions, nature and magnitude of loads, localexperience, availability of materials and experi-enced labor, applicable codes, and cost. Pile driv-ability, strength, and serviceability should also betaken into account. Figure 7.15 presents general

Fig. 7.15 Approximate ranges of design loadsfor vertical piles in axial compression.

*For shaft diameters not exceeding 18 in.† Primary end bearing.‡ Permanent shells only.§ Uncased only.

7.36 n Section Seven

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guidelines and approximate ranges of design loadsfor vertical piles in axial compression. Actual loadsthat can be carried by a given pile in a particularsituation should be assessed in accordance with thegeneral methods and procedures presented in thepreceding and those described in more specializedgeotechnical engineering books.

7.16 Pile-Driving Equipment

Installation of piles by driving is a specialized fieldof construction usually performed by experiencedcontractors with dedicated equipment. The basiccomponents of a pile-driving system are shown inFigure 7.16 and described in the following. Allcomponents of the driving system have some effecton the pile-driving process. The overall stabilityand capacity of the pile-driving crane should beassessed for all stages of loading conditions,including pile pickup and driving.

Lead n The functions of the lead (also knownas leader or guide) are to guide the hammer, main-tain pile alignment, and preserve axial alignmentbetween the hammer and pile. For proper func-tioning, leads should have sufficient strength and

be straight and well greased to allow free hammertravel.

There are four main types of leads: swinging,fixed, semifixed, and offshore. Depending on therelative positions of the crane and the pile, pile size,and other factors, a specific type of lead may haveto be employed. Swinging leads are the simplest,lightest in weight, and most versatile. They do not,however, provide much fixity for prevention oflateral pile movement during driving. Fixed leadsmaintain the position of the pile during drivingand facilitate driving a pile at an inclined angle.However, they are the most expensive type of lead.Semifixed leads have some of the advantages anddisadvantages of the swinging and fixed leads.Offshore leads are used mostly in offshore con-struction to drive large-size steel piles and on landor near shore when a template is used to hold thepile in place. Their use for inclined piles is limitedby the pile flexural strength.

Pile Cap (Helmet) n The pile cap (alsoreferred to as helmet) is a boxlike steel elementinserted in the lead between the hammer and pile(Fig. 7.17). The function of the cap is to house bothhammer and pile cushions and maintain axialalignment between hammer and pile. The size ofthe cap needed depends on the pile size and thejaw-opening size of the lead. In some cases, anadaptor is inserted under the cap to accommodatevarious pile sizes, assuring that the hammer and

Fig. 7.16 Basic components of a pile-drivingrig. (From “The Performance of Pile Driving Systems—Inspection Manual,” FHWA/RD-86/160, FederalHigh-way Administration.) Fig. 7.17 Pile helmet and adjoining parts.

Geotechnical Engineering n 7.37

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pile are concentrically aligned. A poor seating ofthe pile in the cap can cause pile damage andbuckling due to localized stresses and eccentricloading at the pile top.

Cushions n Hammers, except for somehydraulic hammers, include a cushion in thehammer (Fig. 7.17). The function of the hammercushion is to attenuate the hammer impact forcesand protect both the pile and hammer fromdamaging driving stresses. Normally, a steel strikerplate, typically 3 in thick, is placed on top of thecushion to insure uniform cushion compression.Most cushions are produced by specialized manu-facturers and consist of materials such as phenolicor nylon laminate sheets.

For driving precast concrete piles, a pile cushionis also placed at the pile top (Fig. 7.17). The mostcommon material is plywood. It is placed in layerswith total thickness between 4 and 12 in. In somecases, hardwood boards may be used (with thegrain perpendicular to the pile axis) as pilecushions. Specifications often require that a freshpile cushion be used at the start of the driving of apile. The wood used should be dry. The pilecushion should be changed when significantlycompressed or when signs of burning are evident.Cushions (hammer and pile) should be durableand reasonably able to maintain their properties.When a change is necessary during driving, thedriving log should record this. The measuredresistance to driving immediately thereafter shouldbe discounted, especially if the pile is being drivenclose to its capacity, inasmuch as a fresh cushionwill compress significantly more from a hammerblow than would an already compressed one.Hence, measurements of pile movement per blowwill be different.

With the aid of a computer analytical program,such as one based on the wave equation, it ispossible to design a cushion system for a particularhammer and pile that allows maximum energytransfer with minimum risk of pile damage.

Hammer n This provides the energy neededfor pile installation. Basically, an impact pile-driving hammer consists of a striking part, calledthe ram, and a means of imposing impacts in rapidsuccession to the pile.

Hammers are commonly rated by the amount ofpotential energy per blow. This energy basically is

the product of ram weight and drop height(stroke). To a contractor, a hammer is a mass-production machine; hammers with higher effi-ciency are generally more productive and canachieve higher pile capacity. To an engineer, ahammer is an instrument that is used to measurethe quality of the end product, the driven pile.Implicit assumptions regarding hammer perform-ance are included in common pile evaluationprocedures. Hammers with low energy transfer arethe source of poor installation. Hence, pile design-ers, constructors, and inspectors should be familiarwith operating principles and performance charac-teristics of the various hammer types. Followingare brief discussions of the major types of impactpile-driving hammers.

Impact pile-driving hammers rely on a fallingmass to create forces much greater than theirweight. Usually, strokes range between 3 and 10 ft.

These hammers are classified by the mode usedin operating the hammer; that is, the means usedto raise the ram after impact for a new blow. Thereare two major modes: external combustion andinternal combustion. Hammers of each type maybe single or double acting. For single-actinghammers (Fig. 7.18), power is only needed to raisethe ram, whereas the fall is entirely by gravity.Double-acting hammers also apply power to assistthe ram during downward travel. Thus thesehammers deliver more blows per minute thansingle-acting hammers; however, their efficiencymay be lower, since the power source supplies partof the impact energy.

External-combustion hammers (ECHs) rely on apower source external to the hammer for theiroperation. One type is the drop hammer, which israised by a hoist line from the crane supportingthe pile and leads and then dropped to fall underthe action of gravity to impact the pile. Mainadvantages of drop hammers are relatively lowcost and maintenance and the ability to vary thestroke easily. Disadvantages include reduction ofthe effectiveness of the drop due to the cable andwinch assembly required for the operation, slowoperation, and hammer efficiency dependence onoperator’s skills. (The operator must allow thecable to go slack when the hammer is raised to dropheight.) Consequently, use of drop hammers isgenerally limited to small projects involving lightlyloaded piles or sheetpiles.

For some pile drivers, hydraulic pressure is usedto raise the ram. The hammers, known as air/steam

7.38 n Section Seven

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or hydraulic hammers, may be single or doubleacting. Action starts with introduction of themotivefluid (steam, compressed air, or hydraulic fluid)under the piston in the hammer chamber to lift theram. When the ram attains a prescribed height,flow of the motive fluid is discontinued and theram “coasts” against gravity up to the full stroke.At top of stroke, the pressure is vented and the ramfalls under gravity. For double-acting hammers, thepressure is redirected to act on top of the piston andpush the ram downward during its fall. Manyhydraulic hammers are equipped with two strokeheights for more flexibility.

The next cycle starts after impact, and the startshould be carefully controlled. If pressure isintroduced against the ram too early, it will slowdown the ram excessively and reduce the energyavailable to the pile. Known as preadmission, thisis not desirable due to its adverse effect on energytransfer. For some hammers, the ram, immediatelypreceding impact, activates a valve to allow the

motive fluid to enter the cylinder to start the nextcycle. For most hydraulic hammers, the ram posi-tion is detected by proximity switches and the nextcycle is electronically controlled.

Main advantages of external combustion ham-mers include their higher rate of operation thandrop hammers, long track record of performanceand reliability, and their relatively simple design.Disadvantages include the need to have additionalequipment on site, such as boilers and compres-sors, that would not be neededwith another type ofhammer. Also disadvantageous is their relativelyhigh weight, which requires equipment with largelifting capacity.

Diesel hammers are internal-combustion ham-mers (ICHs). The power needed for hammeroperation comes from fuel combustion inside thehammer, therefore eliminating the need for anoutside power source. Basic components of a dieselhammer include the ram, cylinder, impact block,and fuel distribution system. Hammer operation isstarted by lifting the ram with one of the hoist linesfrom the crane or a hydraulic jack to a presetheight. A tripping mechanism then releases theram, allowing it to fall under gravity. During itsdescent, the ram closes cylinder exhaust ports, as aresult of which gases in the combustion chamberare compressed. At some point before impact, theram activates a fuel pump to introduce into thecombustion chamber a prescribed amount of fuel ineither liquid or atomized form. The amount of fueldepends on the fuel pump setting.

For liquid-injection hammers, the impact of theram on the impact block atomizes the fuel. Underthe high pressure, ignition and combustion result.For atomized-fuel-injection hammers, ignition oc-curs when the pressure reaches a certain thresholdbefore impact. The ram impact and the explosiveforce of the fuel drive the pile into the groundwhilethe explosion and pile reaction throw the ramupward past the exhaust ports, exhausting thecombustion gases and drawing in fresh air for thenext cycle. With an open-end diesel (OED), shownin Fig. 7.19, the ram continues to travel upwarduntil arrested by gravity. Then the next cycle starts.The distance that the ram travels upward (stroke)depends on the amount of fuel introduced into thechamber (fuel-pump setting), cushions, pile stiff-ness, and soil resistance. In the case of closed-enddiesel hammers (CEDs), the top of the cylinder isclosed, creating an air-pressure, or bounce, cham-ber. Upward movement of the ram compresses the

Fig. 7.18 Single-acting, external-combustionhammer driving a precast concrete pile.

Geotechnical Engineering n 7.39

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air in the bounce chamber and thus stores energy.The pressure shortens ram stroke and the storedenergy accelerates the ram downward.

The energy rating of diesel hammers is com-monly appraised by observing the ram stroke(or bounce-chamber pressure for closed-end ham-mers). This is an important indication but can bemisleading; for example, when the hammer be-comes very hot during prolonged driving. Becausethe fuel then ignites too early, the ram expendsmore energy to compress the gases and less energyis available for transmission into the pile. The highpressure still causes a relatively high stroke. Thiscondition is commonly referred to as preignition. Incontrast, short ram strokes may be caused by lackof fuel or improper fuel type, lack of compressionin the chamber due to worn piston rings, excessiveram friction, pile stiffness, or lack of soil resistance.

Internal-combustion hammers are advanta-geous because they are entirely self-contained.They are relatively lightweight and thus permit useof smaller cranes than those required for external-combustion hammers. Also, stroke adjustment tosoil resistance with internal-combustion hammersis advantageous in controlling dynamic stressesduring driving of concrete piles. Among disad-vantages is stroke dependence on the hammer-pile-soil system, relatively low blow rate, and potentialcessation of operation when easy driving is en-countered.

Table 7.8 presents the characteristics of impactpile-driving hammers. The hammers are listed byrated energy in ascending order. The table indi-cates for each model the type of hammer: ECH,external combustion hammer, or OED, open-enddiesel hammer; manufacturer; model number; ramweight; and equivalent stroke. Note, however, thatnew models of hammers become available at fre-quent intervals.

Vibratory hammers drive or extract piles byapplying rapidly alternating forces to the pile. Theforces are created by eccentric weights (eccentrics)rotating around horizontal axes. The weights areplaced in pairs so that horizontal centrifugal forcescancel each other, leaving only vertical-force com-ponents. These vertical forces shake piles up anddown and cause vertical pile penetration under theweight of the hammer. The vibration may be eitherlow frequency (less than 50 Hz) or high frequency(more than 100 Hz).

The main parameters that define the character-istics of a vibratory hammer are amplitude pro-duced, power consumption, frequency (vibrationsper minute), and driving force (resultant verticalforce of the rotating eccentrics). Vibratory hammersoffer the advantages of fast penetration, limitednoise, minimal shockwaves induced in the ground,and usually high penetration efficiency in cohe-sionless soils. A disadvantage is limited penetra-tion capacity under hard driving conditions and inclay soils. Also, there is limited experience incorrelating pile capacity with driving energy andpenetration rate. This type of hammer is often usedto install non-load-bearing piles, such as retainingsheetpiles.

(“Vibratory Pile Driving,” J. D. Smart, Ph.D.Thesis, University of Illinois, Urbana, 1969; “Drive-ability and Load Transfer Characteristics of Vibro-Driven Piles,” D. Wong, Ph.D Thesis, University ofHouston, Texas, 1988; various publications of the

Fig. 7.19 Open-end, single-acting diesel hammerdriving a pile.

7.40 n Section Seven

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Table

7.8

ImpactPile-DrivingHam

mers

Rated

Energy

kip-ft

Man

ufacturer

Ham

mer

Model

Ram

Weight,

kips

Equivalen

t

Stroke,

ft

Ham

mer

Type

Rated

Energy

kip-ft

Man

ufacturer

Ham

mer

Model

Ram

Weight,

kips

Equivalen

t

Stroke,

ft

Ham

mer

Type

Rated

Energy

kip-ft

Man

ufacturer

Ham

mer

Model

Ram

Weight,

kips

Equivalen

t

Stroke,

ft

Ham

mer

Type

1.00

MKT

No.5

0.20

5.00

ECH

22.13

IHC

Hydh

SC

303.64

6.08

ECH

32.50

VULCAN

VUL010

10.00

3.25

ECH

2.50

MKT

No.6

0.40

6.25

ECH

22.40

MKT

DE30

2.80

8.00

OED

32.55

FAIRCHLD

F-32

10.85

3.00

ECH

4.15

MKT

No.7

0.80

5.19

ECH

22.50

FEC

FEC

1200

2.75

8.18

OED

32.90

VULCAN

VUL100C

10.00

3.29

ECH

7.26

VULCAN

VUL30C

3.00

2.42

ECH

22.50

ICE

30S

3.00

7.50

OED

33.00

MKT33

DE333020

3.30

10.00

OED

7.26

VULCAN

VUL02

3.00

2.42

ECH

22.99

BERMIN

GH

B23

2.80

8.21

CED

33.18

UDDCOMB

H4H

8.80

3.77

ECH

8.10

LIN

KBELT

LB180

1.73

4.68

CED

22.99

BERMIN

GH

B23

52.80

8.21

CED

34.72

BANUT

6Tonnes

13.23

2.62

ECH

8.13

ICE

180

1.73

4.70

CED

23.12

ICE

422

4.00

5.78

CED

34.72

Pilem

erDKH-4

8.82

3.95

ECH

8.23

DELMAG

D5

1.10

7.48

OED

23.14

BANUT

4Tonnes

8.82

2.62

ECH

35.37

JUNTTAN

HHK

48.82

4.01

ECH

8.65

DAWSON

HPH

1200

2.29

3.77

ECH

23.59

DELMAG

D12

2.75

8.58

OED

35.40

BERMIN

GH

B2505

3.00

11.80

OED

8.75

MKT

9B3

1.60

5.47

ECH

23.80

MKT

DA35BSA

2.80

8.50

OED

35.98

VULCAN

VUL140C

14.00

2.57

ECH

10.50

DELMAG

D6-32

1.32

7.94

OED

23.80

MKT

DE30B

2.80

8.50

OED

37.38

CONMACO

C115

11.50

3.25

ECH

13.11

MKT

10B3

3.00

4.37

ECH

24.38

RAYMOND

R0

7.50

3.25

ECH

37.52

MKT

S14

14.00

2.68

ECH

14.20

MKT

C5-Air

5.00

2.84

ECH

24.40

MKT

C826Stm

8.00

3.05

ECH

37.72

ICE

115

11.50

3.28

ECH

15.00

CONMACO

C50

5.00

3.00

ECH

24.48

RAYMOND

R80CH

8.00

3.06

ECH

38.20

MKT

DA

55B

5.00

7.64

CED

15.00

RAYMOND

R1

5.00

3.00

ECH

24.48

RAYMOND

R80C

8.00

3.06

ECH

38.69

MENCK

MHF3-5

11.02

3.51

ECH

15.00

VULCAN

VUL01

5.00

3.00

ECH

24.48

VULCAN

VUL80C

8.00

3.06

ECH

38.69

MENCK

MHF5-5

11.02

3.51

ECH

15.02

LIN

KBELT

LB312

3.86

3.89

CED

24.76

MENCK

MHF3-3

7.05

3.51

ECH

39.00

VULCAN

VUL012

12.00

3.25

ECH

15.10

VULCAN

VUL50C

5.00

3.02

ECH

24.88

UDDCOMB

H3H

6.60

3.77

ECH

39.25

DELMAG

D16-32

3.52

11.15

OED

16.00

MKT

DE20

2.00

8.00

OED

26.00

ICE

32S

3.00

8.67

OED

40.00

CONMACO

C80E5

8.00

5.00

ECH

16.20

MKT

C5-Steam

5.00

3.24

ECH

28.14

MITSUB.

MH

153.31

8.50

OED

40.00

ICE

40-S

4.00

10.00

OED

16.25

MKT

S-5

5.00

3.25

ECH

28.31

DELMAG

D15

3.30

8.58

OED

40.00

MKT

DA55BSA

5.00

8.00

OED

17.32

DAWSON

HPH

2400

4.19

4.13

ECH

28.92

BANUT

5Tonnes

11.02

2.62

ECH

40.00

MKT40

DE333020

4.00

10.00

OED

17.34

BANUT

3Tonnes

6.61

2.62

ECH

29.25

BERMIN

GH

B225

3.00

9.75

OED

40.00

VULCAN

VUL508

8.00

5.00

ECH

17.60

DELMAG

D8-22

1.76

10.00

OED

29.48

IHC

Hydh

SC

405.51

5.35

ECH

40.31

BERMIN

GH

B300

3.75

10.75

OED

18.00

BERMIN

GH

B200

2.00

9.00

OED

30.36

IHC

Hydh

S40

5.51

5.51

ECH

40.31

BERMIN

GH

B300M

3.75

10.75

OED

18.20

LIN

KBELT

LB440

4.00

4.55

CED

30.37

ICE

520

5.07

5.99

CED

40.49

BANUT

7Tonnes

15.43

2.62

ECH

18.56

ICE

440

4.00

4.64

CED

30.41

HERA

1500

3.37

9.02

OED

40.61

DELMAG

D22

4.91

8.27

OED

19.15

MKT

11B3

5.00

3.83

ECH

30.72

MKT

DA

454.00

7.68

CED

40.62

ICE

640

6.00

6.77

CED

19.18

VULCAN

VUL65C

6.50

2.95

ECH

30.80

MKT

MS-350

7.72

3.99

ECH

40.63

RAYMOND

R3/0

12.50

3.25

ECH

19.50

CONMACO

C65

6.50

3.00

ECH

30.96

MENCK

MHF3-4

8.82

3.51

ECH

41.47

UDDCOMB

H5H

11.00

3.77

ECH

19.50

RAYMOND

R65C

6.50

3.00

ECH

31.33

DELMAG

D12-32

2.82

11.11

OED

42.00

CONMACO

C140

14.00

3.00

ECH

19.50

RAYMOND

R1S

6.50

3.00

ECH

32.00

MKT

DE40

4.00

8.00

OED

42.00

ICE

42-S

4.09

10.27

OED

19.50

RAYMOND

R65CH

6.50

3.00

ECH

32.50

CONMACO

C100

10.00

3.25

ECH

42.00

VULCAN

VUL014

14.00

3.00

ECH

19.50

VULCAN

VUL06

6.50

3.00

ECH

32.50

CONMACO

C565

6.50

5.00

ECH

42.40

DELMAG

D19-32

4.00

10.60

OED

19.57

VULCAN

VUL65CA

6.50

3.01

ECH

32.50

HPSI

650

6.50

5.00

ECH

42.50

MKT

DE50B

5.00

8.50

OED

20.00

MKT20

DE333020

2.00

10.00

OED

32.50

MKT

S10

10.00

3.25

ECH

43.01

MITSUB.

M23

5.06

8.50

OED

21.00

MKT

DA

35B

2.80

7.50

CED

32.50

RAYMOND

R2/

010.00

3.25

ECH

43.20

BERMIN

GH

B4004.8

4.80

9.00

OED

21.20

MKT

C826Air

8.00

2.65

ECH

32.50

VULCAN

VUL506

6.50

5.00

ECH

43.37

BSP

HH

511.02

3.94

ECH

(Tablecontinued)

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Table

7.8

(Continued)

Rated

Energy

kip-ft

Man

ufacturer

Ham

mer

Model

Ram

Weight,

kips

Equivalen

t

Stroke,

ft

Ham

mer

Type

Rated

Energy

kip-ft

Man

ufacturer

Ham

mer

Model

Ram

Weight,

kips

Equivalen

t

Stroke,

ft

Ham

mer

Type

Rated

Energy

kip-ft

Man

ufacturer

Ham

mer

Model

Ram

Weight,

kips

Equivalen

t

Stroke,

ft

Ham

mer

Type

43.40

Pilem

erDKH-5

11.02

3.95

ECH

53.05

JUNTTAN

HHK

613.23

4.01

ECH

69.50

BSP

HH

817.64

3.94

ECH

44.00

HPSI

110

11.00

4.00

ECH

53.75

BERMIN

GH

B400

5.00

10.75

OED

69.65

MENCK

MHF5-9

19.84

3.51

ECH

44.00

MKT

MS500

11.00

4.00

ECH

54.17

MENCK

MHF3-7

15.43

3.51

ECH

70.00

MKT70

DE70/50B

7.00

10.00

OED

44.23

JUNTTAN

HHK

511.03

4.01

ECH

54.17

MENCK

MHF5-7

15.43

3.51

ECH

70.96

HERA

3500

7.87

9.02

OED

44.24

IHC

Hydh

SC60

7.72

5.73

ECH

55.99

FEC

FEC

2800

6.16

9.09

OED

72.18

KOBE

K35

7.72

9.35

OED

44.26

IHC

Hydh

S.60

13.23

3.35

ECH

56.77

HERA

2800

6.29

9.02

OED

72.60

ICE

1070

10.00

7.26

CED

45.00

BERMIN

GH

B4005.0

5.00

9.00

OED

56.88

RAYMOND

R5/

017.50

3.25

ECH

73.00

FEC

FEC

3400

7.48

9.76

OED

45.00

FAIRCHLD

F-45

15.00

3.00

ECH

57.50

CONMACO

C115E

511.50

5.00

ECH

73.66

DELMAG

D30-32

6.60

11.16

OED

45.07

MENCK

MRBS500

11.02

4.09

ECH

58.90

IHC

Hydh

SC

8011.24

5.24

ECH

73.66

DELMAG

D30-23

6.60

11.16

OED

45.35

KOBE

K22-Est

4.85

9.35

OED

59.00

BERMIN

GH

B4005

5.00

11.80

OED

75.00

RAYMOND

R30X

30.00

2.50

ECH

46.43

MENCK

MHF5-6

13.23

3.51

ECH

59.50

MKT

DE70B

7.00

8.50

OED

77.39

MENCK

MHF5-10

22.04

3.51

ECH

46.43

MENCK

MHF3-6

13.23

3.51

ECH

59.60

DELMAG

D30

6.60

9.03

OED

77.42

IHC

Hydh

SC

110

15.21

5.09

ECH

46.84

MITSUB.

MH

255.51

8.50

OED

60.00

CONMACO

C200

20.00

3.00

ECH

77.88

BERMIN

GH

B4505

6.60

11.80

OED

47.20

BERMIN

GH

B3505

4.00

11.80

OED

60.00

HPSI

1500

15.00

4.00

ECH

78.17

BSP

HH

919.84

3.94

ECH

48.50

DELMAG

D22-13

4.85

10.00

OED

60.00

ICE

60S

7.00

8.57

OED

80.00

HPSI

2000

20.00

4.00

ECH

48.50

DELMAG

D22-02

4.85

10.00

OED

60.00

MKT

S20

20.00

3.00

ECH

80.00

ICE

80S

8.00

10.00

OED

48.75

CONMACO

C160

16.25

3.00

ECH

60.00

VULCAN

VUL320

20.00

3.00

ECH

80.41

MITSU8.

M43

9.46

8.50

OED

48.75

RAYMOND

R4/0

15.00

3.25

ECH

60.00

VULCAN

VUL512

12.00

5.00

ECH

81.25

RAYMOND

R8/0

25.00

3.25

ECH

48.75

RAYMOND

R150C

15.00

3.25

ECH

60.00

VULCAN

VUL020

20.00

3.00

ECH

83.82

DELMAG

D36-13

7.93

10.57

OED

48.75

VULCAN

VUL016

16.25

3.00

ECH

60.76

Pilem

erDKH-7

15.43

3.95

ECH

83.82

DELMAG

D36-02

7.93

10.57

OED

49.18

MENCK

MH

687.72

6.37

ECH

60.78

BSP

HH

715.43

3.94

ECH

83.82

DELMAG

D36

7.93

10.57

OED

49.76

UDDCOMB

H6H

13.20

3.77

ECH

61.49

DELMAG

D25-32

5.51

11.16

OED

85.13

MENCK

MHF5-11

24.25

3.51

ECH

50.00

CONMACO

C100E

510.00

5.00

ECH

61.71

MITSUB.

M33

7.26

8.50

OED

85.43

MITSUB.

MH

4510.05

8.50

OED

50.00

FEC

FEC

2500

5.50

9.09

OED

61.91

JUNTTAN

HHK

715.44

4.01

ECH

86.80

Pilem

erDKH-10

22.04

3.95

ECH

50.00

HPSI

1000

10.00

5.00

ECH

61.91

MENCK

MHF5-8

17.64

3.51

ECH

86.88

UDDCOMB

H10H

22.05

3.94

ECH

50.00

MKT50

DE70/50B

5.00

10.00

OED

62.50

CONMACO

C125E

512.50

5.00

ECH

88.00

BERMIN

GH

B550C

11.00

8.00

OED

50.00

VULCAN

VUL510

10.00

5.00

ECH

63.03

FEC

FEC

3000

6.60

9.55

OED

88.42

JUNTTAN

HHK

1022.05

4.01

ECH

50.20

VULCAN

VUL200C

20.00

2.51

ECH

64.00

ICE

160

16.00

4.00

ECH

88.50

DELMAG

D36-32

7.93

11.16

OED

50.69

HERA

2500

5.62

9.02

OED

65.62

MITSUB.

MH

357.72

8.50

OED

88.50

DELMAG

D36-23

7.93

11.16

OED

51.26

DELMAG

D22-23

4.85

10.57

OED

66.00

DELMAG

D30-02

6.60

10.00

OED

90.00

HPSI

225

22.50

4.00

ECH

51.52

KOBE

K25

5.51

9.35

OED

66.00

DELMAG

D30-13

6.60

10.00

OED

90.00

ICE

90-S

9.00

10.00

OED

51.63

ICE

660

7.57

6.82

CED

66.36

IHC

Hydh

S90

9.92

6.69

ECH

90.00

VULCAN

VUL330

30.00

3.00

ECH

51.63

LIN

KBELT

LB660

7.57

6.82

CED

67.77

MENCK

MRBS750

16.53

4.10

ECH

90.00

VULCAN

VUL030

30.00

3.00

ECH

51.65

IHC

Hydh

S70

7.72

6.69

ECH

69.34

UDDCOMB

H8H

17.60

3.94

ECH

90.44

DELMAG

D44

9.50

9.52

OED

51.78

CONMACO

160**

17.26

3.00

ECH

69.43

MENCK

MH

9611.02

6.30

ECH

92.04

BERMIN

GH

B5005

7.80

11.80

OED

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92.75

KOBE

K45

9.92

9.35

OED

138.87

Pilem

erDKH-16

35.27

3.95

ECH

262.11

MENCK

MRBS250

63.93

4.10

ECH

92.87

MENCK

MHF5-12

26.45

3.51

ECH

141.12

MENCK

MH

195

22.05

6.40

ECH

289.55

MENCK

MHU

400

50.71

5.71

ECH

93.28

MENCK

MRBS850

18.96

4.92

ECH

147.18

IHC

Hydh

S200

22.00

6.69

ECH

294.60

IHC

Hydh

S400

44.30

6.65

ECH

93.28

MENCK

MRBS800

18.98

4.92

ECH

147.38

MENCK

MHUT200

26.46

5.57

ECH

295.12

MENCK

MHUT400

52.70

5.60

ECH

95.54

BSP

HH

1124.25

3.94

ECH

149.60

MITSUB.

MH

80B

17.60

8.50

OED

300.00

VULCAN

VUL3100

100.00

3.00

ECH

96.53

DELMAG

D46-13

10.14

9.52

OED

150.00

CONMACO

C5300

30.00

5.00

ECH

300.00

VULCAN

VUL560

62.50

4.80

ECH

100.00

CONMACO

C5200

20.00

5.00

ECH

150.00

MKT150

DE110150

15.00

10.00

OED

325.36

MENCK

MRBS300

66.13

4.92

ECH

100.00

ICE

100

10.00

10.00

OED

150.00

RAYMOND

R60X

60.00

2.50

ECH

347.16

BSP

HA

4088.18

3.94

ECH

100.00

RAYMOND

R40X

40.00

2.50

ECH

150.00

VULCAN

VUL530

30.00

5.00

ECH

350.00

CONMACO

C5700

70.00

5.00

ECH

100.00

VULCAN

VUL520

20.00

5.00

ECH

151.00

IHC

Hydh

SC

200

30.20

5.00

ECH

368.30

IHC

Hydh

S500

55.30

6.66

ECH

101.37

HERA

5000

11.24

9.02

OED

152.06

HERA

7500

16.85

9.02

OED

368.55

MENCK

MHUT500

59.54

6.19

ECH

103.31

IHC

Hydh

SC

150

24.25

4.26

ECH

152.45

DELMAG

D62-02

13.66

11.16

OED

433.64

MENCK

MHU

600

77.16

5.62

ECH

104.80

MENCK

MH

145

16.53

6.34

ECH

152.45

DELMAG

D62-22

13.66

11.16

OED

498.94

MENCK

MRBS460

101.41

4.92

ECH

106.10

JUNTTAN

HHK

1226.46

4.01

ECH

152.45

DELMAG

D62-12

13.66

11.16

OED

500.00

VULCAN

VUL5100

100.00

5.00

ECH

106.20

BERMIN

GH

B5505

9.00

11.80

OED

154.32

HPSI

3005

30.86

5.00

ECH

510.00

CONMACO

C6850

85.00

6.00

ECH

107.18

DELMAG

D46-02

10.14

10.57

OED

154.69

MENCK

MHF10-20

44.07

3.51

ECH

513.34

MENCK

MRBS390

86.86

5.91

ECH

107.18

DELMAG

D46

10.14

10.57

OED

170.00

ICE

205S

20.00

8.50

OED

516.13

MENCK

MHUT700

92.83

5.56

ECH

107.18

DELMAG

D46-23

10.14

10.57

OED

173.58

BSP

HH

20S

44.09

3.94

ECH

542.33

MENCK

MRBS500

110.23

4.92

ECH

110.00

MKT110

DE110150

11.00

10.00

OED

173.60

Pilem

erDKH-20

44.10

3.95

ECH

589.85

IHC

Hydh

S800

81.57

7.23

ECH

112.84

Pilem

erDKH-13

28.66

3.95

ECH

173.58

KOBE

KB80

17.64

9.84

OED

619.18

MENCK

MHUT700

92.83

6.67

ECH

113.16

DELMAG

D46-32

10.14

11.16

OED

176.32

HPSI

3505

35.26

5.00

ECH

631.40

MENCK

MRBS700

154.00

4.10

ECH

113.60

VULCAN

VUL400C

40.00

2.84

ECH

176.97

IHC

Hydh

SC

250

39.24

4.51

ECH

736.91

MENCK

MHUT100

132.30

5.57

ECH

115.57

HERA

5700

12.81

9.02

OED

178.42

HERA

8800

19.78

9.02

OED

737.26

IHCHydh

S1000

101.41

7.27

ECH

116.04

MENCK

MHF10-15

33.06

3.51

ECH

179.16

VULCAN

VUL600C

60.00

2.99

ECH

737.70

MENCK

MHU

1000

126.97

5.81

ECH

120.00

ICE

120S

12.00

10.00

OED

180.00

VULCAN

VUL360

60.00

3.00

ECH

750.00

VULCAN

VUL5150

150.00

5.00

ECH

120.00

VULCAN

VUL340

40.00

3.00

ECH

180.00

VULCAN

VUL060

60.00

3.00

ECH

759.23

MENCK

MRBS600

132.27

5.74

ECH

120.00

VULCAN

VUL040

40.00

3.00

ECH

184.64

IHC

Hydh

S250

27.60

6.69

ECH

867.74

MENCK

MRBS800

176.37

4.92

ECH

121.59

BSP

HH

1430.86

3.94

ECH

186.24

DELMAG

D80-12

17.62

10.57

OED

954.53

MENCK

MRBS880

194.01

4.92

ECH

123.43

MENCK

MRBS110

24.25

5.09

ECH

189.81

MENCK

MRBS180

38.58

4.92

ECH

1177.80

IHC

Hydh

S1600

156.00

7.55

ECH

123.79

JUNTTAN

HHK

1430.87

4.01

ECH

196.64

DELMAG

D80-23

17.62

11.16

OED

1228.87

MENCK

MHU

1700

207.23

5.93

ECH

124.53

DELMAG

D55

11.86

10.50

OED

200.00

VULCAN

VUL540

40.90

4.89

ECH

1547.59

MENCK

MHU

2100

255.80

6.05

ECH

125.70

HERA

6200

13.93

9.02

OED

206.51

IHC

Hydh

S280

29.80

6.93

ECH

1581.83

MENCK

MBS12500

275.58

5.74

ECH

130.18

KOBE

KB60

13.23

9.84

OED

225.00

CONMACO

C5450

45.00

5.00

ECH

1694.97

IHC

Hydh

S2300

226.60

7.48

ECH

132.50

ICE

120S

-15

15.00

8.83

OED

225.95

MENCK

MRBS250

55.11

4.10

ECH

1800.00

VULCAN

VUL6300

300.00

6.00

ECH

135.15

MITSUB.

MH

72B

15.90

8.50

OED

225.95

MENCK

MRBS250

55.11

4.10

ECH

2171.65

MENCK

MHU

3000

363.76

5.97

ECH

135.59

MENCK

MRBS150

33.07

4.10

ECH

245.85

DELMAG

D100-13

22.03

11.16

OED

2210.12

IHC

Hydh

S3000

332.00

6.66

ECH

138.87

BSP

HH

1635.27

3.94

ECH

260.37

BSP

HA

3066.13

3.94

ECH

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Deep Foundations Institute, 120 Charolette Place,Englewood Cliffs, NJ 07632).

Other Pile Driving Accessories n In addi-tion to the basic equipment discussed in the prece-ding, some pile driving requires employment ofspecial accessories, such as an adaptor, follower,mandrel, auger, or water jet.

An adaptor is inserted between a pile helmetand pile head to make it possible for one helmet toaccommodate different pile sizes.

A follower is usually a steel member used toextend a pile temporarily in cases where it isnecessary to drive the pile when the top is belowground level or under water. For efficiency in trans-mitting hammer energy to the pile, stiffness of thefollower should be nearly equal to that of the pile.The follower should be integrated into the drivingsystem so that it maintains axial alignment betweenhammer and pile.

Mandrels are typically used to drive steel shellsor thin-wall pipes that are later filled with concrete.A mandrel is a uniform or tapered, round steeldevice that is inserted into a hollow pile to serve asa rigid core during pile driving.

Water jets or augers are sometimes needed toadvance a pile tip through some intermediatesoil layers. Jet pipes may be integrated into thepile shaft or may be external to the pile. Al-though possibly advantageous in assisting in pilepenetration, jetting may have undesirable effectson pile capacity (compression and particularlyuplift) that should be considered by the engineer.

(Department of Transportation Federal High-way Administration, “The Performance of PileDriving Systems: Inspection Manual,” FHWAReport No. FHWA/RD-86/160, National TechnicalInformation Service, Springfield, VA 22161, www.ntis.gov.)

7.17 Vibration and Noise

With every hammer blow, pile driving producesvibration and noise effects that may extend a longdistance from the source. Hundreds, or eventhousands, of hammer blows are typically neededto drive a single pile. These environmental factorsare increasingly becoming an issue concerning piledriving activities, particularly in urban areas.

Careful planning and execution of pile driving canlimit the potential for real damage, and also forlitigation based on human perception.

A significant portion of the pile driving hammerenergy radiates from the pile into the surroundingground and propagates away as seismic stresswaves. The nature (transient or steady state) andcharacteristics (frequency, amplitude, velocity,attenuation, etc.) of these traveling waves dependon the type and size of hammer (impact or vib-ratory), pile (displacement or non-displacement,impedance, length etc.), and subsurface soil con-ditions. The resulting motion can adversely affectstructures above or below ground surface, under-ground utilities, sensitive equipment and pro-cesses, and annoy the public. Structural damagemay range from superficial plaster cracking tofailure of structural elements. Experience hasshown that damage to structures is not likely tooccur at a distance greater than the pile embeddeddepth, or 50 feet minimum (Wood, 1997). Insituations where liquefaction or shakedown settle-ment of loose granular materials may occur, piledriving vibration effects may extend to more than1000 feet. The following information should berecorded as part of a survey of the pile driving siteand surrounding area: distance to the neareststructure or underground utility, function andcondition of nearby structures and facilities, andground conditions of site and vicinity. The FloridaDepartment of Transportation, for example,requires the monitoring of structures for settlementby recording elevations to 0.001 foot within adistance, in feet, of pile driving operations equal to0.5 times the square root of the hammer energy, infoot-pounds.

People are much more sensitive to groundvibrations than structures. Since they can becomeannoyed with vibrations that are only 1/100th ofthose that might be harmful to most structures,human sensitivity should not be used as a measureof vibration for engineering purposes. Vibrationlimits to prevent damage and human discomfortare not clearly and firmly established. Pile drivingvibrations are typically measured by monitoringground peak particle velocity (ppv) using specialtyequipment. Published limits range from 0.2 inch/second (for historical buildings) to 2 inch/second(for industrial structures); the Florida Departmentof Transportation typically uses 0.5 inch/second asa limiting value. The prediction of the vibrationlevel which may be induced for a particular com-

7.44 n Section Seven

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bination of hammer, pile, and soil is fraught withdifficulties, nevertheless, the literature containsseveral equations for computing predicted peakparticle velocity (Wiss, 1981). Vibration mitigationmeasures include: active isolation screening bymeans of a wave barrier near the pile drivinglocation, passive isolation screening by means of awave barrier near the target affected structure, andpile driving operation controls (jetting, predrilling,change of piling system, hammer type, pile topcushions, and driving sequence). Wave barriers(trenches or sheet pile wall) are attractive, but theyare expensive and difficult to design and imple-ment. Design parameters are available in thetechnical literature (Wood, 1968).

Noise annoys, frustrates and angers people.During the last 30 years there has been increasingconcern with the quality of the environment.Through the Noise Control Act of 1972, theUnited States Congress directed the Environmen-tal Protection Agency (EPA) to publish infor-mation about all identifiable effects of noise andto define acceptable levels which would protectthe public health and welfare. Impact pile drivingis inherently noisy, perhaps the noisiest of allconstruction operations. Noise is an environmen-tal issue in populated areas, and should be ofconcern to those involved in pile drivingactivities. The usual unit of sound measurementis the decibel (dB), and one decibel is the lowestvalue a normal human ear can detect. The ear alsoregisters pitch in addition to loudness. It issensitive to frequencies of 0.01 to 16 kHz andmost responsive at about 1 to 5 kHz. Measuringdevices take this into account, by filteringcomponents outside the most responsive range,and express results in dB(A). Thus, soundintensity, noise, is typically recorded in dB(A).The scale is logarithmic. Apparent loudnessdoubles for each 10 decibel increment. Typicalvalues are: noisy factory 90 dB(A), busy street85 dB(A), radio at full volume 70 dB(A), andnormal speech 35 dB(A). Most people becomeannoyed by steady levels above 55 dB(A). Levelsbelow 80 dB(A) would not cause hearing loss,sustained exposure to levels above 90 dB(A) causephysical and mental discomfort and result inpermanent hearing damage.

Sources of noise in pile driving operationsinclude hammer (or related equipment) exhaust,impact of hammer, and noise radiating from thepile itself. At distances of 10 to 100 feet, pile

driving generally produce levels between 75 to115 dB(A). Typically, a distance of about 300 feetwould be needed for the noise level to be belowthe OSHA allowed 8-hour exposure (90 dB(A)),and the sound from the noisiest hammer/pilewould have to travel miles before decreasing tobelow moderately annoying levels. Sound leveldrops by about 6 dB(A) as the distance doublesfrom the source.

Acoustic shrouds or curtain enclosures havebeen successfully employed to reduce the piledriving noise, by 15 to 30 dB(A), to below annoyinglevels. A reduction of 30 dB(A) would make thenoisiest pile driving operation acceptable to mostpeople located farther than 500 feet from the piledriving operation.

Smoke is another environmental factor tobe considered in planning and executing a piledriving project, especially in metropolitan areas.Sources of smoke are the exhaust from the hammeritself (diesel hammers), the external combustionequipment such as boiler (steam hammers),compressor (air hammers), or power pak (hydrau-lic hammers). Some modern internal combustionhammers use environmentally friendly fuels.

(Wood, R. D., “Dynamic Effects of Pile Installa-tions on Adjacent Structures”, Synthesis of High-way Practice No. 253, National CooperativeHighway Research Program, National AcademyPress,Washington,D.C., 1997, 86pages;Wood, R.D.(1968), “Screening of elastic waves by trenches”,ASCE Journal of the Soil Mechanics and Foun-dations Division, vol. 94, pp 951–979; Wiss, J. F.(1981), “Construction vibrations; state-of-the-art”,ASCE Journal of Geotechnical Engineering,vol. 107, No. GT2, pp 167–181, www. asce.org.)

7.18 Pile-Design Concepts

Methods for evaluating load-carrying capacity andgeneral behavior of piles in a foundation rangefrom simple empirical to techniques that incorpor-ate state-of-the-art analytical and field verifi-cation methods. Approaches to pile engineeringinclude (1) precedence, (2) static-load analysis (3)static-load testing, and (4) dynamic-load analyticaland testing methods. Regardless of the methodselected, the foundation designer should possessfull knowledge of the site subsurface conditions.This requires consultations with a geotechnicalengineer and possibly a geologist familiar with the

Geotechnical Engineering n 7.45

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local area to ensure that a sufficient number ofborings and relevant soil and rock tests areperformed.

Design by precedent includes application ofbuilding code criteria, relevant published data,performance of similar nearby structures, andexperience with pile design and construction.Under some circumstances this approach may beacceptable, but it is not highly recommended.Favorable situations include those involving minorand temporary structures where failure would notresult in appreciable loss of property or any loss oflife and construction sites where long-term experi-ence has been accumulated and documented fora well-defined set of subsurface and loadingconditions.

Static-load analysis for design and predictionof pile behavior is widely used by designers whoare practitioners of geotechnical engineering. Thisapproach is based on soil-mechanics principles,geotechnical engineering theories, pile character-istics, and assumptions regarding pile-soil inter-action. Analysis generally involves evaluations ofthe load-carrying capacity of a single pile, of pile-group behavior, and of foundation settlementunder service conditions. Designs based on thisapproach alone usually incorporate relatively largefactors of safety in determination of allowableworking loads. Safety factors are based on theengineer’s confidence in parameters obtained fromsoil exploration and their representation of thewhole site, anticipated loads, importance of thestructure, and the designer’s experience and sub-jective preferences. Static-load analysis methodsare used in preliminary design to calculaterequired pile lengths for cost estimation andbidding purposes. Final pile design and accept-ability are based on additional methods of verifica-tion. Pile-group behavior and settlement predic-tions are, however, usually based entirely on staticanalyses due to the lack of economical and efficientroutine field verification techniques.

Field testing should be performed on a sufficientnumber of piles to confirm or revise initial designassumptions, verify adequacy of installation equip-ment and procedures, evaluate the effect ofsubsurface profile variations, and form the basisof final acceptance. Traditionally, piles were testedwith a static-load test (by loading in axialcompression, uplift, or laterally). The number ofsuch tests that will be performed on a site islimited, however, due to the expense and time

required when large number of piles are to beinstalled.

See Art. 7.19 for a description of static-loadanalysis and pile testing.

Dynamic-load pile testing and analysis is oftenused in conjunction with or as an alternative tostatic testing. Analytical methods utilizing compu-ters and numerical modeling and based on one-dimensional elastic-wave-propagation theories arehelpful in selecting driving equipment, assessingpile drivability, estimating pile bearing capacity,and determining required driving criteria, that is,blow count. Dynamic analysis is commonly knownas wave equation analysis of pile driving. Fielddynamic testing yields information on driving-system performance of piles: static axial capacity,driving stresses, structural integrity, pile-soil in-teraction, and load-movement behavior.

See Art. 7.20 for a description of dynamicanalysis and pile testing.

(Manual for “Design and Construction ofDriven Pile Foundations,” 1996, Federal HighwayAdministration, Report No. FHWA-H1-96-033,Federal Highway Administration, 400 7th StreetSW, Washington, D.C. 20590)

7.19 Static-Analysis and PileTesting

Static analysis of piles and pile design based on itcommonly employ global factors of safety. Use ofthe load-and-resistance-factors approach is grow-ing, however. Steps involved in static analysisinclude calculation of the static load-carryingcapacity of single piles, evaluation of groupbehavior, and assessment of foundation settlement.Normally, capacity and settlement are treatedseparately and either may control the design. Piledrivability is usually treated as a separate item andis not considered in static-load analysis. See alsoArt. 7.18.

7.19.1 Axial-Load Capacity of SinglePiles

Pile capacity Qu may be taken as the sum of theshaft and toe resistances, Qsu and Qbu respectively.The allowable load Qa may then be determined

7.46 n Section Seven

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from either Eq. (7.35) or (7.36)

Qa ¼ Qsu þQbu

F(7:35)

Qa ¼ Qsu

F1þQbu

F2(7:36)

where F, F1, F2, are safety factors. Typically F forpermanent structures is between 2 and 3 but maybe larger, depending on the perceived reliability ofthe analysis and construction as well as the con-sequences of failure. Equation (7.36) recognizesthat the deformations required to fully mobilizeQsu

and Qbu are not compatible. For example, Qsu maybe developed at displacements less than 0.25 in,whereas Qbu may be realized at a toe displacementequivalent to 5% to 10% of the pile diameter.Consequently, F1 may be taken as 1.5 and F2 as 3.0,if the equivalent single safety factor equals F orlarger. (If Qsu/Qbu , 1.0, F is less than the 2.0usually considered as a minimum safety factor forpermanent structures.)

7.19.2 Shaft Resistance inCohesive Soils

The ultimate stress �ff s of axially loaded piles in co-hesive soils under compressive loads is conven-

tionally evaluated from the ultimate frictionalresistance

Qsu ¼ As�ff s ¼ Asa�ccu (7:37)

where cu ¼ average undrained shear strength ofsoil in contact with shaft surface

As ¼ shaft surface area

a ¼ shear-strength (adhesion) reductionfactor

One relationship for selection of a is shown inFig. 7.20. This and similar relationships are em-pirical and are derived from correlations of load-test data with the cu of soil samples tested in thelaboratory. Some engineers suggest that �ff s isinfluenced by pile length and that a limiting valueof 1 ton/ft2 be set for displacement piles less than50 ft long and reduced 15% for each 50 ft ofadditional length. This suggestion is rejected byother engineers on the presumption that it neglectsthe effects of pile residual stresses in evaluation ofthe results of static-load tests on piles.

The shaft resistance stress �ff s for cohesive soilsmay be evaluated from effective-stress concepts:

�ff s ¼ bs 0vo (7:38)

Fig. 7.20 Variation of shear-strength (adhesion) reduction factor a with undrained shear strength.(After “Recommended Practice for Planning, Designing, and Constructing Fixed Off-Shore Platforms,” AmericanPetroleum Institute, Dallas.)

Geotechnical Engineering n 7.47

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where s 0vo ¼ effective overburden pressure of soil

b ¼ function of effective friction angle,stress history, length of pile, andamount of soil displacement inducedby pile installation

b usually ranges between 0.22 and 0.35 forintermediate-length displacement piles driven innormally consolidated soils, whereas for pilessignificantly longer than 100 ft, b may be as smallas 0.15. Derivations of b are given by G. G.Meyerhof, “Bearing Capacity and Settlement ofPile Foundations,” ASCE Journal of GeotechnicalEngineering Division, vol. 102, no. GT3, 1976;J. B. Burland, “Shaft Friction of Piles in Clay,”Ground Engineering, vol. 6, 1973; “Soil Capacity forSupporting Deep Foundation Members in Clay,”STP 670, ASTM.

Both the a and b methods have been applied inanalysis of n discrete soil layers:

Qsu ¼Xni¼1

Asi�ff si (7:39)

The capacity of friction piles driven in cohesivesoils may be significantly influenced by the elapsedtime after pile driving and the rate of load ap-plication. The frictional capacity Qs of displace-ment piles driven in cohesive soils increases withtime after driving. For example, the pile capacityafter substantial dissipation of pore pressuresinduced during driving (a typical design assump-tion) may be three times the capacity measuredsoon after driving. This behavior must be con-sidered if piles are to be rapidly loaded shortlyafter driving and when load tests are interpreted.

Some research indicates that frictional capacityfor tensile load Qut may be less than the shaftfriction under compression loading Qsu. In theabsence of load-test data, it is therefore appropriateto take Qut, as 0.80Qsu and ignore the weight of thepile. Also, Qut is fully developed at average piledeformations of about 0.10 to 0.15 in, about one-half those developed in compression. Expanded-base piles develop additional base resistance andcan be used to substantially increase uplift resis-tances.

(V. A. Sowa, “Cast-In-Situ Bored Piles,” Cana-dianGeotechnical Journal, vol. 7, 1970; G.G.Meyerhofand J. I. Adams, “The Ultimate Uplift Capacity ofFoundations,” Canadian Geotechnical Journal, vol. 5,no. 4, 1968.)

7.19.3 Shaft Resistance inCohesionless Soils

The shaft resistance stress �ff s is a function of the soil-shaft friction angle d, deg, and an empirical lateralearth-pressure coefficient K:

�ff s ¼ K �ss 0vo tan d � fl (7:40)

At displacement-pile penetrations of 10 to 20pile diameters (loose to dense sand), the averageskin friction reaches a limiting value fl. Primarilydepending on the relative density and texture ofthe soil, fl has been approximated conservatively byusing Eq. (7.40) to calculate �ff s. This approachemploys the same principles and involves the samelimitations discussed in Art. 7.8.2.

For relatively long piles in sand, K is typicallytaken in the range of 0.7 to 1.0 and d is taken to beabout f 2 5, where f0 is the angle of internalfriction, deg. For piles less than 50 ft long, K is morelikely to be in the range of 1.0 to 2.0 but can begreater than 3.0 for tapered piles.

Empirical procedures have also been used toevaluate �ff s from in situ tests, such as cone penetra-tion, standard penetration, and relative density tests.Equation (7.41), based on standard penetration tests,as proposed by Meyerhof, is generally conservativeand has the advantage of simplicity.

�ff s ¼�NN

50(7:41)

where �NN ¼ average average standard penetrationresistance within the embedded length of pile and �ff sis given in tons/ft2. (G. G. Meyerhof, “BearingCapacity and Settlement of Pile Foundations,” ASCEJournal of Geotechnical Engineering Division, vol. 102,no. GT3, 1976.)

7.19.4 Toe Capacity Load

For piles installed in cohesive soils, the ultimate toeload may be computed from

Qbu ¼ Abq ¼ AbNccu (7:42)

where Ab ¼ end-bearing area of pile

q ¼ bearing capacity of soil

Nc ¼ bearing-capacity factor

cu ¼ undrained shear strength of soil withinzone 1 pile diameter above and 2diameters below pile tip

7.48 n Section Seven

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Although theoretical conditions suggest that Nc

may vary between about 8 and 12, Nc is usuallytaken as 9.

For cohesionless soils, the toe resistance stress qis conventionally expressed by Eq. (7.43) in termsof a bearing-capacity factor Nq and the effectiveoverburden pressure at the pile tip s 0

vo.

q ¼ Nqs0vo � ql (7:43)

Some research indicates that, for piles in sands, q,like �ff s, reaches a quasi-constant value ql after pen-etrations of the bearing stratum in the range of 10 to20 pile diameters. Approximately,

ql ¼ 0:5Nq tanf (7:44)

where f is the friction angle of the bearing soilsbelow the critical depth. Values of Nq applicable topiles are given in Fig. 7.21. Empirical correlations ofCPT data with q and ql have also been applied topredict successfully end-bearing capacity of pilesin sand. (G. G. Meyerhof, “Bearing Capacity andSettlement of Pile Foundations,” ASCE Journal ofGeotechnical Engineering Division, vol. 102, no.GT3, 1976.)

7.19.5 Pile Settlement

Prediction of pile settlement to confirm allowableloads requires separation of the pile load into shaft

friction and end-bearing components. Since q and�ff s at working loads and at ultimate loads aredifferent, this separation can only be qualitativelyevaluated from ultimate-load analyses. Avariety ofmethods for settlement analysis of single piles havebeen proposed, many of which are empirical orsemiempirical and incorporate elements of elasticsolutions.

(H. Y. Fang, “Foundation Engineering Hand-book,” 2nd ed., VanNostrandReinhold,NewYork.)

7.19.6 Groups of Piles

A pile group may consist of a cluster of piles orseveral piles in a row. The response of an indi-vidual pile in a pile group, where the piles aresituated close to one another, may be influenced bythe response and geometry of neighboring piles.Piles in such groups interact with one anotherthrough the surrounding soil, resulting in what iscalled the pile-soil-pile interaction, or group effect.The efficiency of a pile group (hg) is defined as theratio of the actual capacity of the group to thesummation of the capacities of the individual pilesin the group when tested as single piles. The pile-soil-pile interaction has two components: pileinstallation, and loading effects. Analytical modelsdeveloped to analyze the pile-soil-pile interactionby considering strain superposition in the soil massneglect the effect of installation and the alterationof the failure zone around an individual pile bythose of neighboring piles.

In loose sand, the group efficiency in com-pression exceeds unity, with the highest valuesoccurring at a pile center to center spacing(s) todiameter or width (d) ratio (s/d) of 2. Generally,higher efficiencies occur with an increase in thenumber of piles in the group. However, in densesand, efficiency may be either greater or less thanunity, although the trend is toward hg . 1. Anefficiency smaller than one is probably due todilatancy and would generally be expected forbored or partially jetted piles. Conventionalpractice for the analysis of pile groups in sand hasbeen based on assigning a conservative upperbound for hg of unity for driven piles and 0.67 forbored piles.

Piles in clay always yield values of groupefficiencies less than unity with a distinctive trendtoward block failure in square groups with an (s/d)ratio of less than 2. Historically, the geotechnicalpractice was based on a value of hg of unity for pile

Fig. 7.21 Bearing-capacity factor for granularsoils related to angle of internal friction.

Geotechnical Engineering n 7.49

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groups in clay, provided that block failure does notoccur and that sufficient time has elapsed betweeninstallation and the first application of load topermit excess pore pressure to dissipate.

Several efficiency formulas have been publishedin the literature. These formulas are mostly basedon relating the group efficiency to the spacingbetween the piles and generally yield efficiencyvalues of less than unity, regardless of the pile/soilconditions. A major apparent shortcoming in mostof the efficiency formulas is that they do notaccount for the characteristics of the soil in contactwith the pile group. Comparison of differentefficiency formulas show considerable differencein their results. There is no comprehensive mathe-matical model available for computing the effi-ciency of a pile group. Any general group efficiencyformula that only considers the planar geometry ofthe pile group should be considered with caution.Soil characteristics, time-dependent effects, capcontact, order of pile driving, and the increase oflateral pressure influence the efficiency of a pilegroup.

The pile group efficiency formula developed bySayed and Bakeer (1992) accounts for the three-dimensional geometry of the pile group, soilstrength and time-dependent change, and type ofembedding soil (cohesive and cohensionless). For atypical configuration of pile group, the groupefficiency hg is expressed as:

hg ¼ 1� (1� h0s � K ) � r (7:45)

where h0s ¼ geometric efficiency

K ¼ group interaction factor, and

r ¼ friction factor

This equation is particularly applicable for com-puting the efficiency of pile groups where aconsiderable percentage of the load is carriedthrough shaft resistance. For an end bearing pilegroup, the term r becomes practically equal to zero,and accordingly, the formula yields a value of hg ofone. The formula does not account for the contri-bution of pile cap resistance to the overall bearingcapacity of the pile group (neglected due to thepotential of erosion or loss of support from settle-ment of the soil).

For a pile group arranged in a rectangular orsquare array, the geometric efficiency h0

s is definedas h0

s ¼ Pg=SPp, where Pg ¼ the perimeter of thepile group; and SPp is the summation of the

perimeters of the individual piles in the group.Generally, h0

s increases with an increase in the pilespacing-to-diameter (width) ratio s/d, and itstypical values are between 0.6 and 2.5.

The factor K is a function of the method of pileinstallation, pile spacing, and soil type. It is alsoused to model the change in soil strength due topile driving (e.g., compaction in cohesionless soilsor remolding in cohesive soils). The value of Kmayrange from 0.4 to 9, where higher values areexpected in dense cohesionless or stiff cohesivesoils and smaller values are expected in loose orsoft soils. Avalue of 1 is obtained for piles driven insoft clay and a value greater than 1 is expected forsands. The appropriate value of K is determinedaccording to the relative density of the sand or theconsistency of the clay. For example, a value of 2 to3 is appropriate in medium-dense sand. Thesevalues were back-calculated from the results ofseveral load tests on pile groups.

The friction factor r is defined as Qsu/Qu, whereQsu and Qu are the ultimate shaft resistance andtotal capacity of a single pile, respectively. Thisfactor can be used to introduce the effect of timeinto the analysis when it is important to assess theshort-term as well as the long-term efficiencies of apile group. This is achieved by considering the gainor loss in the shear strength of the soil in thecalculation of Qsu and Qu. Typical values of r mayrange from zero for end-bearing piles to one forfriction piles, with typical values of greater than0.60 for friction or floating type foundations.

Pile dynamic measurements and related analy-sis (i.e., PDA and CAPWAP) made at the End OfInitial Driving (EOID) and the Beginning OfRestrike (BOR) can provide estimates of the frictionfactor r for the short-term and/or long-term con-ditions, respectively. For bored piles in cohesivesoils, some remolding and possibly lateral stressrelief usually occur during construction. It issuggested that the friction factor r be determinedfrom a total stress analysis to calculate the short-term efficiency. The long-term efficiency should bebased on an effective stress approach. Two typesof triaxial tests, unconsolidated-undrained (UU)and consolidated-undrained with pore-pressuremeasurement (CU), can be used to provide therequired strength parameters for the analysis.Moreover, high-strain dynamic tests can be per-formed on bored piles to provide similar infor-mation on the friction factor r, analogous tothat obtained for driven piles. Some geotechnical

7.50 n Section Seven

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engineers may prefer to use a total stress approachto compute the ultimate load capacity for both theshort and long-term capacities. One can stillcompute the friction factor r for the short-termand/or long-term conditions provided that ade-quate information is available regarding thethixotropic gain of strength with time.

The three-dimensional modeling of pile groupsincorporating the efficiency formula can be per-formed using the Florida-Pier (FLPIER) computerprogram developed by the Florida Department ofTransportation (FDOT) at the University of Florida.FLPIER considers both axial and lateral pile-soilinteraction and group effects. Pile-soil-pile inter-action effects are considered through p-y multi-pliers which are assigned for each row within thegroup for lateral loading and group efficiency hg

for axial loading. FHWA’s computer programCOM624 is also available for modeling the lateralpile-soil interaction.

(S. M. Sayed and R. M. Bakeer (1992),“Efficiency Formula For Pile Groups,” Journal ofGeotechnical Engineering, ASCE, 118 No. 2, pages278–299; S. T. Wang and L. C. Reese, L. C. (1993)“COM624P–Laterally Loaded Pile Analysis Pro-gram for the Microcomputer, Version 2.0,” FHWAOffice of Technology Applications, Publication No.FHWA-SA-91-048, Washington, D.C. 20590; Flor-ida Department of Transportation FDOT (1995)“User’s Manual for Florida Pier Program”, www.dot.state.fl.us)

A very approximate analysis of group settle-ment, applicable to friction piles, models the pilegroup as a raft of equivalent plan dimensionssituated at a depth below the surface equal to two-thirds the pile length. Subsequently, conventionalsettlement analyses are employed (see Arts. 7.12and 7.13).

Negative Skin Friction, Dragload, andDowndrag n Influenced by consolidation inducedby placement of fill and/or lowering of the watertable, soils along the upper portion of a pile willtend to compress and move down relative to thepile. In the process, load is transferred to the pilethrough negative skin friction. The permanent load(dead load) on the pile and the dragload imposedby the negative skin friction are transferred to thelower portion of the pile and resisted by means ofpositive shaft resistance and by toe resistance.A point of equilibrium, called the neutral plane,

exists where the negative skin friction changes overinto positive shaft resistance. This is where there isno relative movement between the pile and the soil,which means that if the neutral plane is locatedin non-settling soil, then, the pile does not settle.If on the other hand, the soil experiences settlementat the level of the neutral plane, the pile will settlethe same amount, i.e., be subjected to downdrag.Dragload is of concern if the sum of the dragloadand the dead load exceeds the structural strengthof the pile. The following must be considered:

1. Sum of dead plus live loads is smaller than thepile capacity divided by an appropriate factor ofsafety. The dragload is not included with theseloads.

2. Sum of dead load and dragload is smaller thanthe structural strength with an appropriatefactor of safety. The live load is not includedbecause live load and drag load can not coexist.

3. The settlement of the pile (pile group) is smallerthan a limiting value. The live load anddragload are not included in this analysis.

A procedure for construing the neutral plane anddetermining pile allowable load is illustrated inFig. 7.22. The diagrams assume that above theneutral plane, the unit negative skin friction, qn,and positive shaft resistance, rs, are equal, which isan assumption on the safe side. A key factor is theestimate of the pile toe resistance, Rt: If the pile toeresistance is small, the neutral plane lies higherthan when the toe resistance is large. Further, if thepile toe is located in a non-settling soil, the pilesettlement will be negligible and only a function ofthe pile toe penetration necessary to mobilize thepile and bearing resistance. The maximum nega-tive skin friction that can be developed on a singlepile can be calculated with Eq. (7.38) with b factorsfor clay of 0.20 to 0.25, for silt of 0.25 to 0.35, and forsand of 0.35 to 0.50.

A very approximate method of pile-groupanalysis calculates the upper limit of group dragload Qgd from

Qgd ¼ AFgFHF þ PHcu (7:46)

HF, gF, and AF represent the thickness, unit weight,and area of fill containedwithin the group. P,H, andcu are the circumference of the group, the thicknessof the consolidating soil layers penetrated by thepiles, and their undrained shear strength, respect-

Geotechnical Engineering n 7.51

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ively. Such forces as Qgd could only be approachedfor the case of piles driven to rock through heavilysurcharged, highly compressible subsoils.

(H. G. Poulos and E. H. Davis, “Elastic Solutionsfor Soil and Rock Mechanics,” and K. Terzaghi andR. B. Peck, “Soil Mechanics and EngineeringPractice,” John Wiley & Sons, Inc., New York; J. E.Garlanger and W. T. Lambe, Symposium on Down-drag of Piles, Research Report 73–56, Soils Publi-cation no. 331, Massachusetts Institute of Tech-nology, Cambridge, 1973; B. H. Fellenius, “Basics ofFoundation Design,” BiTech Publishers, Rich-mond, BC, Canada, 1999).

7.19.7 Design of Piles for LateralLoads

Piles and pile groups are typically designed tosustain lateral loads by the resistance of vertical

piles, by inclined, or batter, piles, or by a com-bination. Tieback systems employing ground anchoror deadmen reactions are used in conjunction withlaterally loaded sheetpiles (rarely with foundationpiles).

Lateral loads or eccentric loading produceoverturning moments and uplift forces on a groupof piles. Under these circumstances, a pile mayhave to be designed for a combination of bothlateral and tensile load.

Inclined Piles n Depending on the degree ofinclination, piles driven at an angle with thevertical can have a much higher lateral-loadcapacity than vertical piles since a large part ofthe lateral load can be carried in axial compression.To minimize construction problems, however, pilebatters (rake) should be less than 1 horizontal to 2vertical.

Fig. 7.22 Construing the Neutral Plane and determining allowable load (Guidelines for Static PileDesign, – A Continuing Education Short Course Text, B. H. Fellenius, Deep Foundations Institute, 1991(www.dfi.org)).

7.52 n Section Seven

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Evaluation of the load distribution in a pilegroup consisting of inclined piles or combinedvertical and batter piles is extremely complexbecause of the three-dimensional nature andindeterminancy of the system. A variety ofcomputer solutions have become available andallow a rational evaluation of the load distributionto inclined group piles. The same methods of axial-capacity evaluation developed for vertical piles areapplied to inclined-pile design, although higherdriving energy losses during construction suggestthat inclined piles would have a somewhatreduced axial-load capacity for the same terminalresistance. (A. Hrennikoff, “Analysis of PileFoundations with Batter Piles,” ASCE Transactions,vol. 115, 1950.)

Laterally Loaded Vertical Piles n Vertical-pile resistance to lateral loads is a function of boththe flexural stiffness of the pile, the stiffness of thebearing soil in the upper 4D to 6D length of pile,where D ¼ pile diameter, and the degree of pile-head fixity. The lateral-load design capacity is alsorelated to the amount of lateral deflection permit-ted and, except under very exceptional circum-stances, the tolerable-lateral-deflection criteria willcontrol the lateral-load design capacity.

Design loads for laterally loaded piles areusually evaluated by beam theory for both anelastic and nonlinear soil reaction, although elasticand elastoplastic continuum solutions are avail-able. Nonlinear solutions require characterization ofthe soil reaction p versus lateral deflection y alongthe shaft. In obtaining these solutions, degradationof the soil stiffness by cyclic loading is an importantconsideration.

The lateral-load vs. pile-head deflection rela-tionship is readily developed from charted non-

dimensional solutions of Reese and Matlock. Thesolution assumes the soil modulus K to increaselinearly with depth z; that is, K ¼ nhz, where nh ¼coefficient of horizontal subgrade reaction. Acharacteristic pile length T is calculated from

T ¼ffiffiffiffiffiEI

nh

s(7:47)

where EI ¼ pile stiffness. The lateral deflection y ofa pile with head free to move and subject to a lateralload Pt and moment Mt applied at the ground lineis given by

y ¼ AyPtT3

EIþ ByMt

T2

EI(7:48)

where Ay and By are nondimensional coefficients.Nondimensional coefficients are also available forevaluation of pile slope, moment, shear, and soilreaction along the shaft.

For positive moment,

M ¼ AmPtT þ BmMt (7:49)

Positive Mt and Pt values are represented byclockwise moment and loads directed to the righton the pile head at the ground line. The coefficientsapplicable to evaluation of pile-head deflection andto the maximum positive moment and its approxi-mate position on the shaft z/T, where z ¼ distancebelow the ground line, are listed in Table 7.9.

The negative moment imposed at the pile headby pile-cap or other structural restraint can beevaluated as a function of the head slope (rotation)from

�Mt ¼ AuPtT

Bu� usEI

BuT(7:50)

Table 7.9 Deflection, Moment, and Slope Coefficients

zmax Ay By Au Bu Am* Bm* z/T*

2 4.70 3.39 23.40 23.21 0.51 0.84 0.853 2.65 1.77 21.75 21.85 0.71 0.60 1.494 2.44 1.63 21.65 21.78 0.78 0.70 1.32

.5 2.43 1.62 21.62 21.75 0.77 0.69 1.32

*Coefficients for maximum positive moment are located at about the values given in the table for z/T.

Source: L. C. Reese and H. Matlock, “Non-Dimensional Solutions for Laterally Loaded Piles with Soil Modulus AssumedProportional to Depth,” 8th Texas Conference of Soil Mechanics and Foundation Engineering, University of Texas, 1956.

Geotechnical Engineering n 7.53

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where us, rad, represents the counterclockwise (þ)rotation of the pile head and Au and Bu arecoefficients (see Table 7.9). The influence of thedegrees of fixity of the pile head on y andM can beevaluated by substituting the value of 2Mt fromEq. (7.50) in Eqs. (7.48) and (7.49). Note that for thefixed-head case:

yf ¼ PtT3

EIAy �

AuBy

Bu

� �(7:51)

Improvement of Lateral Resistance n

The lateral-load capacity of a specific pile type canbe most effectively increased by increasing thediameter, i.e., the stiffness and lateral-bearing area.Other steps are to improve the quality of thesurficial bearing soils by excavation and replace-ment or in-place densification, to add reinforce-ment, and to increase the pile-head fixity condition.

Typical lateral-load design criteria for buildingslimit lateral pile-head deformations to about 1⁄4 in.Associated design loads for foundation pilesdriven in medium dense sands or medium claysare typically in the range of 2 to 4 tons, althoughsignificantly higher values have been justified byload testing or detailed analyses or a combination.

Resistance of pile groups to lateral loads is notwell-documented by field observations. Results ofmodel testing and elastic analysis, however,indicate that pile spacings less than about 8 pilediameters D in the direction of loading reduce thesoil modulus K. The reduction factors are assumedto vary linearly from 1.0 at 8D to 0.25 at a 3Dspacing if the number of piles in the group is 5 ormore and the passive resistance of the pile cap isignored. The effect of this reduction is to “soften”the soil reaction and produce smaller lateralresistance for a given group deflection. Elasticanalyses also confirm the long-held judgment thatbatter piles in the center of a pile group are largelyineffective in resisting lateral load.

(B. B. Broms, “Design of Laterally Loaded Piles,”ASCE Journal of Soil Mechanics and FoundationEngineeringDivision, vol. 91, no. SM3, 1965.H. Y. Fang,“Foundation Engineering Handbook,” Van NostrandReinhold, New York. B. H. Fellenius, “Guidelines forStatic Pile Design,” Deep Foundation Institute, 120Charolette Place, Englewood Cliffs, NJ 07632(www.dfi.org). H. G. Poulos and E. H. Davis, “ElasticSolutions for Soil and RockMechanics,” JohnWiley &Sons, Inc., New York. L. C. Reese and R. C. Welch,“Lateral Loading of Deep Foundations in Stiff Clay,”

ASCE Journal of Geotechnical Engineering, vol. 101, no.GT7, 1975.)

7.19.8 Static-Load Pile Testing

Because of the inherent uncertainty in static pile-design methods and the influence of constructionprocedures on the behavior of piles, static-loadtests are desirable or may be required. Static-loadtests are almost always conducted on single piles;testing of pile groups is very rare.

Engineers use static-load tests to determine theresponse of a pile under applied loads. Axialcompression testing is the most common, althoughwhen other design considerations control, uplift orlateral loading tests are also performed. In somespecial cases, testing is performed with cyclicloadings or with combined loads; for example,axial and lateral loading. Pile testing may beperformed during the design or construction phaseof a project so that foundation design data andinstallation criteria can be developed or verified, orto prove the adequacy of a pile to carry design load.

Use of static-load pile testing is limited byexpense and time required for the tests and analyses.For small projects, when testing costs add signifi-cantly to the foundation cost, the increased costoften results in elimination of pile testing. Forprojects involving a large number of piles, static-load pile tests usually are performed, but only a fewpiles are tested. (A typical recommendation is that ofthe total number of piles to be installed in normalpractice 1% be tested, but the percentage of pilestested in actual practice may be much lower.) Thenumber and location of test piles should bedetermined by the foundation design engineer afterevaluation of the variability of subsurface con-ditions, pile loadings, type of pile, and installationtechniques. Waiting time between pile installationand testing generally ranges from several days toseveral weeks, depending on pile type and soilconditions.

The foundation contractor generally is respon-sible for providing the physical setup for conduct-ing a static-load test. The foundation designershould supervise the testing.

Standards detailing procedures on how toarrange and conduct static-load pile tests include“Standard Test Method for Piles under Static AxialCompression Load,” ASTM D1143 (www.astm.org); “Standard Method of Testing Individual Pilesunder Static Axial Tension Load,” ASTM D3689;

7.54 n Section Seven

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and “Standard Method of Testing Piles underLateral Loads,” ASTM D3966. See also “StaticTesting of Deep Foundations,” U. S. FederalHighway Administration, Report No. FHWA-SA-91-042 (www.fhwa.gov), 1992; “Axial Pile LoadingTest—Part 1: Static Loading,” International Societyfor Soil Mechanics and Foundation Engineering,1985; and “Canadian Foundation EngineeringManual,” 2nd ed., Canadian Geotechnical Society,1985.

Load Application n In a static-load pile test, ahydraulic jack, acting against a reaction, appliesload at the pile head. The reaction may be providedby a kentledge, or platform loaded with weights(Fig. 7.23), or by a steel frame supported by reactionpiles (Fig. 7.24), or by ground anchors. The distanceto be used between the test pile and reaction-system supports depends on the soil conditionsand the level of loading but is generally three pilediameters or 8 ft, whichever is greater. It may benecessary to have the test configuration evaluatedby a structural engineer.

Hydraulic jacks including their operationshould conform to “Safety Code for Jacks,” ANSIB30. 1, American National Standards Institute. Thejacking system should be calibrated (with loadcells, gages, or machines having an accuracy of atleast 2%) within a 6-month period prior to piletesting. The available jack extension should be at

least 6 in. The jack should apply the load at thecenter of the pile (Fig. 7.25). When more than onejack is needed for the test, all jacks should bepressurized by a common device.

Loads should be measured by a calibratedpressure gage and also by a load cell placedbetween the jack and the pile. Internal forces in thepile may be measured with strain gages installedalong the pile. Two types of test-loading pro-cedures are used: the maintained load (ML) and theconstant rate of penetration (CRP) methods.

In the ML method, load is applied in incre-ments of 25% of the anticipated pile capacity untilfailure occurs or the load totals 200% of the designload. Each increment ismaintained until pile move-ment is less than 0.01 in/h or for 2 h, whicheveroccurs first. The final load is maintained for 24 h.Then, the test load is removed in decrements of25% of the total test load, with 1 h between dec-rements. This procedure may require from 1 to 3days to complete. According to some practices, theML method is changed to the CRP procedure assoon as the rate exceeds 0.8 in/h.

Tests that consist of numerous load incre-ments (25 to 40 increments) applied at constanttime intervals (5 to 15 min) are termed quick tests.

In the CRP procedure, the pile is continuouslyloaded so as to maintain a constant rate ofpenetration into the ground (typically between0.01 and 0.10 in/min for granular soils and 0.01 to0.05 in/min for cohesive soils). Loading is con-tinued until no further increase is necessary forcontinuous pile penetration at the specified rate.As long as pile penetration continues, the load

Fig. 7.24 Reaction piles used in static-load teston a pile.

Fig. 7.23 Static-load test on a pile with deadweight as the reaction load.

Geotechnical Engineering n 7.55

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inducing the specified penetration rate is main-tained until the total pile penetration is at least 15%of the average pile diameter or diagonal dimen-sion, at which time the load is released. Also, if,under the maximum applied load, penetrationceases, the load is released.

Alternatively, for axial-compression static-loadtests, sacrificial jacks or other equipment, such asthe Osterberg Cell, may be placed at the bottomof the pile to load it (J. O. Osterberg, “New LoadCell Testing Device,” Deep Foundations Institute(www.dfi.org)). One advantage is automatic sep-aration of data on shaft and toe resistance. Anotheris elimination of the expense and time required forconstructing a reaction system, inasmuch as soilresistance serves as a reaction. A disadvantage isthat random pile testing is not possible since theloading apparatus and pile installations must beconcurrent.

Penetration Measurements n The axialmovement of the pile head under applied load maybe measured by mechanical dial gages or electro-mechanical devices mounted on an independentlysupported (and protected) reference beam. Figure7.25 shows a typical arrangement of equipment

and instruments at the pile head. The gages shouldhave at least 2 in of travel (extendable to 6 in) andtypically a precision of at least 0.001 in. For redun-dancy, measurements may also be taken with asurveyor’s rod and precise level and referenced tofixed benchmarks. Another alternative is a tightlystretched piano wire positioned against a mirrorand scale that are attached to the side of the pile.Movements at locations along the pile length andat the pile toe may be determined with the use oftelltales.

For the ML or quick-test procedures, pile move-ments are recorded before and after the appli-cation of each load increment. For the CRPmethod,readings of pile movement should be taken at leastevery 30 s.

Pile-head transverse displacements should bemonitored and controlled during the test. Forsafety and proper evaluation of test results, move-ments of the reaction supports should also bemonitored during the test.

Interpretation of Test Results n A con-siderable amount of data is generated during astatic-load test, particularly with instrumentedpiles. The most widely used procedure for pre-senting test results is the plot of pile-head load vs.

Fig. 7.25 Typical rearrangement of loading equipment and instrumentation at pile head for acompression static-load test. (From “Static Testing of Deep Foundations,” FHWA SA-91-042, Federal HighwayAdministration.)

7.56 n Section Seven

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movement. Other results that may be plottedinclude pile-head time vs. movement and loadtransfer (from instrumentation along the pileshaft). Shapes of load vs. movement plots varyconsiderably; so do the procedures for evaluatingthem for calculation of limit load (often mistakenlyreferred to as failure load).

Problems in data interpretation arise from thelack of a universally recognized definition of failure.For a pile that has a load-carrying capacity greaterthan that of the soil, failure may be considered tooccur when pile movement continues undersustained or slightly increasing load (pile plun-ging). In general, the term failure load should bereplaced with interpreted failure load for evaluationsfrom plots of pile load vs. movement. Thedefinition of interpreted failure load should bebased on mathematical rules that produce repea-table results without being influenced by thesubjective interpretation of the engineer. In theoffset limit method, interpreted failure load isdefined as the value of the load ordinate of the loadvs. movement curve at r þ 0.15 þ D/120, where ris the movement, in, at the termination of elasticcompression andD is the nominal pile diameter, in.One advantage of this technique is the ability totake pile stiffness into consideration. Anotheradvantage is that maximum allowable pile move-ment for a specific allowable load can be calculatedprior to proof testing of a pile. Interpretationmethods that rely on extrapolation of the load-movement curve should be avoided. (“Guidelinesfor the Interpretation and Analysis of the StaticLoading Test,” 1990, B. H. Fellenius, Deep Foun-dation Institute, www.dfi.com).

The test report should include the following aswell as other relevant data:

1. Information on general site subsurface con-ditions, emphasizing soil data obtained fromexploration near the test pile

2. Descriptions of the pile and pile installationprocedure

3. Dates and times of pile installation and statictesting

4. Descriptions of testing apparatus and testingprocedure

5. Calibration certificates

6. Photographs of test setup

7. Plots of test results

8. Description of interpretation methods

9. Name of testing supervisor

The cost, time, and effort required for a static-load test should be carefully weighed against themany potential benefits. A static-load test on asingle pile, however, does not account for theeffects of long-term settlement, downdrag loads,time-dependent soil behavior, or pile group action,nor does the test eliminate the need for an adequatefoundation design.

7.20 Dynamic Pile Testingand Analysis

Simple observations made during impact piledriving are an important and integral part of thepile installation process. In its most basic form,dynamic-load pile testing encompasses visualobservations of hammer operation and pilepenetration during pile driving. Some engineersapply equations based on the Newtonian physicsof rigid bodies to the pile movements recordedduring pile driving to estimate the load-carryingcapacity of the pile. The basic premise is that theharder it is to drive the pile into the ground, themore load it will be able to carry. The equations,generally known as energy formulas, typicallyrelate hammer energy and work done on the pile tosoil resistance. More than 400 formulas have beenproposed, including the widely used and simpleEngineering News formula.

This method of estimating load capacity,however, has several shortcomings. These includeincomplete, crude, and oversimplified represen-tation of pile driving, pile and soil properties, andpile-soil interaction. Often, the method has beenfound to be grossly inaccurate and unreliable tothe extent that many engineers believe that itshould be eliminated from contemporary practice.

Modern rational dynamic testing and analysisincorporates pile dynamic measurements analyzedwith one-dimensional elastic stress wave propa-gation principles and theories. Such testingmethods have become routine procedures in con-temporary foundation engineering practice world-wide. They are covered in many codes andspecifications. (ASTM D 4945-96: Standard TestMethod for High-Strain Dynamic Testing of Piles;“Application of Stress Wave Theory to Piles:Quality Assurance on Land and Offshore Piling,”

Geotechnical Engineering n 7.57

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Proceedings of 6th International Conference, SanPaulo, Brazil, 2000, A. A. Balkema Publishers,www.balkema.nl).

7.20.1 Wave Equation

In contrast to the deficiencies of the energyformulas, analysis of pile-driving blow count orpenetration per blow yields more accurate esti-mates of the load-carrying capacity of a pile, ifbased on accurate modeling and rational prin-ciples. One such type of analysis employs thewave equation based on a concept developed byE.A. Smith (ASCE Journal of Geotechnical EngineeringDivision, August 1960). Analysis is facilitated by useof computer programs such as GRLWEAP (GobleRausche Likins and Associates, Inc., Cleveland,Ohio) that simulate and analyze impact piledriving. Sophisticated numerical modeling, advan-ced analytical techniques, and one-dimensionalelastic-wave-propagation principles are required.Computations can be performed with personalcomputers. A substantial improvement that thewave equation offers over the energy approach isthe ability to model all hammer, cushion, pile-cap,and pile and soil components realistically.

Figure 7.26 illustrates the lumped-mass modelused in wave equation analyses. All componentsthat generate, transmit, or dissipate energy arerepresented by a spring, mass, or dashpot. Thesepermit representation of mass, stiffness, andviscosity.

A series of masses and springs represent themass and stiffness of the pile. Elastic springs andlinear-viscous dashpots model soil-resistanceforces along the pile shaft and under the toe. Thesprings represent the displacement-dependentstatic-loaded components, and the dashpots theloading-dependent dynamic components. Springsmodel stiffness and coefficient of restitution (toaccount for energy dissipation) of hammer and pilecushions. A single mass represents the pile-cap. Forexternal-combustion hammers, the representationis straightforward: a stocky ram, by a single mass;the hammer assembly (cylinder, columns, etc.), bymasses and springs. For internal-combustionhammers, the modeling is more involved. Theslender ram is divided into several segments. Thegas pressure of the diesel combustion cycle iscalculated according to the thermodynamic gas lawfor either liquid or atomized fuel injection.

The parameters needed for execution of a waveequation analysis with the GRLWEAP computerprogram are:

Hammer: Model and efficiency

Hammer and Pile Cushions: Area, thickness, elasticmodulus, and coefficient of restitution

Pile Cap: Weight, including all cushions and anyinserts

Pile: Area, elastic modulus, and density, all as afunction of length

Soil: Total static capacity, percent shaft resistanceand its distribution, quake and damping constantsalong the shaft and under the toe

In practice, wave equation analysis is employedto deal with the following questions:

1. If the input to the computer program provides acomplete description of hammer, cushions, pilecap, pile, and soil, can the pile be driven safelyand economically to the required staticcapacity?

2. If the input provides measurements of pilepenetration during pile driving or restrikingblow count, what is the static-load capacity ofthe pile?

For case 1, pile design and proper selection ofhammer and driving system can be verified toensure that expected pile-driving stresses are belowallowable limits and reasonable blow count isattainable before actual field work starts. For case 2,given field observations made during pile driving,the analysis is used as a quality-control tool toevaluate pile capacity.

Generally, wave equation analysis is applied toa pile for the cases of several static-load resistancescovering a wide range of values (at a constant pilepenetration corresponding to the expected finalpile-toe depth). Analysis results are then plotted asa bearing graph relating static pile capacity anddriving stresses to blow counts.

Figure 7.27 presents a bearing graph from ananalysis of a single-acting external-combustionhammer (Vulcan 012) and a precast concrete pile(18 in square, 95 ft long).

Foradieselhammer, thestrokeorbounce-chamberpressure is also included in the plot. Alternatively, foran open-end diesel hammer (or any hammer with

7.58 n Section Seven

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variable stroke), the analysis may be performed witha constant pile static capacity and various strokes.In this way, the required blow count can be obtainedas a function of the actual stroke.

Wave equation analysis may also be based onpile penetration (commonly termed pile drivabil-ity). In this way, variations of soil resistance with

depth can be taken into account. Analysis resultsare obtained as a function of pile penetration.

Pile specifications prescribe use of waveequation analysis to determine suitability of a pile-driving system. Although it is an excellent tool foranalysis of impact pile driving, the wave equationapproach has some limitations. These are mainly

Fig. 7.26 Lumped-mass model of a pile used in wave equation analysis. (a) Rectangular block withspring represents mass and stiffness; dashpot, the loading-dependent dynamic components; small squareblock with spring, the soil resistance forces along the pile shaft. (b) Variation of dynamic soil resistancewith pile velocity. (c) Variation of static soil resistance with pile displacement.

Geotechnical Engineering n 7.59

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due to uncertainties in quantifying some of therequired inputs, such as hammer performance andsoil parameters. The hammer efficiency valueneeded in the analysis is usually taken as theaverage value observed in many similar situations.Also, soil damping and quake values (maximumelastic soil deformation) needed in modeling soilbehavior cannot be readily obtained from standardfield or laboratory soil tests or related to otherconventional engineering soil properties.

Dynamic-load pile testing and data analysisyield information regarding the hammer, drivingsystem, and pile and soil behavior that can be used

to confirm the assumptions of wave equationanalysis. Dynamic-load pile tests are routinelyperformed on projects around the world for thepurposes of monitoring and improving pileinstallation and as construction control procedures.Many professional organizations have establishedstandards and guidelines for the performance anduse of this type of testing; for example, ASTM(D4945), Federal Highway Administration (“Man-ual on Design and Construction of Driven PileFoundation”). Dynamic-load testing methods arealso effectively employed for evaluating cast-in-place piles (“Dynamic Load Testing of Drilled

Fig. 7.27 Bearing graph derived from a wave equation analysis. (a) Variation of pile tensile andcompressive stresses with blows per foot. (b) Ultimate capacity of pile indicated by blows per foot. (c) Skinfriction distribution along the tested pile. Driving was done with a Vulcan hammer, model 012, with 67%efficiency. Helmet weighed 2.22kips (k). Stiffness of hammer cushion was 5765k/in and of pile cushion,1620k/in. Pile was 95ft long and had a top area of 324in2. Other input parameters were quake (soilmaximum elastic deformation), 0.100in for shaft resistance and 0.150in for toe resistance; soil dampingfactor, 0.150s/ft for shaft and toe resistance.

7.60 n Section Seven

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Shaft—Final Report,” Department of Civil Engin-eering, University of Florida, Gainesville 1991).

Main objectives of dynamic-load testing includeevaluation of driving resistance and static-loadcapacity; determination of pile axial stresses duringdriving; assessment of pile structural integrity; andinvestigation of hammer and driving system per-formance.

(ASTM D 4945-96: Standard Test Method forHigh-Strain Dynamic Testing of Piles; “DynamicTesting of Pile Foundations During Construction,”M. Hussein and G. Likins, Proceedings of ASCEStructures Congress XIII, Boston, MA, 1995).

7.20.2 Case Method

A procedure developed at Case Institute ofTechnology (now Case Western Reserve Univer-sity), Cleveland, Ohio, by a research team headedby G. G. Goble, enables calculation of pile staticcapacity from measurements of pile force andacceleration under hammer impacts during piledriving. The necessary equipment and analyticalmethods developed have been expanded toevaluate other aspects of the pile-driving process.These procedures are routinely applied in the fieldusing a device called the Pile Driving Analyzer(PDA). As an extension of the original work theresearchers developed a computer program knownas the CAse Pile Wave Analysis Program (CAP-WAP), which is described later.

Measurements of pile force and velocity recordsunder hammer impacts are the basis for moderndynamic pile testing. Data are obtained with theuse of reusable strain transducers and acceler-ometers. Strain gages are bolted on the pile shaft,usually at a distance of about two pile diametersbelow the pile head. The PDA serves as a dataacquisition system and field computer that pro-vides signal conditioning, processing, and cali-bration of measurement signals. It convertsmeasurements of pile strains and acceleration topile force and velocity records. Dynamic recordsand testing results are available in real timefollowing each hammer impact and are perma-nently stored in digital form. Using wave propa-gation theory and some assumptions regardingpile and soil, the PDA applies Case methodequations and computes in a closed-form solutionsome 40 variables that fully describe the conditionof the hammer-pile-soil system in real timefollowing each hammer impact.

When a hammer or drop weight strikes the pilehead, a compressive-stress wave travels down thepile shaft at a speed c, which is a function of the pileelastic modulus and mass density (Art. 6.82.1). Theimpact induces at the pile head a force F and aparticle velocity v. As long as the wave travels inone direction, force and velocity are proportional;that is, F ¼ Zv, where Z is the pile impedance, andZ ¼ EA/c, where A is the cross-sectional area of thepile and E is its elastic modulus. Changes inimpedance in the pile shaft and pile toe, and soil-resistance forces, produce wave reflections. Thereflected waves arrive at the pile head after impactat a time proportional to the distance of theirlocation from the toe. Soil-resistance forces orincrease in pile impedance cause compressive-wave reflections that increase pile force anddecrease velocity. Decrease in pile impedance hasthe opposite effect.

For a pile of length L, impedance Z, and stress-wave velocity c, the PDA computes total soilresistance from measured force and velocityrecords during the first stress-wave cycle; that is,when 0 , t � 2L/c, where t is time measured fromstart of hammer impact. This soil resistanceincludes both static and viscous components. Inthe computation of pile bearing capacity understatic load RS at the time of testing, effects of soildamping must be considered. Damping is associ-ated with velocity. By definition, the Case methoddamping force is equal to ZJcvb, where Jc is thedimensionless Case damping factor, and vb the piletoe velocity, which can be computed frommeasured data at the pile head by applying wavemechanics principles. The static capacity of a pilecan be calculated from:

RS ¼ 1

2[(1� Jc)(Ft1 þ Zvt1)þ (1þ Jc)(Ft2 � Zvt2)]

(7:51a)

where t2 ¼ t1 þ 2L/c and t1 is normally the time ofthe first relative velocity peak. The dampingconstant Jc is related to soil grain size and may betaken for clean sands as 0.10 to 0.15, for silty sandsas 0.15 to 0.25, for silts as 0.25 to 0.40, for silty claysas 0.4 to 0.7, and for clays as 0.7 to 1.0.

The computedRS value is the pile static capacityat the time of testing. Time-dependent effects can beevaluated by testing during pile restrikes. Forthis purpose, the pile must have sufficient pene-tration under the hammer impact to achieve full

Geotechnical Engineering n 7.61

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mobilization of soil-resistance forces. (F. Rausche,G. Goble, and G. Likins, “Dynamic Determinationof Pile Capacity,” ASCE Journal of GeotechnicalEngineering Division, vol. 111, no. 3, 1985.)

The impact of a hammer subjects piles to a com-plex combination of compression, tension, torsional,and bending forces. Maximum pile compressivestress at the transducers’ location is directly obtainedfrom the measured data as the maximum recordedforce divided by the pile area. For piles with mainlytoe soil resistance, the compressive force at the piletoe is calculated from pile-head measurements andone-dimensional wave propagation considerations.Maximum tension force in the pile shaft can becomputed frommeasurements near the pile head byconsidering the magnitude of both upward- anddownward-traveling force components. Pile damageoccurs if driving stresses exceed the strength of thepile material.

For a pile with a uniform cross-sectional areainitially, damage after driving can be indicated by achange in area. Since pile impedance is pro-portional to pile area, a change in impedancewould indicate pile damage. Hence, a driven pilecan be tested for underground damage bymeasuring changes in pile impedance. Changes inpile impedance cause wave reflections and changesin the upward-traveling wave measured at the pilehead. From the magnitude and time after impact ofthe relative wave changes, the extent and locationof impedance change and hence of pile damage canbe determined. Determination of pile damage canbe assisted with the use of the PDA, whichcomputes a relative integrity factor (unity foruniform piles and zero for a pile end) based onmeasured data near the pile head. (F. Rausche andG. G. Goble, “Determination of Pile Damage byTop Measurements,” ASTM STP-670; “StructuredFailure of Pile Foundations During Installation,”M. J. Hussein and G. G. Goble, ASCE, Proceedingsof the Construction Congress VI, Orlando, FL,2000.)

The PDA also is helpful in determining theenergy actually received by a pile from a hammerblow. Whereas hammers are assigned an energyrating by manufacturers, only the energy reachingthe pile is of significance in effecting pilepenetration. Due to many factors related to thehammer mechanical condition, driving-systembehavior, and general dynamic hammer-cushions-pile-soil incompatibility, the percentage of potentialhammer energy that actually reaches the pile is

quite variable and often less than 50%. (“ThePerformance of Pile Driving Systems—MainReport,” vol. 1–4, FHWADTFH 61-82-1-00059,Federal Highway Administration.) Figure 7.28presents a summary of data obtained on hundredsof sites to indicate the percentage of all hammers ofa specific type with an energy-transfer efficiencyless than a specific percentage. Given records ofpile force and velocity, the PDA calculates thetransferred energy as the time integral of theproduct of force and velocity. The maximumtransferred energy value for each blow representsthe single most important parameter for an overallevaluation of driving-system performance.

7.20.3 CAPWAP Method

The CAse Pile Wave Analysis Program (CAPWAP)combines field-measured dynamic-load data andwave-equation-type analytical procedures to pre-dict pile static-load capacity, soil-resistance distri-bution, soil-damping and quake values, pile loadvs. movement plots, and pile-soil load-transfercharacteristics. CAPWAP is a signal-matching orsystem identification method; that is, its results arebased on a best possible match between a compu-ted variable and its measured equivalent.

The pile is modeled with segments about 3 ftlong with linearly elastic properties. Piles withnonuniform cross sections or composite construc-tion can be accurately modeled. Static and dynamicforces along the pile shaft and under its toerepresent soil resistance. Generally, the soil modelfollows that of the Smith approach (Art. 7.20.1)with modifications to account for full pile-penetration and rebound effects, including radi-ation damping. At the start of the analysis, anaccurate pile model (incorporating splices, ifpresent) is established and a complete set of soilconstants is assumed. The hammer model used forthe wave equation method is replaced by themeasured velocity imposed as a boundary con-dition. The program calculates the force necessaryto induce the imposed velocity. Measured andcalculated forces are compared. If they do notagree, the soil model is adjusted and the analysisrepeated. This iterative process is continued untilno further improvement in the match can beobtained. The total number of unknowns to beevaluated during the analysis is Ns þ 18, where Ns

is the number of soil elements. Typically, one soil

7.62 n Section Seven

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element is placed at every 6 ft of pile penetrationplus an additional one under the toe.

Results that can be obtained from a CAPWAPanalysis include the following:

Comparisons of measured values with correspond-ing computed values

Soil-resistance forces and their distribution forstatic loads

Soil-stiffness and soil-damping parameters alongthe pile shaft and under its toe

Forces, velocities, displacements, and energies as afunction of time for all pile segments

Simulation of the relationship between static loadsand movements of pile head and pile toe

Pile forces at ultimate soil resistance

Correlations between CAPWAP predicted valuesand results from static-load tests indicate verygood agreement. (ASCE Geotechnical Special Pub-lication No. 40, 1994.)

7.20.4 Low-Strain Dynamic IntegrityTesting

The structural integrity of driven or cast-in-placeconcrete piles may be compromised duringinstallation. Piles may also be damaged afterinstallation by large lateral movements fromimpacts of heavy equipment or from slope orretaining-wall failures. Procedures such as exca-vation around a suspect pile or drilling and coringthrough its shaft are crude methods for investi-gating possible pile damage. Several testingtechniques are available, however, for evaluation

Fig. 7.28 Comparison of performance of two types of hammers when driving steel or concrete piles.The percentile indicates the percentage of all hammers in each case with a rated transfer efficiency lessthan a specific percentage.

Geotechnical Engineering n 7.63

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of the structural integrity of deep foundationelements in a more sophisticated manner(W. G. Fleming, A. J. Weltmen, M. F. Randolph,and W. K. Elson, “Piling Engineering,” SurreyUniversity Press, London). Some of these tests,though, require that the pile be prepared orinstrumented before or during installation. Theserequirements make random application prohibi-tively expensive, if not impossible. A convenientand economical method is the low-strain pulse-echo technique, which requires relatively littleinstrumentation and testing effort and which isemployed in low-strain, dynamic-load integritytesting. This method is based on one-dimensionalwave mechanics principles and the measurementof dynamic-loading effects at the pile head underimpacts of a small hand-held hammer. (ASTM D5882-00: Standard Test Method for Low-StrainIntegrity Testing of Piles.)

The following principle is utilized: Whenimpacted at the top, a compressive-stress wavetravels down the pile shaft at a constant speed cand is reflected back to the pile head from the toe.Changes in pile impedance Z change wavecharacteristics and indicate changes in pile cross-sectional size and quality and thus possible piledamage (Art. 7.20.2). Low-strain integrity testing isbased on the premise that changes in pileimpedance and soil-resistance forces producepredictable wave reflections at the pile head. Thetime after impact that the reflected wave isrecorded at the pile head can be used to calculatethe location on the pile of changes in area or soilresistance.

Field equipment consists of an accelerometer,a hand-held hammer (instrumented or withoutinstrumentation), dedicated software, and a PileIntegrity Tester (Fig. 7.29), a data acquisitionsystem capable of converting analog signals todigital form, data processing, and data storage. Pilepreparation involves smoothing and leveling of asmall area of the pile top. The accelerometer isaffixed to the pile top with a jell-type material, andhammer blows are applied to the pile head.Typically, pile-head data resulting from severalhammer blows are averaged and analyzed.

Data interpretation may be based on records ofpile-top velocity (integral of measured accelera-tion), data in time or frequency domains, or morerigorous dynamic analysis. For a specific stress-wave speed (typically 13,000 ft/s), records ofvelocity at the pile head can be interpreted for pile

nonuniformities and length. As an example,Fig. 7.30 shows a plot in which the abscissa istime, measured starting from impact, and theordinate is depth below the pile top. The times that

Fig. 7.30 Graph relatesdistance fromapile headto a depthwhere a change in the pile cross section orsoil resistance occurs to the time it takes an impactpulse applied at the pile head and traveling at avelocity c to reach and then be reflected from thechange back to the pile head. Line I indicatesthe reflection due to impedance, II the reflectiondue to passive resistance R (modeled velocityproportional), and III the reflection from thepile toe.

Fig. 7.29 Pile Integrity Tester. (Courtesy of PileDynamics, Inc., Cleveland, Ohio.)

7.64 n Section Seven

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changes in wave characteristics due to pileimpedance or soil resistance are recorded at thepile head are represented along the time axis bysmall rectangles. The line from the origin extendingdownward to the right presents the position of thewave traveling with velocity c after impact. Wherea change in pile impedance Z occurs, at depth a andtime a/c, a line (I) extends diagonally upward to theright and indicates that the wave reaches the piletop at time 2a/c. Hence, with the time and wavevelocity known, the distance a can be calculated.Similarly, from time 2b/c, as indicated by line II, thedistance b from the pile top of the change in soilresistance R can be computed. Line III indicatesthat the wave from the toe at distance L from thepile head reaches the head at time 2L/c.

Dynamic analysis may be done in a signal-matching process or by a method that generates apile impedance profile from the measured pile-topdata. (F. Rausche et al., “A Formalized Procedurefor Quality Assessment of Cast-in-Place ShaftsUsing Sonic Pulse Echo Methods,” TransportationResearch Board, Washington, D.C. 1994.)

The low-strain integrity method is applicable toconcrete (cast-in-place and driven) and wood piles.Usually, piles are tested shortly after installationso that deficiencies may be detected early andcorrective measures taken during foundation con-struction and before erection of the superstructure.As for other nondestructive testing methods, theresults of measurements recorded may be dividedinto four main categories: (1) clear indication of asound pile, (2) clear indication of a seriousdefect, (3) indication of a somewhat defective pile,and (4) records do not support any conclusions.The foundation engineer, taking into considerationstructural, geotechnical, and other relevant factors,should decide between pile acceptability orrejection.

The low-strain integrity method may be used todetermine the length and condition of piles underexisting structures. (M. Hussein, G. Likins, andG. Goble, “Determination of Pile Lengths underExisting Structures,” Deep Foundations Institute,1992, www.dfi.org.)

The method has some limitations. For example,wave reflections coming from locations greaterthan about 35 pile diameters may be too weak to bedetected at the pile head with instrumentscurrently available. Also, gradual changes in pileimpedance may escape detection. Furthermore, themethodmay not yield reliable results for steel piles.

Concrete-filled steel pipe piles may be evaluatedwith this method.

7.21 Specification Notes

Specifications for pile installation should providerealistic criteria for pile location, alignment, andminimum penetration or termination drivingresistance. Particular attention should be given toprovisions for identification of pile heave andrelaxation and for associated remedial measures.Corrective actions for damaged or out-of-positionpiles should also be identified. Material quality andquality control should be addressed, especially forcast-in-place concrete piles. Tip protection of pilesis an important consideration for some types ofhigh-capacity, end-bearing piles or piles driventhrough obstructions. Other items that may beimportant are criteria for driving sequence inpile groups, preexcavation procedures, protectionagainst corrosive subsoils, and control of piledriving in proximity to open or recently concretedpile shells. Guidelines for selected specificationitems are in Table 7.10.

The following is a list of documents containingsample guidelines, standards, and specificationsrelated to deep foundations design and construc-tion. Clear, comprehensive, reasonable, and fairproject specifications greatly reduce the potentialfor disputes and costly delays.

“Guidelines for writing Construction Specificationsfor Piling” — Deep Foundations Institute, www.dfi.org.

“Recommended Design Specifications for DrivenBearing Piles” — Pile Driving Contractors Associ-ation, www.piledrivers.org.

“Standard Guidelines for the Design and Installa-tion of Pile Foundations” — American Society ofCivil Engineers, www.asce.org.

“Standards and Specifications for the FoundationDrilling Insustry” — International Association ofFoundation Drilling, www.adsc-iafd.com.

“Standard Specifications for Road and BridgeConstruction, Section 455: Structures Foundations”— Florida Department of Transportation, www.dot.state.fl.us.

“Standard Specification for the Construction ofDrilled Piers”—Americal Concrete Institute, www.aci-int.org.

Geotechnical Engineering n 7.65

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“Recommended Practice for Design, Manufactur-ing, and Installation of Prestressed ConcretePiling” — Precast/Prestressed Concrete Institute,www.pci.org.

“International Building Code — Chapter 18: Soilsand Foundations” — International Code Council,5203 Leesburg Pike, Falls Church, Virginia 22041.

7.22 Drilled Shafts

Drilled shafts are commonly used to transfer largeaxial and lateral loads to competent bearingmaterials by shaft or base resistance or both. Alsoknown as drilled piers, drilled-in caissons, or large-diameter bored piles, drilled shafts are cylindrical,cast-in-place concrete shafts installed by large-diameter, auger drilling equipment. Shaft diam-eters commonly range from 2.5 to 10 ft and lengthsfrom 10 to 150 ft, although shafts with dimensionswell outside these ranges can be installed. Shaftsmay be of constant diameter (straight shafts, Fig.7.31a) or may be underreamed (belled Fig. 7.31b) orsocketed into rock (Fig. 7.31c). Depending on load

requirements, shafts may be concreted with orwithout steel reinforcement.

Under appropriate foundation conditions, asingle drilled shaft is well-suited for support ofvery heavy concentrated loads; 2000 tons with rockbearing is not unusual.

Subsurface conditions favoring drilled shaftsare characterized by materials and groundwater

Fig. 7.31 Types of drilled shafts.

Table 7.10 Guide to Selected Specification Provisions

Position . . .within 6in of plan location (3in for pile groups with less than 5 piles)Plumbness . . . deviation from vertical shall not exceed 2% in any interval (4% from axis

for batter piles)Pile-hammer assembly Verification of the suitability of the proposed hammer assembly to drive the

designated piles shall be provided by wave equation or equivalentanalysis subject to the engineer’s approval.

Pile-driving leaders All piles shall be driven with fixed leaders sufficiently rigid to maintain pileposition and axial alignment during driving.

Driving criteria . . . to an elevation of at least ___ and/or to a terminal driving resistance of __blows/__in

Indicator piles Before start of production driving, indicator piles should be driven atlocations determined by the engineer. Continuous driving resistancerecords shall be maintained for each indicator pile.

Preboring Preboring immediately preceding pile installation shall extend to elevation___. The bore diameter for friction piles shall not be less than 1 or morethan 2in smaller than the pile diameter.

Heave* The elevation of pile butts or tips of CIPC pile shells shall be establishedimmediately after driving and shall be resurveyed on completion of thepile group. Should heave in excess of 1⁄4 in be detected, the piles shall beredriven to their initial elevation or as directed by the engineer.

Relaxation or setup† Terminal driving resistance of piles in-place for at least 24h shall be redrivenas directed by the engineer.

* Can be initially conducted on a limited number of pile groups and subsequently extended to all piles if required.† May be specified as part of initial driving operations.

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conditions that do not induce caving or squeezingof subsoils during drilling and concrete placement.High-capacity bearing levels at moderate depthsand the absence of drilling obstructions, such asboulders or rubble, are also favorable conditions.Current construction techniques allow drilledshafts to be installed in almost any subsurfacecondition, although the cost-effectiveness or relia-bility of the system will vary significantly.

7.22.1 Construction Methods forDrilled Shafts

In stable soil deposits, such as stiff clays, concretewith or without reinforcement may be placed inuncased shafts. Temporary casing, however, maybe employed during inspection of bearing con-ditions. Temporary casing may also be installed inthe shaft during or immediately after drilling toprevent soil intrusion into the concrete duringplacement. During this process, the height ofconcrete in the casing should at all times besufficient so that the weight more than counter-balances hydrostatic heads imposed by ground-water or by fluid trapped in the annular spacebetween the soil and the casing. The lack ofattention to this requirement is perhaps the greatestcontributor to drilled-shaft failures.

Unstable soil conditions encountered within alimited interval of shaft penetration can be handledby advancing a casing into stable subsoils belowthe caving zone via vibratory driving or byscrewing down the casing with a torque-barattachment. The hole is continued by augeringthrough the casing, which may subsequently beretracted during concrete placement or left in place.A shaft through unstable soils may also be ad-vanced without a casing if a weighted drilling fluid(slurry) is used to prevent caving.

For a limited unstable zone, underlain by rela-tively impervious soils, a casing can be screwedinto these soils so as to form a water seal. Thisallows the drilling slurry to be removed and theshaft continued through the casing and completedby normal concreting techniques.

The shaft may also be advanced entirely byslurry drilling techniques. With this method, con-crete is tremied in place so as to completely dis-place the slurry.

Reinforcing steel must be carefully designed tobe stable under the downward force exerted by

the concrete during placement. Utilization of rein-forcement that is not full length is not generallyrecommended where temporary casing is used tofacilitate concrete placement or slurry methods ofconstruction.

Concrete can be placed in shafts containing notmore than about 4 in of water (less for belledshafts). Free-fall placement can be used if an un-obstructed flow is achieved. Bottom-dischargehoppers centered on the shaft facilitate unrestrictedflow, whereas flexible conduits (elephant trunks)attached to the hopper can be used to guide con-crete fall in heavily reinforced shafts. Rigid tremiepipes are employed to place concrete in water or inslurry-filled shafts.

Equipment and Tools n Large-diameterdrills are crane- and truck-mounted, dependingon their size and weight. The capacity of the drill israted by its maximum continuous torque, ft-lb, andthe force exerted on the drilling tool. This force isthe weight of the Kelly bar (drill stem) plus theforce applied with some drills by their Kelly-bardown-crowd mechanism.

Downward force on the auger is a function ofthe Kelly-bar length and cross section. TelescopingKelly bars with cross sections up to 12 in squarehave been used to drill 10-ft-diameter shafts inearth to depths over 220 ft. Solid pin-connectedKelly sections up to 8 in square have also beeneffectively used for drilling deep holes. Additionaldown-crowd forces exerted by some drills are onthe order of 20 to 30 kips (crane-mounts) and 15 to50 kips (truck-mounts).

Drilling tools consisting of open helix (single-flight) and bucket augers are typically used forearth drilling and may be interchanged duringconstruction operations. To drill hard soil and softand weathered rock more efficiently, flight augersare fitted with hard-surfaced teeth. This type ofauger can significantly increase the rate of advancein some materials and provides a more equitabledefinition of “rock excavation” when compared tothe refusal of conventional earth augers. Flightaugers allow a somewhat faster operation and insome circumstances have a superior penetrationcapability. Bucket augers are usually more efficientfor excavating soft soils or running sands andprovide a superior bottom cleanout.

Belled shafts in soils and soft rocks are con-structed with special underreaming tools. Theseare usually limited in size to a diameter three times

Geotechnical Engineering n 7.67

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the diameter of the shaft. Hand mining techniquesmay be required where hard seams or otherobstructions limit machine belling.

Cutting tools consisting of roller bits or corebarrels are typically used to extend shafts intoharder rock and to form rock sockets. Multirollerbits are often used with a reverse circulation type ofrotary drilling rig. This technique, together withpercussion bits activated by pneumatically pow-ered drills, usually provide the most rapid advancein rock but have the disadvantage of requiringspecial drill types that may not be efficient in earthdrilling.

7.22.2 Construction QualityControl and Assurancefor Drilled Shafts

During the preparation of drilled-shaft designand construction specifications, special attentionshould be given to construction-related designfeatures, including shaft types, shaft-diametervariations, site trafficability, ground-loss potential,and protection of adjacent facilities. Proper tech-nical specifications and contract provisions for suchitems as payment for rock excavation and drillingobstructions are instrumental in preventing sig-nificant cost overruns and associated claims.

Some cost-reduction or quality-related precau-tions in preparation of drilled-shaft designs andspecifications are:

1. Minimize the number of different shaft sizes;extra concrete quantities for diameters largerthan actually needed are usually far less costlythan use of a multiplicity of drilling tools andcasings.

2. Delete a requirement for concrete vibration anduse concrete slumps not less than 6+1 in.

3. Do not leave a casing in place in oversized holesunless pressure grouting is used to preventground loss.

4. Shaft diameters should be at least 2.5 ft,preferably 3 ft.

Tolerances for location of drilled shafts shouldnot exceed 3 in or 1⁄24 of the shaft diameter, which-ever is less. Vertical deviation should not be morethan 2% of the shaft length or 12.5% of thediameter, whichever controls, except for specialconditions.

Provisions for proof testing are extremelyimportant for shafts designed for high-capacityend bearing. This is particularly true for bearingmaterials that may contain discontinuities or haverandom variations in quality. Small percussiondrills (jackhammers) are often used for proof test-ing and may be supplemented by diamond coring,if appropriate.

Because many drilled-shaft projects involvevariations in bearing levels that cannot be quan-tified during the design stage, the limitations of thebearing level and shaft quantities estimated for bid-ding purposesmust be clearly identified. Variationsin bearing level and quality are best accommodatedby specifications and contract provisions that facili-tate field changes. The continuous presence ofa qualified engineer inspector, experienced indrilled-shaft construction, is required to ensure thequality and cost-effectiveness of the construction.

7.22.3 Drilled-Shaft Design

Much of the design methodology for drilled shaftsis similar to that applied to pile foundationsand usually differs only in the manner in whichthe design parameters are characterized. Conse-quently, drilled-shaft design may be based onprecedent (experience), load testing, or staticanalyses. The ultimate-load design approach iscurrently the most common form of static analysisapplied, although load-deformation compatibilitymethods are being increasingly used (see Art. 7.18).

7.22.4 Skin Friction in CohesiveSoils

Ultimate skin friction for axially loaded shaftsdrilled into cohesive soils is usually evaluated byapplication of an empirically derived reduction(adhesion) factor to the undrained shear strength ofthe soil in contact with the shaft [see Eq. (7.37)]. Forconventionally drilled shafts in stiff clays (cu � 0.50tons/ft2), the adhesion factor a has been observedto range usually between 0.3 and 0.6.

Based on analysis of high-quality load-testresults, primarily in stiff, fissured Beaumont andLondon clays, a factors of 0.5 and 0.45 have beenrecommended. Unlike the criteria applied to piledesign, these factors are independent of cu, but arelargely dependent on construction methods andpractices. Reese has recommended that the shaftlength assumed to be effective in transferring load

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should be reduced by 5 ft to account for baseinteraction effects and that a similar reduction beapplied to account for surface effects such as soilshrinkage.

Table 7.11 lists recommended a factors forstraight shafts as a function of the normalized shearstrength cu=s

0vo (Art. 7.5.1) and plasticity index Ip

(Art. 7.4). These factors reflect conventional dry-hole construction methods and the influence of thestress history and plasticity of the soil in contactwith the shaft. The a factors in Table 7.11 may belinearly interpreted for specific values of cu=s

0vo and

Ip . To account for tip effects, the part of the shaftlocated 1 diameter above the base should beignored in evaluation of Qsu (see Art. 7.17).

Where shafts are drilled to bearing on relativelyincompressible materials, the amount of relativemovement between the soil and the shaft may beinsufficient to develop a substantial portion of theultimate skin friction, particularly for short, verystiff shafts. Under these circumstances, Qs shouldbe ignored in design or analyzed with load-dis-placement compatibility procedures.

Because belled shafts usually require largerdeformations than straight shafts to develop designloads, Qsu may be reduced for such shafts as aresult of a progressive degradation at relativedeformations greater than that required to developpeak values. Limited data on belled vs. straightshafts suggest a reduction in the a factor on theorder of 15% to account for the reduced shaftfriction of the belled shafts. It is also conservative toassume that there is no significant load transfer byfriction in that portion of the shaft located about 1diameter above the top of the bell.

There is some evidence that when shaft drillingis facilitated by the use of a weighted drilling fluid(mud slurry), there may be a substantial reductionin Qsu, presumably as a result of entrapment of the

slurry between the soil and shaft concrete. Wherethis potential exists, it has been suggested that the afactor be reduced about 40%.

(A. W. Skempton, “Summation, Symposium onLarge Bored Piles,” Institute of Civil Engineers,London; L. C. Reese, F. T. Toma, and M. W. O’Neill,“Behavior of Drilled Piers under Axial Loading,”ASCE Journal of Geotechnical Engineering Division,vol. 102, no. GT5, 1976; W. S. Gardner, “Investi-gation of the Effects of Skin Friction on thePerformance of Drilled Shafts in Cohesive Soils,”Report to U.S. Army Engineers Waterways Exper-iment Station (Contract no. DACA 39-80-C-0001),vol. 3, Vicksburg, Miss., 1981.)

7.22.5 Skin Friction in CohesionlessSoils

Ultimate skin friction in cohesionless soils can beevaluated approximately with Eq. (7.40). In theabsence of more definitive data, K in Eq. (7.40) maybe taken as 0.6 for loose sands and 0.7 for mediumdense to dense sands, on the assumption that thesoil-shaft interface friction angle is taken as f0 � 58.As for piles, limited test data indicate that theaverage friction stress fsu is independent of over-burden pressure for shafts drilled below a criticaldepth zc of from 10 (loose sand) to 20 (dense sand)shaft diameters. The limiting skin friction fi forshafts with ratio of length to diameter L/D � 25should appreciably not exceed 1.0 ton/ft2. Averagefsu may be less than 1.0 ton/ft2 for shafts longerthan about 80 ft.

Equation (7.52) approximately represents acorrelation between fsu and the average standardpenetration test blow count �NN within the em-bedded pile length recommended for shafts in sandwith effective L/D � 10. The stress fsu so computed,however, is less conservative than the foregoingdesign approach, particularly for �NN � 30 blows perfoot.

fsu ¼ 0:03 �NN � 1:6 tons=ft2 (7:52)

7.22.6 End Bearing on Soils

End bearing of drilled shafts in cohesive soils istypically evaluated as described for driven piles[Eq. (7.42)]. The shear-strength term in this equa-tion represents the average cu within a zone of 2

Table 7.11 Adhesion Factorsa forDrilled Shafts

PlasticityNormalized Shear Strength cu=s

0vo*

Index 0.3 or Less 1.0 2.5 or More

20 0.33 0.44 0.5530 0.36 0.48 0.6060 0.40 0.52 0.65

* Based on UU tests on good-quality samples selected so thatcu is not significantly influenced by the presence of fissures.

Geotechnical Engineering n 7.69

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diameters below the shaft space. For smaller shafts,the suggested reduction factor is 0.8.

End bearing in cohesionless soils can be esti-mated in accordance with Eq. (7.43) with the samecritical-depth limitations described for pile foun-dations. Nq for drilled shafts, however, has beenobserved to be significantly smaller than thatapplied to piles (see Fig. 7.19). Meyerhof hassuggested that Nq should be reduced by 50%.Alternatively, qu can be expressed in terms of theaverage SPT blow count �NN as:

qu ¼ 0:67 �NN � 40 tons=ft2 (7:53)

where qu ¼ ultimate base resistance at a settle-ment equivalent to 5% of the base diameter.(G. G. Meyerhof, “Bearing Capacity and Settlementof Pile Foundations,” ASCE Journal of GeotechnicalEngineering Division, vol. 102, no. GT3, 1976.)

7.22.7 Shaft Settlement

Drilled-shaft settlements can be estimated by load-deformation compatibility analyses (see Art. 7.18).Other methods used to estimate settlement ofdrilled shafts, singly or in groups, are identical tothose used for piles (Art. 7.19.5). These includeelastic, semiempirical elastic, and load-transfersolutions for single shafts drilled in cohesive orcohesionless soils. (H. G. Poulos and E. H. Davis,“Elastic Solutions for Soil and Rock Mechanics,”John Wiley & Sons, Inc., New York; A. S. Vesic,“Principles of Pile Foundation Design,” Soil Mech-anics Series no. 38, Duke University, Durham, N.C.,1975; H. Y. Fang, “Foundation Engineering Hand-book,” 2nd ed., VanNostrandReinhold,NewYork.)

Resistance to tensile and lateral loads bystraight-shaft drilled shafts should be evaluatedas described for pile foundations (see Art. 7.19).

7.22.8 Rock-Supported Shafts

Drilled shafts may be designed to be supported onrock or to be socketed into rock. Except for long,relatively small-diameter (comparatively compres-sible) shafts, conventional design ignores the skinfriction of belled or straight shafts founded onrelatively incompressible materials. Where shaftsare socketed in rock, the design capacity is con-sidered a combination of the sidewall shearingresistance (bond) and the end bearing of the socket.

In practice, both end-bearing and rock-socketdesigns are based on local experience, presumptivevalues in codes or semi empirical design methods.Latter methods based on field load test results arepreferred for design efficiency.

Bearing values on rock given in design codestypically range from 50 to 100 tons/ft2 for massivecrystalline rock, 20 to 50 tons/ft2 for sound foliatedrock, 15 to 25 tons/ft2 for sound sedimentary rock,8 to 10 tons/ft2 for soft and fractured rock, and 4 to8 tons/ft2 for soft shales.

The supporting ability of a specific rock type isprimarily dependent on the frequency, orientation,and size of the discontinuities within the rock massand the degree of weathering of the rock minerals.Consequently, application of presumptive bearingvalues is not recommended without specific localperformance correlations. (R. W. Woodward, W. S.Gardner, and D. M. Greer, “Drilled Pier Founda-tions,” McGraw-Hill Book Company, New York(books.mcgraw-hill.com).)

Some analyses relate the bearing values qu injointed rock to the uniaxial compressive (UC)strength of representative rock cores. These anal-yses indicate that qu should not be significantly lessthan UC, possibly excluding weak sedimentaryrocks such as compacted shales and siltstones.With a safety factor of 3, the maximum allowablebearing value qa can be taken as: qa � 0.3UC. Inmost instances, however, the compressibility of therockmass rather than rock strength governs. Elasticsolutions can be used to evaluate the settlement ofshafts bearing on rock if appropriate deformationmoduli of the rock mass Er can be determined.(H. G. Poulos and E. H. Davis, “Elastic Solutions forSoil and Rock Mechanics,” John Wiley & Sons,Inc., New York (www.wiley.com); D. U. Deere,A. J. Hendron, F. D. Patton, and E. J. Cording,“Breakage of Rock,” Eighth Symposium on RockMechanics, American Institute of Mining and Met-allurgical Engineers, Minneapolis, Minn., 1967;F. H. Kulhawy, “Geotechnical Model for RockFoundation Settlement,” ASCE Journal of Geotechni-cal Engineering Division, vol. 104, no. GT2, 1978.)

Concrete-rock bond stresses fR used for thedesign of rock sockets have been empiricallyestablished from a limited number of load tests.Typical values range from 70 to 200 psi, increasingwith rock quality. For good-quality rock, fR may berelated to the 28-day concrete strength f 0c and to theuniaxial compressive (UC) strength of rock cores.For rock with RQD � 50% (Table 7.3), fR can be

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estimated as 0:05f 0c or 0.05UC strength, whichever issmaller, except that fR should not exceed 250 psi. Asshown by Fig. 7.31, the ultimate rock-concrete bondfRu is significantly higher than fR except for veryhighUC values. (P. Rosenberg and N. L. Journeaux,“Friction and End-Bearing Tests on Bedrock forHigh-Capacity Socket Design,” Canadian Geotechni-cal Journal, vol. 13, no. 3, 1976.)

Design of rock sockets is conventionallybased on

Qd ¼ pdsLsfR þ p

4d2s qa (7:54)

where Qd ¼ allowable design load on rock socket

ds ¼ socket diameter

Ls ¼ socket length

fR ¼ allowable concrete-rock bond stress

qa ¼ allowable bearing pressure on rock

Load-distribution measurements show, however,that much less of the load goes to the base than isindicated by Eq. (7.54). This behavior is demon-strated by the data in Table 7.12, where Ls/ds is theratio of the shaft length to shaft diameter and Er/Ep

is the ratio of rock modulus to shaft modulus. Thefinite-element solution summarized in Table 7.12probably reflects a realistic trend if the averagesocket-wall shearing resistance does not exceed theultimate fR value; that is, slip along the socketsidewall does not occur.

A simplified design approach, taking intoaccount approximately the compatibility of thesocket and base resistance, is applied as follows:

1. Proportion the rock socket for design load Qd

with Eq. (7.54) on the assumption that theend-bearing stress is less than qa [say qa/4, which

is equivalent to assuming that the base loadQb ¼ (p=4)d2s qa=4].

2. Calculate Qb ¼ RQd, where R is the base-loadratio interpreted from Table 7.12.

3. If RQd does not equal the assumed Qb, repeatthe procedure with a new qa value until anapproximate convergence is achieved and q � qa.

The final design should be checked against theestablished settlement tolerance of the drilled shaft.(B. Ladanyi, discussion of “Friction and End-Bearing Tests on Bedrock,” Canadian GeotechnicalJournal, vol. 14, no. 1, 1977; H. G. Poulos andE. H. Davis, “Elastic Solutions for Rock and SoilMechanics,” John Wiley & Sons, Inc., New York(www.wiley.com).)

Following the recommendations of Rosenbergand Journeaux, a more realistic solution by theabove method is obtained if fRu is substituted for fR.Ideally, fRu should be determined from load tests. Ifthis parameter is selected from Fig. 7.32 or fromother data that are not site-specific, a safety factorof at least 1.5 should be applied to fRu in recognitionof the uncertainties associated with theUC strengthcorrelations. (P. Rosenberg and N. L. Journeaux,“Friction and End-Bearing Tests on Bedrock forHigh-Capacity Socket Design,” Canadian Geotechni-cal Journal, vol. 13, no. 3, 1976.)

7.22.9 Testing of Drilled Shafts

Static-load capacity of drilled shafts may beverified by either static-load or dynamic-loadtesting (Arts. 7.19 and 7.20). Testing by applyingstatic loads on the shaft head (conventional static-load test) or against the toe (Osterberg cell) pro-vides information on shaft capacity and generalbehavior. Dynamic-load testing in which pile-headforce and velocity under the impact of a fallingweight are measured with a Pile Driving Analyzerand subsequent analysis with the CAPWAPmethod (Art. 7.20.3) provide information on thestatic-load capacity and shaft-movement and shaft-soil load-transfer relationships of the shaft.

Structural integrity of a drilled shaft may beassessed after excavation or coring through the shaft.Low-strain dynamic-load testing with a Pile Integ-rity Tester (Art. 7.20.4), offers many advantages,however. Alternative integrity evaluation methodsare parallel seismic or cross-hole sonic logging.

For parallel seismic testing, a small casing isinserted into the ground near the tested shaft and

Table 7.12 Percent of Base Load Transmitted toRock Socket

Er/Ep

Ls/ds 0.25 1.0 4.0

0.5 54* 48 441.0 31 23 181.5 17* 12 8*2.0 13* 8 4

* Estimated by interpretation of finite-element solution; forPoisson’s ratio ¼ 0.26.

Geotechnical Engineering n 7.71

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to a greater depth than the shaft length. A hydro-phone is lowered into the casing to pick up thesignals resulting from blows on the shaft head froma small hand-held hammer. Inasmuch aswave velo-city in the soil and shaft are different, the unknownlength of the pile can be discerned from a series ofmeasurements. One limitation of this method is theneed to bore a hole adjacent to the shaft to be tested.

Cross-hole testing requires two full-length lon-gitudinal access tubes in the shaft. A transmitter islowered through one of the tubes to send a signalto a receiver lowered into the other tube. Thearrival time and magnitude of the received signalare interpreted to assess the integrity of the shaft

between the two tubes. For large-diameter shafts,more than two tubes may be needed for thoroughshaft evaluation. A disadvantage of this method isthe need to form two or more access tubes in theshaft during construction. Furthermore, randomtesting or evaluations of existing shafts may not bepossible with this method.

(C. L. Crowther, “Load Testing of DeepFoundations,” John Wiley & Sons, Inc., New York(www.wiley.com); “New Failure Load Criterion forLarge Diameter Bored Piles in Weathered Geo-meterials,” by Charles W. W. Ng, et al., ASCEJournal of Geotechnical and Geoenvironmental Engin-eering, vol. 127, no. 6, June 2001.)

Fig. 7.32 Chart relates the bond of a rock socket to the unconfined compressive strength of cores.

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Retaining Methods ForExcavation

The simplest method of retaining the sides of anexcavation in soil is to permit the soil to form anatural slope that will be stable even in thepresence of water. When there is insufficient spacefor this inside the excavation or when theexcavation sides must be vertical, constructionsuch as that described in the following mustbe used.

7.23 Caissons

Load-bearing enclosures known as caissons areformed in the ground, usually to protect excavationfor a foundation, aid construction of the substruc-ture, and serve as part of the permanent structure.Sometimes, a caisson is used to enclose a subsurfacespace to be used for such purposes as a pump well,machinery pit, or access to a deeper shaft or tunnel.Several caissons may be aligned to form a bridgepier, bulkhead, seawall, foundation wall for abuilding, or impervious core wall for an earth dam.

For foundations, caissons are used to facilitateconstruction of shafts or piers extending from nearthe surface of land or water to a bearing stratum.This type of construction can carry heavy loads togreat depths. Built of common structural materials,they may have any shape in cross section. Theyrange in size from about that of a pile to over 100 ftin length and width. Some small ones are con-sidered drilled shafts. Previously described con-struction methods for drilled shafts are employedbased on subsurface conditions and drilled shaftdepth and diameter.

Caissons often are installed by sinking themunder their own weight or with a surcharge. Theoperation is assisted by jacking, jetting, excavating,and undercutting. Care must be taken during thisoperation to maintain alignment. The caissons maybe built up as they sink, to permit construction tobe carried out at the surface, or they may becompletely prefabricated. Types of caissons usedfor foundation work are as follows:

Chicago caissons are used for constructingfoundation shafts through a thick layer of clay tohardpan or rock. The method is useful where thesoil is sufficiently stiff to permit excavation forshort distances without caving. A circular pit about5 ft deep is dug and lined with wood staves. This

vertical lagging is braced with two rings made withsteel channels. Then, 5 ft of soil is removed, and theoperation is repeated. If the ground is poor, shorterlengths are dug until the bearing stratum isreached. If necessary, the caissons can be belled atthe bottom to carry large loads. Finally, the hole isfilled with concrete. Minimum economical diam-eter for hand digging is 4 ft.

Sheeted piers or caissons are similarly con-structed, but the vertical lagging of wood or steel isdriven down during or before excavation. Thissystem usually is used for shallow depths in wetground.

In dry ground, horizontal wood sheeting maybe used. This is economical and necessary wherethere is inadequate vertical clearance. Louveredconstruction should be used to provide drainageand to permit packing behind the wood sheetingwhere soil will not maintain a vertical face longenough to permit insertion of the next sheet. Thistype of construction requires over-excavating sothat the wood sheets can be placed. Openings mustbe wide enough between sheets to allow backfillingand tamping, to correct the excavation irregula-rities and equalize pressure on all sides. Smallblocks may be inserted between successive sheetsto leave packing gaps. If the excavation is large,soldier beams, vertical cantilevers, can be driven tobreak up the long sheeting spans.

Benoto caissons up to 39 in in diameter may besunk through water-bearing sands, hardpan, andboulders to depths of 150 ft. Excavation is donewith a hammer grab, a single-line orange-peelbucket, inside a temporary, cylindrical steel casing.The hammer grab is dropped to cut into or breakup the soil. After impact, the blades close aroundthe soil. Then, the bucket is lifted out anddischarged. Boulders are broken up with heavy,percussion-type drills. Rock is drilled out by churndrills. To line the excavation, a casing is boltedtogether in 20-ft-deep sections, starting with acutting edge. A hydraulic attachment oscillates thecasing continuously to ease sinking and with-drawal, while jacks force the casing into theground. As concrete is placed, the jacks withdrawthe casing in a way that allows concreting of thecaisson. Benoto caissons are slower to place andmore expensive than drilled shafts, except in wetgranular material and where soil conditions are tootough for augers or rotating-bucket diggers.

Open caissons (Fig. 7.33) are enclosures withouttop and bottom during the lowering process. When

Geotechnical Engineering n 7.73

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used for pump wells and shafts, they often arecylindrical. For bridge piers, these caissons usuallyare rectangular and compartmented. The compart-ments serve as dredging wells, pipe passages, andaccess shafts. Dredging wells usually have 12- to16-ft clear openings to facilitate excavation withclamshell or orange-peel buckets.

An open caisson may be a braced steel shell thatis filled with concrete, except for the wells, as it issunk into place. Or a caisson may be constructedentirely of concrete.

Friction along the caisson sides may range from300 to over 1000 lb/ft2. So despite steel cuttingedges at the wall bottoms, the caissonmay not sink.Water and compressed-air jets may be used tolubricate the soil to decrease the friction. For thatpurpose, vertical jetting pipes should be embeddedin the outer walls.

If the caisson does not sink under its ownweight with the aid of jets when soil within hasbeen removed down to the cutting edge, thecaisson must be weighted. One way is to build ithigher, to its final height, if necessary. Otherwise, aplatform may have to be built on top and weightspiled on it, a measure that can be expensive.

Care must be taken to undercut the edgesevenly, or the caisson will tip. Obstructions andvariations in the soil also can cause uneven sinking.

When the caisson reaches the bearing strata, thebottom is plugged with concrete (Fig. 7.41b). Theplug may be placed by tremie or made by injectinggrout into the voids of coarse aggregate.

When a caisson must be placed through water,marine work sometimes may be converted to aland job by construction of a sand island. Fill isplaced until it projects above the water surface.

Fig. 7.33 Construction with an open concrete caisson.

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Then, the caisson is constructed and sunk as usualon land.

Pneumatic caissons contain at the base aworking chamber with compressed air at apressure equal to the hydrostatic pressure of thewater in the soil. Without the balancing pressure,the water would force soil from below up into acaisson. A working chamber clear of water alsopermits hand work to remove obstructions thatbuckets, air lifts, jets, and divers cannot. Thus, thedownward course of the caisson can be bettercontrolled. But sinking may be slower and moreexpensive, and compressed-air work requiresprecautions against safety and health hazards.

Access to the working chamber for workers,materials, and equipment is through air locks,usually placed at the top of the caisson (Fig. 7.34).Steel access cylinders 3 ft in diameter connect theair locks with the working chamber in largecaissons.

Entrance to the working chamber requires onlya short stay for a worker in an air lock. But thereturn stop may be lengthy, depending on thepressure in the chamber, to avoid the bends, orcaisson disease, which is caused by air bubbles in

muscles, joints, and the blood. Slow decompressiongives the body time to eliminate the excess air. Inaddition to slow decompression, it is necessary torestrict the hours worked at various pressures andlimit the maximum pressure to 50 psi aboveatmospheric or less. The restriction on pressurelimits the maximum depth at which compressed-air work can be done to about 115 ft. A medical, orrecompression, lock is also required on the site fortreatment of workers attacked by the bends.

Floating caissons are used when it is desirableto fabricate caissons on land, tow them intoposition, and sink them through water. They areconstructed much like open or pneumatic caissonsbut with a “false” bottom, “false” top, or buoyantcells. When floated into position, a caisson must bekept in alignment as it is lowered. A number ofmeans may be used for the purpose, includinganchors, templates supported on temporary piles,anchored barges, and cofferdams. Sinking gener-ally is accomplished by adding concrete to thewalls. When the cutting edges reach the bottom, thetemporary bulkheads at the base, or false bottoms,are removed since buoyancy no longer is necessary.With false tops, buoyancy is controlled withcompressed air, which can be released when thecaisson sits on the bottom. With buoyant cells,buoyancy is gradually lost as the cells are filledwith concrete.

Closed-box caissons are similar to floatingcaissons, except the top and bottom are permanent.Constructed on land, of steel or reinforced concrete,they are towed into position. Sometimes, the sitecan be dredged in advance to expose soil that cansafely support the caisson and loads that will beimposed on it. Where loads are heavy, however,this may not be practicable; then, the box caissonmay have to be supported on piles, but allowancecan be made for its buoyancy. This type of caissonhas been used for breakwaters, seawalls, andbridge-pier foundations.

Potomac caissons have been used in wide tidalrivers with deep water underlain by deep, softdeposits of sand and silt. Large timber mats areplaced on the river bottom, to serve as a templatefor piles and to retain tremie concrete. Long, steelpipe or H piles are driven in clusters, vertical andbattered, as required. Prefabricated steel or con-crete caissons are set on the mat over the pile clus-ters, to serve as permanent forms for concreteshafts to be supported on the piles. Then, concreteis tremied into the caissons. Since the caissons are

Fig. 7.34 Pneumatic caisson.Pressure inworkingchamber is above atmospheric.

Geotechnical Engineering n 7.75

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used only as forms, construction need not be soheavy as for conventional construction, where theymust withstand launching and sinking stresses,and cutting edges are not required.

(H. Y. Fang, “Foundation Engineering Hand-book,” 2nd ed., Van Nostrand Reinhold Company,New York.)

7.24 Dikes and Cribs

Earth dikes, when fill is available, are likely to bethe least expensive for keeping water out of anexcavation. If impervious material is not easilyobtained, however, a steel sheetpile cutoff wall mayhave to be driven along the dike, to permit pumpsto handle the leakage. With an impervious core inthe dike, wellpoints, deep-well pumps, or sumpsand ditches may be able to keep the excavationunwatered.

Timber cribs are relatively inexpensive excava-tion enclosures. Built on shore, they can be floatedto the site and sunk by filling with rock. The waterside may be faced with wood boards for water-tightness (Fig. 7.35). For greater watertightness,two lines of cribs may be used to support two linesof wood sheeting between which clay is tamped toform a “puddle” wall. Design of timber cribs

should provide ample safety against overturningand sliding.

7.25 Cofferdams

Temporary walls or enclosures for protecting anexcavation are called cofferdams. Generally, one ofthe most important functions is to permit work tobe carried out on a nearly dry site.

Cofferdams should be planned so that they canbe easily dismantled for reuse. Since they aretemporary, safety factors can be small, 1.25 to 1.5,when all probable loads are accounted for in thedesign. But design stresses should be kept lowwhen stresses, unit pressure, and bracing reactionsare uncertain. Design should allow for constructionloads and the possibility of damage from construc-tion equipment. For cofferdams in water, thedesign should provide for dynamic effect offlowing water and impact of waves. The height ofthe cofferdam should be adequate to keep outfloods that occur frequently.

7.25.1 Double-Wall Cofferdams

These may be erected in water to enclose largeareas. Double-wall cofferdams consist of two linesof sheetpiles tied to each other; the space between

Fig. 7.35 Timber crib with stone filling.

7.76 n Section Seven

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is filled with sand (Fig. 7.36). For sheetpiles drivento irregular rock, or gravel, or onto boulders, thebottom of the space between walls may be pluggedwith a thick layer of tremie concrete to seal gapsbelow the tips of the sheeting. Double-wallcofferdams are likely to be more watertight thansingle-wall ones and can be used to greater depths.

A berm may be placed against the outside faceof a cofferdam for stability. If so, it should beprotected against erosion. For this purpose, riprap,woven mattresses, streamline fins or jetties, orgroins may be used. If the cofferdam rests on rock,a berm needs to be placed on the inside only ifrequired to resist sliding, overturning, or shearing.On sand, an ample berm must be provided so thatwater has a long path to travel to enter thecofferdam (Fig. 7.36). (The amount of percolation isproportional to the length of path and the head.)Otherwise, the inside face of the cofferdam maysettle, and the cofferdam may overturn as waterpercolates under the cofferdam and causes a quick,or boiling, excavation bottom. An alternative to awide berm is wider spacing of the cofferdamwalls.This is more expensive but has the added advan-tage that the top of the fill can be used by con-struction equipment and for construction plant.

7.25.2 Cellular Cofferdams

Used in construction of dams, locks, wharves, andbridge piers, cellular cofferdams are suitable forenclosing large areas in deep water. These en-closures are composed of relatively wide units.

Average width of a cellular cofferdam on rockshould be 0.70 to 0.85 times the head of wateragainst the outside. When constructed on sand, acellular cofferdam should have an ample berm onthe inside to prevent the excavation bottom frombecoming quick (Fig. 7.37d).

Steel sheetpiles interlocked form the cells. Onetype of cell consists of circular arcs connected bystraight diaphragms (Fig. 7.37a). Another typecomprises circular cells connected by circular arcs(Fig. 7.37b). Still another type is the cloverleaf,composed of large circular cells subdivided bystraight diaphragms (Fig. 7.37c). The cells are filledwith sand. The internal shearing resistance of thesand contributes substantially to the strength of thecofferdam. For this reason, it is unwise to fill acofferdamwith clay or silt. Weepholes on the insidesheetpiles drain the fill, thus relieving the hydro-static pressure on those sheets and increasing theshear strength of the fill.

In circular cells, lateral pressure of the fill causesonly ring tension in the sheetpiles. Maximum stressin the pile interlocks usually is limited to 8000 lb/lin in. This in turn limits the maximum diameter ofthe circular cells. Because of numerous uncertain-ties, this maximum generally is set at 60 ft. Whenlarger-size cells are needed, the cloverleaf type maybe used.

Circular cells are preferred to the diaphragmtype because each circular cell is a self-supportingunit. It may be filled completely to the top beforeconstruction of the next cell starts. (Unbalancedfills in a cell may distort straight diaphragms.)

Fig. 7.36 Double-wall cofferdam.

Geotechnical Engineering n 7.77

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When a circular cell has been filled, the top may beused as a platform for construction of the next cell.Also, circular cells require less steel per linear footof cofferdam. The diaphragm type, however, maybe made as wide as desired.

When the sheetpiles are being driven, care mustbe taken to avoid breaking the interlocks. Thesheetpiles should be accurately set and plumbedagainst a structurally sound template. They shouldbe driven in short increments, so that when unevenbedrockor boulders are encountered, driving canbestopped before the cells or interlocks are damaged.Also, all the piles in a cell should be started until thecell is ringed. This can reduce jamming troubleswith the last piles to be installed for the cell.

7.25.3 Single-Wall Cofferdams

These form an enclosure with only one line ofsheeting. If there will be no water pressure on thesheeting, they may be built with soldier beams(piles extended to the top of the enclosure) andhorizontal wood lagging (Fig. 7.38). If there will bewater pressure, the cofferdam may be constructedof sheetpiles. Although they require less wallmaterial than double-wall or cellular cofferdams,single-wall cofferdams generally require bracingon the inside. Also, unless the bottom is driven intoa thick, impervious layer, theymay leak excessively

at the bottom. There may also be leakage atinterlocks. Furthermore, there is danger of floodingand collapse due to hydrostatic forces when thesecofferdams are unwatered.

For marine applications, therefore, it is advan-tageous to excavate, drive piles, and place a seal oftremie concrete without unwatering single-wallsheetpile cofferdams. Often, it is advisable to pre-dredge the area before the cofferdam is construc-ted, to facilitate placing of bracing and to removeobstructions to pile driving. Also, if blasting isnecessary, it would severely stress the sheeting andbracing if done after they were installed.

For buildings, single-wall cofferdams must becarefully installed. Small movements and conse-quent loss of ground usually must be prevented toavoid damaging neighboring structures, streets,and utilities. Therefore, the cofferdams must beamply braced. Sheeting close to an existing struc-ture should not be a substitute for underpinning.

Bracing n Cantilevered sheetpiles may be usedfor shallow single-wall cofferdams in water or onland where small lateral movement will not betroublesome. Embedment of the piles in the bottommust be deep enough to insure stability. Designusually is based on the assumptions that lateralpassive resistance varies linearly with depth andthe point of inflection is about two-thirds the

Fig. 7.37 Cellular sheetpile cofferdam.

7.78 n Section Seven

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embedded length below the surface. In general,however, cofferdams require bracing.

Cofferdams may be braced in many ways.Figure 7.39 shows some commonly used methods.Circular cofferdams may be braced with horizontalrings (Fig. 7.39a). For small rectangular cofferdams,horizontal braces, or wales, along sidewalls andend walls may be connected to serve only as struts.For larger cofferdams, diagonal bracing (Fig. 7.39b)or cross-lot bracing (Fig. 7.39d and e) is necessary.When space is available at the top of an excavation,pile tops can be anchored with concrete dead men(Fig. 7.39c). Where rock is close, the wall can be tiedback with tensionedwires or bars that are anchoredin grouted sockets in the rock (Fig. 7.40). See alsoArt. 7.40.4.

Horizontal cross braces should be spaced tominimize interference with excavation, form con-struction, concreting, and pile driving. Spacing of12 and 18 ft is common. Piles and wales selectedshould be strong enough as beams to permit suchspacing. In marine applications, divers often haveto install the wales and braces underwater. Toreduce the amount of such work, tiers of bracing

may be prefabricated and lowered into thecofferdam from falsework or from the top set ofwales and braces, which is installed above thewater surface. In some cases, it may be advan-tageous to prefabricate and erect the whole cage ofbracing before the sheetpiles are driven. Then, thecage, supported on piles, can serve also as atemplate for driving the sheetpiles.

All wales and braces should be forced intobearing with the sheeting by wedges and jacks.

When pumping cannot control leakage into acofferdam, excavation may have to be carried outin compressed air. This requires a sealed workingchamber, access shafts, and air locks, as for pneu-matic caissons (Art. 7.23). Other techniques, such asuse of a tremie concrete seal or chemical solidifica-tion or freezing of the soil, if practicable, however,will be more economical.

Braced sheetpiles may be designed as continu-ous beams subjected to uniform loading for earthand to loading varying linearly with depth forwater (Art. 7.27). (Actually, earth pressure dependson the flexibility of the sheeting and relativestiffness of supports.) Wales may be designed for

Fig. 7.38 Soldier beams and wood lagging retain the sides of an excavation.

Geotechnical Engineering n 7.79

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uniform loading. Allowable unit stresses in thewales, struts, and ties may be taken at half theelastic limit for the materials because the construc-tion is temporary and the members are exposed toview. Distress in a member can easily be detectedand remedial steps taken quickly.

Soldier beams and horizontal wood sheetingare a variation of single-wall cofferdams often usedwhere impermeability is not required. The soldierbeams, or piles, are driven vertically into theground to below the bottom of the proposed

excavation. Spacing usually ranges from 5 to 10 ft(Table 7.13). (The wood lagging can be used in thethicknesses shown in Table 7.13 because of archingof the earth between successive soldier beams.)

As excavation proceeds, the wood boards areplaced horizontally between the soldiers (Fig. 7.38).Louvers or packing spaces, 1 to 2 in high, are leftbetween the boards so that earth can be tampedbehind them to hold them in place. Hay may alsobe stuffed behind the boards to keep the groundfrom running through the gaps. The louvers permit

Fig. 7.39 Types of cofferdam bracing include compression rings; bracing, diagonal (rakers) or cross lot;wales, and tiebacks.

7.80 n Section Seven

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drainage of water, to relieve hydrostatic pressureon the sheeting and thus allow use of a lighterbracing system. The soldiers may be braceddirectly with horizontal or inclined struts; or walesand braces may be used.

Advantages of soldier-beam constructioninclude fewer piles; the sheeting does not have to

extend below the excavation bottom, as dosheetpiles; and the soldiers can be driven moreeasily in hard ground than can sheetpiles. Varyingthe spacing of the soldiers permits avoidance ofunderground utilities. Use of heavy sections forthe piles allows wide spacing of wales and braces.But the soldiers and lagging, as well as sheetpiles,are no substitute for underpinning; it is necessaryto support and underpin even light adjoiningstructures.

Liner-plate cofferdams may be used for exca-vating circular shafts. The plates are placed inhorizontal rings as excavation proceeds. Stampedfrom steel plate, usually about 16 in high and 3 ftlong, light enough to be carried by one person, linerplates have inward-turned flanges along all edges.Top and bottom flanges provide a seat forsuccessive rings. End flanges permit easy bolting

Fig. 7.40 Vertical section shows prestressed tiebacks for soldier beams.

Table 7.13 Usual Maximum Spans ofHorizontal Sheeting with Soldier Piles, ft

NominalThickness ofSheeting, in

In Well-DrainedSoils

In CohesiveSoils with LowShear Resistance

2 5 4.53 8.5 64 10 8

Geotechnical Engineering n 7.81

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Fig. 7.41 Slurry-trenchmethod for constructing a continuous concrete wall: (a) Excavating one section;(b) concreting one section while another is being excavated.

7.82 n Section Seven

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of adjoining plates in a ring. The plates also arecorrugated for added stiffness. Large-diametercofferdams may be constructed by bracing theliner plates with steel beam rings.

Vertical-lagging cofferdams, with horizontal-ring bracing, also may be used for excavatingcircular shafts. The method is similar to that usedfor Chicago caissons (Art. 7.23). It is similarlyrestricted to soils that can stand without support indepths of 3 to 5 ft for a short time.

Slurry trenches may be used for constructingconcrete walls. The method permits building a wallin a trench without the earth sides collapsing.While excavation proceeds for a 24- to 36-in-widetrench, the hole is filled with a bentonite slurrywith a specific gravity of 1.05 to 1.10 (Fig. 7.41a).The fluid pressure against the sides and caking ofbentonite on the sides prevent the earth walls of thetrench from collapsing. Excavation is carried out asection at a time. A section may be 20 ft long and asmuch as 100 ft deep. When the bottom of the wall isreached in a section, reinforcing is placed in thatsection. (Tests have shown that the bond of thereinforcing to concrete is not materially reduced bythe bentonite.) Then, concrete is tremied into thetrench, replacing the slurry, which may flow intothe next section to be excavated or be pumped intotanks for reuse in the next section (Fig. 7.41b). Themethod has been used to construct cutoffs fordams, cofferdams, foundations, walls of buildings,and shafts.

7.26 Soil Solidification

Grouting is the injection of cement or chemicalsinto soil or rock to enhance engineering properties.During the past 20 years significant developmentsin materials and equipment have transformedgrouting from art to science. See Federal HighwayAdministration publication “Ground Improve-ment Technical Summaries” Publication No.FHWA-SA-98-086 December 1999 for an extensiveprimer on grouting techniques, applications andprocedures.

Freezing is another means of solidifying water-bearing soils where obstructions or depth precludepile driving. It can be used for deep shaft exca-vations and requires little material for temporaryconstruction; the refrigeration plant has highsalvage value. But freezing the soil may take avery long time. Also, holes have to be drilled below

the bottom of the proposed excavation for insertionof refrigeration pipes.

(L. White and E. A. Prentis, “Cofferdams,”Columbia University Press, New York; H. Y. Fang,“Foundation Engineering Handbook,” 2nd ed.,Van Nostrand Reinhold Company, New York.)

7.27 Lateral Active Pressureson Retaining Walls

Water exerts against a vertical surface a horizontalpressure equal to the vertical pressure. At anylevel, the vertical pressure equals the weight of a1-ft2 column of water above that level. Hence, thehorizontal pressure p, lb/ft3 at any level is

p ¼ wh (7:55)

where w ¼ unit weight of water, lb/ft3

h ¼ depth of water, ft

The pressure diagram is triangular (Fig. 7.42).Equation (7.55) also can be written

p ¼ Kwh (7:56)

where K ¼ pressure coefficient ¼ 1.00.

Fig. 7.42 Pressure diagram for water.

Geotechnical Engineering n 7.83

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The resultant, or total, pressure, lb/lin ft, rep-resented by the area of the hydrostatic-pressurediagram, is

P ¼ Kwh2

2(7:57)

It acts at a distance h/3 above the base of thetriangle.

Soil also exerts lateral pressure. But the amountof this pressure depends on the type of soil, itscompaction or consistency, and its degree ofsaturation, and on the resistance of the structureto the pressure. Also, the magnitude of passivepressure differs from that of active pressure.

Active pressure tends to move a structure inthe direction in which the pressure acts. Passivepressure opposes motion of a structure.

Retaining walls backfilled with cohesionlesssoils (sands and gravel) tend to rotate slightlyaround the base. Behind such a wall, a wedge ofsandABC (Fig. 7.43a) tends to shear along planeAC.C. A. Coulomb determined that the ratio of slidingresistance to sliding force is a minimum when ACmakes an angle of 458 þ f/2 with the horizontal,

where f is the angle of internal friction of thesoil, deg.

For triangular pressure distribution (Fig. 7.43b),the active lateral pressure of a cohesionless soil at adepth h, ft, is

p ¼ Kawh (7:58)

where Ka ¼ coefficient of active earth pressure

w ¼ unit weight of soil, lb/ft3

The total active pressure, lb/lin ft, is

Ea ¼ Kawh2

2(7:59)

Because of frictional resistance to sliding at theface of the wall, Ea is inclined at an angle dwith thenormal to the wall, where d is the angle of wallfriction, deg (Fig. 7.43a). If the face of the wall isvertical, the horizontal active pressure equalsEa cos d. If the face makes an angle b with thevertical (Fig. 7.43a), thepressureequalsEa cos (d þ b).The resultant acts at a distance of h/3 above the baseof the wall.

If the ground slopes upward from the top of thewall at an angle a, deg, with the horizontal, then forcohesionless soils

Fig. 7.43 Free-standing wall with sand backfill (a) is subjected to (b) triangular pressure distribution.

7.84 n Section Seven

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Table 7.14 Active-Lateral-Pressure Coefficients Ka

f

108 158 208 258 308 358 408

a ¼ 0 0.70 0.59 0.49 0.41 0.33 0.27 0.22a ¼ 108 0.97 0.70 0.57 0.47 0.37 0.30 0.24

b ¼ 0 a ¼ 208 — — 0.88 0.57 0.44 0.34 0.27a ¼ 308 — — — — 0.75 0.43 0.32a ¼ f 0.97 0.93 0.88 0.82 0.75 0.67 0.59

a ¼ 0 0.76 0.65 0.55 0.48 0.41 0.43 0.29a ¼ 108 1.05 0.78 0.64 0.55 0.47 0.38 0.32

b ¼ 108 a ¼ 208 — — 1.02 0.69 0.55 0.45 0.36a ¼ 308 — — — — 0.92 0.56 0.43a ¼ f 1.05 1.04 1.02 0.98 0.92 0.86 0.79

a ¼ 0 0.83 0.74 0.65 0.57 0.50 0.43 0.38a ¼ 108 1.17 0.90 0.77 0.66 0.57 0.49 0.43

b ¼ 208 a ¼ 208 — — 1.21 0.83 0.69 0.57 0.49a ¼ 308 — — — — 1.17 0.73 0.59a ¼ f 1.17 1.20 1.21 1.20 1.17 1.12 1.06

a ¼ 0 0.94 0.86 0.78 0.70 0.62 0.56 0.49a ¼ 108 1.37 1.06 0.94 0.83 0.74 0.65 0.56

b ¼ 308 a ¼ 208 — — 1.51 1.06 0.89 0.77 0.66a ¼ 308 — — — — 1.55 0.99 0.79a ¼ f 1.37 1.45 1.51 1.54 1.55 1.54 1.51

Table 7.15 Angles of Internal Friction and Unit Weights of Soils

Types of SoilDensity orConsistency

Angle of InternalFriction f, deg

Unit Weight w,lb/ft3

Coarse sand or sand and gravel Compact 40 140Loose 35 90

Medium sand Compact 40 130Loose 30 90

Fine silty sand or sandy silt Compact 30 130Loose 25 85

Uniform silt Compact 30 135Loose 25 85

Clay-silt Soft to medium 20 90–120

Silty clay Soft to medium 15 90–120

Clay Soft to medium 0–10 90–120

Geotechnical Engineering n 7.85

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Ka ¼ cos2 (f� b)

cos2 b cos (dþb) 1þffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffisin (fþ d) sin (f� a)

cos (dþ b) cos (a� b)

s" #�2

(7:60)

The effect of wall friction on Ka is small and usuallyis neglected. For d ¼ 0,

Ka ¼ cos2 (f� b)

cos3 b 1þffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffisinf sin (f� a)

cosb cos (a� b)

s" #2(7:61)

Table 7.14 lists values of Ka determined from Eq.(7.61). Approximate values of f and unit weightsfor various soils are given in Table 7.15.

For level ground at the top of the wall (a ¼ 0):

Ka ¼ cos2 (f� b)

cos3 b 1þ sinf

cosb

� �2(7:62)

If in addition, the back face of the wall is vertical(b ¼ 0), Rankine’s equation is obtained:

Ka ¼ 1� sinf

1þ sinf(7:63)

Coulomb derived the trigonometric equivalent:

Ka ¼ tan2 458� f

2

� �(7:64)

The selection of the wall friction angle should becarefully applied as it has a significant effect on theresulting earth pressure force.

Unyielding walls retaining walls backfilledwith sand and gravel, such as the abutment wallsof a rigid-frame concrete bridge or foundationwalls braced by floors, do not allow shearingresistance to develop in the sand along planes thatcan be determined analytically. For such walls,triangular pressure diagramsmay be assumed, andKa may be taken equal to 0.5.

Braced walls retaining cuts in sand (Fig. 7.44a)are subjected to earth pressure gradually and

Fig. 7.44 Braced wall retaining sand (a) may have to resist pressure distributions of the type shown in(b). (c) Uniform pressure distribution may be assumed for design.

7.86 n Section Seven

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develop resistance in increments as excavationproceeds and braces are installed. Such walls tendto rotate about a point in the upper portion. Hence,the active pressures do not vary linearly withdepth. Field measurements have yielded a varietyof curves for the pressure diagram, of which twotypes are shown in Fig. 7.44b. Consequently, someauthorities have recommended a trapezoidal pres-sure diagram, with a maximum ordinate

p ¼ 0:8Kawh (7:65)

Ka may be obtained from Table 7.14. The totalpressure exceeds that for a triangular distribution.

Figure 7.45 shows earth-pressure diagramsdeveloped for a sandy soil and a clayey soil. Inboth cases, the braced wall is subjected to a 3-ft-deep surcharge, and height of wall is 34 ft. For thesandy soil (Fig. 7.45a), Fig. 7.45b shows the pressurediagram assumed. The maximum pressure can be

obtained from Eq. (7.65), with h ¼ 34 þ 3 ¼ 37 ftand Ka assumed as 0.30 and w as 110 lb/ft3.

p1 ¼ 0:8� 0:3� 110� 37 ¼ 975 lb=ft2

The total pressure is estimated as

P ¼ 0:8� 975� 37 ¼ 28,900 lb=lin ft

The equivalent maximum pressure for a trapezoi-dal diagram for the 34-ft height of the wall then is

p ¼ 28,900

0:8� 34¼ 1060 lb=ft2

Assumption of a uniform distribution (Fig. 7.44c),however, simplifies the calculations and has little orno effect on the design of the sheeting and braces,which should be substantial to withstand construc-tion abuses. Furthermore, trapezoidal loadingterminating at the level of the excavation may notapply if piles are driven inside the completed

Fig. 7.45 Assumed trapezoidal diagrams for a bracedwall in soils described by boring logs in (a) and (d).

Geotechnical Engineering n 7.87

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excavation. The shocks may temporarily decreasethe passive resistance of the sand in which the wallis embedded and lower the inflection point. Thiswould increase the span between the inflectionpoint and the lowest brace and increase thepressure on that brace. Hence, uniform pressuredistribution may be more applicable than trapezoi-dal for such conditions.

See Note for free-standing walls.Flexible retaining walls in sand cuts are sub-

jected to pressures that depend on the fixity of theanchorage. If the anchor moves sufficiently or thetie from the anchor to the upper portion of thebulkhead stretches enough, the bulkhead mayrotate slightly about a point near the bottom. In thatcase, the sliding-wedge theory may apply. Thepressure distribution may be taken as triangular,and Eqs. (7.58) to (7.64) may be used. But if theanchor does not yield, then pressure distributionsmuch like those in Fig. 7.44b for a braced cut mayoccur. Either a trapezoidal or uniform pressuredistribution may be assumed, with maximumpressure given by Eq. (7.65). Stresses in the tieshould be kept low because it may have to resistunanticipated pressures, especially those resultingfrom a redistribution of forces from soil arching.(Federal Highway Administration GeotechnicalEngineering Circular No.4 “Ground Anchors andAnchored Systems” FHWA-IF-99015, June 1999).

Free-standing walls retaining plastic-clay cuts(Fig. 7.46a) may have to resist two types of activelateral pressure, both with triangular distribution. Inthe short term the shearing resistance is due to co-hesion only, a clay bank may be expected to standwith a vertical face without support for a height, ft, of

h0 ¼ 2c

w(7:66)

where 2c ¼ unconfined compressive strength ofclay, lb/ft2

w ¼ unit weight of clay, lb/ft3

So if there is a slight rotation of the wall about itsbase, the upper portion of the clay cut will standvertically without support for a depth h0. Belowthat, the pressure will increase linearly with depthas if the clay were a heavy liquid (Fig. 7.46b):

p ¼ wh� 2c

The total pressure, lb/lin ft, then is

Ea ¼ w

2h� 2c

w

� �2

(7:67)

It acts at a distance (h 2 2c/w)/3 above the base ofthe wall. These equations assume wall friction iszero, the back face of the wall is vertical, and theground is level.

Fig. 7.46 Free-standing wall retaining clay (a) may have to resist the pressure distribution shown in (b)or (d). For mixed soils, the distribution may approximate that shown in (c).

7.88 n Section Seven

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In time the clay will reach its long termstrength and the pressure distribution may becomeapproximately triangular (Fig. 7.46d) from the topof the wall to the base. The pressures then may becalculated from Eqs. (7.58) to (7.64) with an ap-parent angle of internal friction for the soil (forexample, see the values of f in Table 7.15). The wallshould be designed for the pressures producingthe highest stresses and overturning moments.

Note: The finer the backfill material, the morelikely it is that pressures greater than active willdevelop, because of plastic deformations, water-level fluctuation, temperature changes, and othereffects. As a result, it would be advisable to use indesign at least the coefficient for earth pressure atrest:

Ko ¼ 1� sinf (7:68)

The safety factor should be at least 2.5.Clay should not be used behind retaining walls,

where other economical alternatives are available.The swelling type especially should be avoidedbecause it can cause high pressures and progres-sive shifting or rotation of the wall.

For a mixture of cohesive and cohesionlesssoils, the pressure distribution may temporarilybe as shown in Fig. 7.46c. The height, ft, of theunsupported vertical face of the clay is

h00 ¼ 2c

w tan (458� f=2)(7:69)

The pressure at the base is

p ¼ wh tan2 458� f

2

� �� 2c tan 458� f

2

� �(7:70)

The total pressure, lb/lin ft, is

Ea ¼ w

2h tan 458� f

2

� �� 2c

w

� �2(7:71)

It acts at a distance (h� h00)/3 above the base of thewall.

Braced walls retaining clay cuts (Fig. 7.47a) alsomay have to resist two types of active lateralpressure. As for sand, the pressure distributionmay temporarily be approximated by a trapezoi-dal diagram (Fig. 7.47b). On the basis of fieldobservations, R. B. Peck has recommended a max-imum pressure of

p ¼ wh� 4c (7:72)

and a total pressure, lb/lin ft, of

Ea ¼ 1:55h

2(wh� 4c) (7:73)

[R. B. Peck, “Earth Pressure Measurements in OpenCuts, Chicago (Ill.) Subway,” Transactions, AmericanSociety of Civil Engineers, 1943, pp. 1008–1036.]

Figure 7.47c shows a trapezoidal earth-pressurediagram determined for the clayey-soil condi-tion of Fig. 7.47d. The weight of the soil is takenas 120 lb/ft3; c is assumed as zero and the active-

Fig. 7.47 Braced wall retaining clay (a) may have to resist pressures approximated by the pressuredistribution in (b) or (d). Uniform distribution (c) may be assumed in design.

Geotechnical Engineering n 7.89

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lateral-pressure coefficient as 0.3. Height of thewall is 34 ft, surcharge 3 ft. Then, the maximumpressure, obtained from Eq. (7.58) since the soil isclayey, not pure clay, is

p1 ¼ 0:3� 120� 37 ¼ 1330 lb=ft2

From Eq. (7.73) with the above assumptions, thetotal pressure is

P ¼ 1:55

2� 37� 1330 ¼ 38,100 lb=lin ft

The equivalent maximum pressure for a trapezoi-dal diagram for the 34-ft height of wall is

p ¼ 38,100

34� 2

1:55¼ 1450 lb=ft2

To simplify calculations, a uniform pressure dis-tribution may be used instead (Fig. 7.47c).

If after a time the clay should attain a con-solidated equilibrium state, the pressure distribu-tion may be better represented by a triangulardiagram ABC (Fig. 7.47d), as suggested by G. P.Tschebotarioff. The peak pressure may be assumedat a distance of kh ¼ 0.4h above the excavation levelfor a stiff clay; that is, k ¼ 0.4. For a medium clay, kmay be taken as 0.25, and for a soft clay, as 0. Forcomputing the pressures, Ka may be estimatedfrom Table 7.14 with an apparent angle of frictionobtained from laboratory tests or approximatedfrom Table 7.15. The wall should be designed forthe pressures producing the highest stresses andoverturning moments.

See also Note for free-standing walls.Flexible retaining walls in clay cuts and

anchored near the top similarly should be checkedfor two types of pressures. When the anchor islikely to yield slightly or the tie to stretch, thepressure distribution in Fig. 7.47d with k ¼ 0 maybe applicable. For an unyielding anchor, any of thepressure distributions in Fig. 7.47 may be assumed,as for a braced wall. The safety factor for design ofties and anchorages should be at least twice thatused in conventional design. See also Note for free-standing walls.

Backfill placed against a retaining wall shouldpreferably be sand or gravel (free draining) tofacilitate drainage. Also, weepholes should beprovided through the wall near the bottom and adrain installed along the footing, to conduct waterfrom the back of the wall and prevent buildup ofhydrostatic pressures.

Saturated or submerged soil imposes substan-tially greater pressure on a retaining wall than dryor moist soil. The active lateral pressure for a soil-fluid backfill is the sum of the hydrostatic pressureand the lateral soil pressure based on the buoyedunit weight of the soil. This weight roughly may be60% of the dry weight.

Surcharge, or loading imposed on a backfill,increases the active lateral pressure on a wall andraises the line of action of the total, or resultant,pressure. A surcharge ws, lb/ft2, uniformly dis-tributed over the entire ground surface may betaken as equivalent to a layer of soil of the sameunit weightw as the backfill and with a thickness ofws/w. The active lateral pressure, lb/ft2, due to thesurcharge, from the backfill surface down, thenwillbe Kaws. This should be added to the lateralpressures that would exist without the surcharge.Ka may be obtained from Table 7.14.

(A. Caquot and J. Kerisel, “Tables for Calc-ulation of Passive Pressure, Active Pressure,and Bearing Capacity of Foundations,” Gauthier-Villars, Paris.)

7.28 Passive Lateral Pressureon Retaining Walls andAnchors

As defined in Art. 7.27, active pressure tends tomove a structure in the direction in which pressureacts, whereas passive pressure opposes motion of astructure.

Passive pressures of cohesionless soils, resistingmovement of a wall or anchor, develop because ofinternal friction in the soils. Because of frictionbetween soil and wall, the failure surface is curved,not plane as assumed in the Coulomb sliding-wedge theory (Art. 7.27). Use of the Coulombtheory yields unsafe values of passive pressurewhen the effects of wall friction are included.

Total passive pressure, lb/lin ft, on a wall oranchor extending to the ground surface (Fig. 7.48a)may be expressed for sand in the form

P ¼ Kpwh2

2(7:74)

where Kp ¼ coefficient of passive lateral pressure

w ¼ unit weight of soil, lb/ft3

h ¼ height of wall or anchor to groundsurface, ft

7.90 n Section Seven

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The pressure distribution usually assumed forsand is shown in Fig. 7.48b. Table 7.16 lists values ofKp for a vertical wall face (b ¼ 0) and horizontalground surface (a ¼ 0), for curved surfaces offailure. (Many tables and diagrams for determiningpassive pressures are given in A. Caquot and J.Kerisel, “Tables for Calculation of Passive Pressure,Active Pressure, and Bearing Capacity of Founda-tions,” Gauthier-Villars, Paris.)

Since a wall usually transmits a downwardshearing force to the soil, the angle of wall frictiond correspondingly is negative (Fig. 7.48a). For embed-ded portions of structures, such as anchored sheetpilebulkheads, d and the angle of internal friction f ofthe soil reach their peak values simultaneously indense sand. For those conditions, if specific infor-mation is not available, d may be assumed as 22⁄3f(for f . 308). For such structures as a heavy anchorblock subjected to a horizontal pull or thrust, d maybe taken as2f/2 for dense sand. For those cases, thewall friction develops as the sand is pushed upwardby the anchor and is unlikely to reach its maximum

value before the internal resistance of the sand isexceeded.

When wall friction is zero (d ¼ 0), the failuresurface is a plane inclined at an angle of 458 2 f/2with the horizontal. The sliding-wedge theory thenyields

Kp ¼ cos2 (fþ b)

cos3 b 1�ffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffisinf sin (fþ a)

cosb cos (a� b)

s" #2(7:75)

When the ground is horizontal (a ¼ 0):

Kp ¼ cos2 (fþ b)

cos3 b(1� sinf= cosb)2(7:76)

If, in addition, the back face of the wall is vertical(b ¼ 0):

Kp ¼ 1þ sinf

1� sinf¼ tan2 458þ f

2

� �¼ 1

Ka(7:77)

Fig. 7.48 Passive pressure on a wall (a) may vary as shown in (b) for sand or as shown in (c) for clay.

Table 7.16 Passive Lateral-Pressure Coefficients Kp*

f ¼ 108 f ¼ 158 f ¼ 208 f ¼ 258 f ¼ 308 f ¼ 358 f ¼ 408

d ¼ 0 1.42 1.70 2.04 2.56 3.00 3.70 4.60d ¼ 2f/2 1.56 1.98 2.59 3.46 4.78 6.88 10.38d ¼ 2f 1.65 2.19 3.01 4.29 6.42 10.20 17.50

* For vertical wall face (b ¼ 0) and horizontal ground surface (a ¼ 0).

Geotechnical Engineering n 7.91

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The first line of Table 7.16 lists values obtainedfrom Eq. (7.77).

Continuous anchors in sand (f 5 3388888), whensubjected to horizontal pull or thrust, develop pas-sive pressures, lb/lin ft, of about

P ¼ 1:5wh2 (7:78)

where h ¼ distance from bottom of anchor to thesurface, ft.

This relationship holds for ratios of h to height d, ft,of anchor of 1.5 to 5.5, and assumes a horizontalground surface and vertical anchor face.

Square anchors within the same range of h/ddevelop about

P ¼ 2:50þ h

8d

� �2

dwh2

2(7:79)

where P ¼ passive lateral pressure, lb

d ¼ length and height of anchor, ft

Passive pressures of cohesive soils, resistingmovement of a wall or anchor extending to theground surface, depend on the unit weight of thesoil w and its unconfined compressive strength 2c,psf. At a distance h, ft, below the surface, thepassive lateral pressure, psf, is

p ¼ whþ 2c (7:80)

The total pressure, lb/lin ft, is

P ¼ wh2

2þ 2ch (7:81)

and acts at a distance, ft, above the bottom of thewall or anchor of

�xx ¼ h(whþ 6c)

3(whþ 4c)

The pressure distribution for plastic clay is shownin Fig. 7.48c.

Continuous anchors in plastic clay, when sub-jected to horizontal pull or thrust, develop passivepressures, lb/lin ft, of about

P ¼ cd 8:7� 11,600

(h=dþ 11)3

� �(7:82)

where h ¼ distance from bottom of anchor tosurface, ft

d ¼ height of anchor, ft

Equation (7.82) is based on tests made with hori-zontal ground surface and vertical anchor face.

Safety factors should be applied to the passivepressures computed from Eqs. (7.74) to (7.82) fordesign use. Experience indicates that a safety factorof 2 is satisfactory for clean sands and gravels. Forclay, a safety factor of 3 may be desirable because ofuncertainties as to effective shearing strength.

(G. P. Tschebotarioff, “Soil Mechanics, Founda-tions, and Earth Structures,” McGraw-Hill BookCompany, New York (books.mcgraw-hill.com);K. Terzaghi and R. B. Peck, “Soil Mechanics inEngineering Practice,” John Wiley & Sons, Inc.,New York (www.wiley.com); Leo Casagrande,“Comments on Conventional Design of RetainingStructures,” ASCE Journal of Soil Mechanics andFoundations Engineering Division, 1973, pp. 181–198;H. Y. Fang, “Foundation Engineering Handbook,”2nd ed., Van Nostrand Reinhold Company, NewYork.)

7.29 Vertical Earth Pressureon Conduit

The vertical load on an underground conduitdepends principally on the weight of the prism ofsoil directly above it. But the load also is affected byvertical shearing forces along the sides of thisprism. Caused by differential settlement of theprism and adjoining soil, the shearing forces maybe directed up or down. Hence, the load on theconduit may be less or greater than the weight ofthe soil prism directly above it.

Conduits are classified as ditch or projecting,depending on installation conditions that affect theshears. A ditch conduit is a pipe set in a relativelynarrow trench dug in undisturbed soil (Fig. 7.49).Backfill then is placed in the trench up to theoriginal ground surface. A projecting conduit is apipe over which an embankment is placed.

A projecting conduit may be positive ornegative, depending on the extent of the embank-ment vertically. A positive projecting conduit isinstalled in a shallow bed with the pipe top abovethe surface of the ground. Then, the embankmentis placed over the pipe (Fig. 7.50a). A negative pro-jecting conduit is set in a narrow, shallow trenchwith the pipe top below the original ground surface(Fig. 7.50b). Then, the ditch is backfilled, afterwhich the embankment is placed. The load on theconduit is less when the backfill is not compacted.

7.92 n Section Seven

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Load on underground pipe also may be reducedby the imperfect-ditchmethod of construction. Thisstarts out as for a positive projecting conduit, withthe pipe at the original ground surface. Theembankment is placed and compacted for a fewfeet above the pipe. But then, a trench as wide asthe conduit is dug down to it through the com-pacted soil. The trench is backfilled with a loose,compressible soil (Fig. 7.50c). After that, the em-bankment is completed.

The load, lb/lin ft, on a rigid ditch conduit maybe computed from

W ¼ CDwhb (7:83)

and on a flexible ditch conduit from

W ¼ CDwhD (7:84)

where CD ¼ load coefficient for ditch conduit

w ¼ unit weight of fill, lb/ft3

h ¼ height of fill above top of conduit, ft

b ¼ width of ditch at top of conduit, ft

D ¼ outside diameter of conduit, ft

From the equilibrium of vertical forces, includingshears, acting on the backfill above the conduit, CD

may be determined:

CD ¼ 1� e�kh=b

k

b

h(7:85)

where e ¼ 2.718

k ¼ 2Katan u

Ka ¼ coefficient of active earth pressure [Eq.(7.64) and Table 7.14]

u ¼ angle of friction between fill and adja-cent soil (u � f, angle of internal frictionof fill)

Table 7.17 gives values of CD for k ¼ 0.33 for cohe-sionless soils, k ¼ 0.30 for saturated topsoil, andk ¼ 0.26 and 0.22 for clay (usual maximum andsaturated).

Fig. 7.50 Type of projecting conduit depends on method of backfilling.

Fig. 7.49 Ditch conduit.

Geotechnical Engineering n 7.93

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Vertical load, lb/lin ft, on conduit installed bytunneling may be estimated from

W ¼ CDb(wh� 2c) (7:86)

where c ¼ cohesion of the soil, or half the uncon-fined compressive strength of the soil, psf. The loadcoefficient CD may be computed from Eq. (7.85) orobtained from Table 7.17 with b ¼ maximumwidthof tunnel excavation, ft, and h ¼ distance fromtunnel top to ground surface, ft.

For a ditch conduit, shearing forces extend fromthe pipe top to the ground surface. For a projectingconduit, however, if the embankment is sufficientlyhigh, the shear may become zero at a horizontalplane below grade, the plane of equal settlement.Load on a projecting conduit is affected by thelocation of this plane.

Vertical load, lb/lin ft, on a positive projectingconduit may be computed from

W ¼ CPwhD (7:87)

where CP ¼ load coefficient for positive projectingconduit. Formulas have been derived for CP andthe depth of the plane of equal settlement. Theseformulas, however, are too lengthy for practicalapplication, and the computation does not appearto be justified by the uncertainties in actual relativesettlement of the soil above the conduit. Tests maybe made in the field to determine CP. If so, thepossibility of an increase in earth pressure withtime should be considered. For a rough estimate,

CP may be assumed as 1 for flexible conduit and 1.5for rigid conduit.

The vertical load, lb/lin ft, on negative project-ing conduit may be computed from

W ¼ CNwhb (7:88)

where CN ¼ load coefficient for negative projectingconduit

h ¼ height of fill above top of conduit, ft

b ¼ horizontal width of trench at top ofconduit, ft

The load on an imperfect ditch conduit may beobtained from

W ¼ CNwhD (7:89)

where D ¼ outside diameter of conduit, ft.Formulas have been derived for CN, but they are

complex, and insufficient values are available forthe parameters involved. As a rough guide, CN

may be taken as 0.9 when depth of cover exceedsconduit diameter. (See also Art. 10.31.)

Superimposed surface loads increase the loadon an underground conduit. The magnitude of theincrease depends on the depth of the pipe belowgrade and the type of soil. For moving loads, animpact factor of about 2 should be applied. Asuperimposed uniform load w0, lb/ft2, of largeextent may be treated for projecting conduit as anequivalent layer of embankment with a thickness,ft, of w0=w. For ditch conduit, the load due to the

Table 7.17 Load Coefficients CD for Ditch Conduit

Clay

h/b Cohesionless Soils Saturated Topsoil k ¼ 0.26 k ¼ 0.22

1 0.85 0.86 0.88 0.892 0.75 0.75 0.78 0.803 0.63 0.67 0.69 0.734 0.55 0.58 0.62 0.675 0.50 0.52 0.56 0.606 0.44 0.47 0.51 0.557 0.39 0.42 0.46 0.518 0.35 0.38 0.42 0.479 0.32 0.34 0.39 0.4310 0.30 0.32 0.36 0.4011 0.27 0.29 0.33 0.3712 0.25 0.27 0.31 0.35

Over 12 3.0b/h 3.3b/h 3.9b/h 4.5b/h

7.94 n Section Seven

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soil should be increased by bw0e2kh/b, wherek ¼ 2Katan u, as in Eq. (7.85). The increase due toconcentrated loads can be estimated by assumingthe loads to spread out linearly with depth, at anangle of about 308 with the vertical. (See alsoArt. 7.11).

(M. G. Spangler, “Soil Engineering,” Interna-tional Textbook Company, Scranton, Pa.; “Hand-book of Steel Drainage and Highway ConstructionProducts,” American Iron and Steel Institute,Washington, D. C. (www.steel.org).)

7.30 Dewatering Methodsfor Excavations

The main purpose of dewatering is to enableconstruction to be carried out under relatively dryconditions. Good drainage stabilizes excavatedslopes, reduces lateral loads on sheeting andbracing, and reduces required air pressure intunneling. Dewatering makes excavated materiallighter and easier to handle. It also prevents loss ofsoil below slopes or from the bottom of theexcavation, and prevents a “quick” or “boiling”bottom. In addition, permanent lowering of thegroundwater table or relief of artesian pressuremay allow a less expensive design for the structure,especially when the soil consolidates or becomescompact. If lowering of the water level or pressurerelief is temporary, however, the improvementof the soil should not be considered in founda-tion design. Increases in strength and bearing capa-city may be lost when the soil again becomessaturated.

To keep an excavation reasonably dry, thegroundwater table should be kept at least 2 ft, andpreferably 5 ft, below the bottom in most soils.

Site investigations should yield informationuseful for deciding on the most suitable and eco-nomical dewatering method. Important is a knowl-edge of the types of soil in and below the site,probable groundwater levels during construction,permeability of the soils, and quantities of water tobe handled. A pumping test may be desirable forestimating capacity of pumps needed and drainagecharacteristics of the ground.

Many methods have been used for dewateringexcavations. Those used most often are listed inTable 7.18 with conditions for which they generallyare most suitable. (See also Art. 7.37.)

In many small excavations, or where there aredense or cemented soils, water may be collected inditches or sumps at the bottom and pumped out.This is the most economical method of dewatering,and the sumps do not interfere with futureconstruction as does a comprehensive wellpointsystem. But the seepage may slough the slopes,unless they are stabilized with gravel, and mayhold up excavation while the soil drains. Also,springs may develop in fine sand or silt and causeunderground erosion and subsidence of theground surface.

For sheetpile-enclosed excavations in pervioussoils, it is advisable to intercept water before itenters the enclosure. Otherwise, the water willput high pressures on the sheeting. Seepagealso can cause the excavation bottom to becomequick, overloading the bottom bracing, or createpiping, undermining the sheeting. Furthermore,pumping from the inside of the cofferdam islikely to leave the soil to be excavated wet andtough to handle.

Wellpoints often are used for lowering the watertable in pervious soils. They are not suitable,however, in soils that are so fine that they will flowwith the water or in soils with low permeability.Also, other methods may be more economical fordeep excavations, very heavy flows, or consider-able lowering of the water table (Table 7.18).

Wellpoints are metal well screens about 2 to 3 inin diameter and up to about 4 ft long. A pipeconnects each wellpoint to a header, from whichwater is pumped to discharge (Fig. 7.51). Eachpump usually is a combined vacuum and centri-fugal pump. Spacing of wellpoints generally rangesfrom 3 to 12 ft c to c.

Awellpoint may be jetted into position or set ina hole made with a hole puncher or heavy steelcasing. Accordingly, wellpoints may be self-jettingor plain-tip. To insure good drainage in fine anddirty sands or in layers of silt or clay, the wellpointand riser should be surrounded by sand to justbelow the water table. The space above the filtershould be sealed with silt or clay to keep air fromgetting into the wellpoint through the filter.

Wellpoints generally are relied on to lowerthe water table 15 to 20 ft. Deep excavations may bedewatered with multistage wellpoints, with onerow of wellpoints for every 15 ft of depth. Or whenthe flow is less than about 15 gal/min per well-point, a single-stage system of wellpoints may beinstalled above the water table and operated with

Geotechnical Engineering n 7.95

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jet-eductor pumps atop the wellpoints. Thesepumps can lower the water table up to about 100 ft,but they have an efficiency of only about 30%.

Deep wells may be used in pervious soils fordeep excavations, large lowering of the water table,and heavy water flows. They may be placed alongthe top of an excavation to drain it, to interceptseepage before pressure makes slopes unstable,and to relieve artesian pressure before it heaves theexcavation bottom.

Usual spacing of wells ranges from 20 to 250 ft.Diameter generally ranges from 6 to 20 in. Wellscreens may be 20 to 75 ft long, and they are sur-rounded with a sand-gravel filter. Generally, pump-ing is done with a submersible or vertical turbinepump installed near the bottom of each well.

Figure 7.52 shows a deep-well installation usedfor a 300-ft-wide by 600-ft-long excavation for

a building of the Smithsonian Institution, Washing-ton, D. C. Two deep-well pumps lowered thegeneral water level in the excavation 20 ft. The wellinstallation proceeded as follows: (1) Excavation towater level (elevation 0.0). (2) Driving of sheetpilesaround the well area (Fig. 7.52a). (3) Excavationunderwater inside the sheetpile enclosure to ele-vation 237.0 ft (Fig. 7.52b). Bracing installed asdigging progressed. (4) Installation of a wire-mesh-wrapped timber frame, extending from elevation0.0 to 237.0 (Fig. 7.52c). Weights added to sink theframe. (5) Backfilling of space between sheetpilesand mesh with 3⁄16 - to

3⁄8 -in gravel. (6) Removal ofsheetpiles. (7) Installation of pump and start ofpumping.

Vacuum well or wellpoint systems may be usedto drain silts with low permeability (coefficientbetween 0.01 and 0.0001 mm/s). In these systems,

Table 7.18 Methods for Dewatering Excavations

Saturated-Soil Conditions Dewatering Method Probably Suitable

Surface water Ditches; dikes; sheetpiles and pumps orunderwater excavation and concrete tremie seal

Gravel Underwater excavation, grout curtain; gravitydrainage with large sumps with gravel filters

Sand (except very fine sand) Gravity drainageWaterbearing strata near surface; water table

does not have to be lowered more than 15ftWellpoints with vacuum and centrifugal pumps

Waterbearing strata near surface; water table tobe lowered more than 15ft, low pumpingrate

Wellpoints with jet-eductor pumps

Excavations 30ft or more below water table;artesian pressure; high pumping rate; largelowering of water table—all where adequatedepth of pervious soil is available forsubmergence of well screen and pump

Deep wells, plus, if necessary, wellpoints

Sand underlain by rock near excavation bottom Wellpoints to rock, plus ditches, drains, automatic“mops”

Sand underlain by clay Wellpoints in holes 3 or 4 into the clay, backfilledwith sand

Silt; very find sand (permeability coefficientbetween 0.01 and 0.0001mm/s)

For lifts up to 15ft, wellpoints with vacuum; forgreater lifts, wells with vacuum; sumps

Silt or silty sand underlain by pervious soil At top of excavation, and extending to thepervious soil, vertical sand drains plus wellpoints or wells

Clay-silts, silts Electro-osmosisClay underlain by pervious soil At top of excavation, wellpoints or deep wells

extending to pervious soilDense or cemented soils: small excavations Ditches and sumps

7.96 n Section Seven

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wells or wellpoints are closely spaced, and avacuum is held with vacuum pumps in the wellscreens and sand filters. At the top, the filter, well,and risers should be sealed to a depth of 5 ft withbentonite or an impervious soil to prevent loss ofthe vacuum. Water drawn to the well screens ispumped out with submersible or centrifugalpumps.

Where a pervious soil underlies silts or siltysands, vertical sand drains and deep wells canteam up to dewater an excavation. Installed at thetop, and extending to the pervious soil, the sandpiles intercept seepage and allow it to draindown to the pervious soil. Pumping from thedeep wells relieves the pressure in that deep soillayer.

For some silts and clay-silts, electrical drainagewith wells or wellpoints may work, whereasgravity methods may not (Art. 7.37). In saturated

clays, thermal or chemical stabilization may benecessary (Arts. 7.38 and 7.39).

Small amounts of surface water may be re-moved from excavations with “mops.” Surroundedwith gravel to prevent clogging, these drains areconnected to a header with suction hose or pipe.For automatic operation, each mop should beopened and closed by a float and float valve.

When structures on silt or soft material arelocated near an excavation to be dewatered, careshould be taken that lowering of the water tabledoes not cause them to settle. It may be necessary tounderpin the structures or to pump dischargewater into recharge wells near the structures tomaintain the water table around them.

(L. Zeevaert, “Foundation Engineering for Diffi-cult Subsoil Conditions,” H. Y. Fang, “FoundationEngineering Handbook,” 2nd ed., Van NostrandReinhold Company, New York.)

Fig. 7.51 Wellpoint system for dewatering an excavation.

Geotechnical Engineering n 7.97

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Underpinning

The general methods and main materials used togive additional support at or below grade to struc-tures are called underpinning. Usually, the addedsupport is applied at or near the footings.

7.31 UnderpinningProcedures

Underpinning may be remedial or precautionary.Remedial underpinning adds foundation capacityto an inadequately supported structure. Precau-tionary underpinning is provided to obtain ade-quate foundation capacity to sustain higher loads,as a safeguard against possible settlement duringadjacent excavating, or to accommodate changes inground conditions. Usually, this type of under-pinning is required for the foundations of a struc-ture when deeper foundations are to be con-structed nearby for an addition or anotherstructure. Loss of ground, even though small, into

an adjoining excavation may cause excessivesettlement of existing foundations.

Presumably, an excavation influences an exist-ing substructure when a plane through the outer-most foundations, on a 1-on-1 slope for sand or a1-on-2 slope for unconsolidated silt or soft clay,penetrates the excavation. For a cohesionless soil,underpinning exterior walls within a 1-on-1 slopeusually suffices; interior columns are not likely tobe affected if farther from the edge of the exca-vation than half the depth of the cut.

The commonly accepted procedures of struc-tural and foundation design should be used forunderpinning. Data for computing dead loads maybe obtained from plans of the structure or a fieldsurvey. Since underpinning is applied to existingstructures, some of which may be old, engineers incharge of underpinning design and constructionshould be familiar with older types of constructionas well as the most modern.

Before underpinning starts, the engineersshould investigate and record existing defects inthe structure. Preferably, the engineers should be

Fig. 7.52 Deep-well installation used at Smithsonian Institution, Washington, D.C. (Spencer, White &Prentis, Inc.)

7.98 n Section Seven

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accompanied in this investigation by a represen-tative of the owner. The structure should bethoroughly inspected, from top to bottom, inside(if possible) and out. The report should includenames of inspectors, dates of inspection, anddescription and location of defects. Photographsare useful in verifying written descriptions ofdamaged areas. The engineers shouldmark existingcracks in such a way that future observationswould indicate whether they are continuing toopen or spread.

Underpinning generally is accompanied bysome settlement. If design and field work aregood, the settlement may be limited to about 1⁄4 to3⁄8 in. But as long as settlement is uniform in astructure, damage is unlikely. Differential settle-ment should be avoided. To check on settlement,elevations of critical points, especially columnsand walls, should be measured frequently duringunderpinning. Since movement may also occurlaterally, the plumbness of walls and columns alsoshould be checked.

One of the first steps in underpinning usually isdigging under a foundation, which decreases itsload-carrying capacity temporarily. Hence, pre-liminary support may be necessary until under-pinning is installed. This support may be providedby shores, needles, grillages, and piles. Sometimes,it is desirable to leave them in place as permanentsupports.

Generally, it is advisable to keep preliminarysupports at a minimum, for economy and to avoidinterference with other operations. For the pur-pose, advantage may be taken of arching actionand of the ability of a structure to withstandmoderate overloads. Also, columns centrally sup-ported on large spread footings need not be shoredwhen digging is along an edge and involves only asmall percent of the total footing area. A large partof the column load is supported by the soil directlyunder the column.

When necessary, weak portions of a structure,especially masonry, should be repaired or strength-ened before underpinning starts.

7.32 Shores

Installed vertically or on a slight incline, shores areused to support walls or piers while underpinningpits are dug (Fig. 7.53a). Good bearing should beprovided at top and bottom of the shores. One way

of providing bearing at the top is to cut a niche andmortar a steel bearing plate against the upper face.An alternate to the plate is a Z shape, made byremoving diagonally opposite half flanges from anH beam. When the top of the shore is cut to fitbetween flange and web of the Z, movement of theshore is restrained. For a weak masonry wall, theload may have to be distributed over a larger area.One way of doing this is to insert a few lintel anglesabout 12 in apart vertically and bolt them to avertical, heavy timber or steel distributing beam.The horizontal leg of an angle on the beam thentransmits the load to a shore.

Inclined shores on only one side of a wall requiresupport at the base for horizontal as well as verticalforces. One way is to brace the shores against anopposite wall at the floor. Preferably, the base of eachshore should sit on a footing perpendicular to theaxis of the shore. Sized to provide sufficient bearingon the soil, the footing may be made of heavy tim-bers, steel beams, or reinforced concrete, dependingon the load on the shore.

Loads may be transferred to a shore by wedgesor jacks. Oak wedges are suitable for light loads;forged steel wedges and bearing plates are desir-able for heavy loads. Jacks, however, offer greaterflexibility in length adjustments and allow correc-tions during underpinning for settlement of shorefootings.

(H. A. Prentis and L. White, “Underpinning,”Columbia University Press, New York; M. J. Tom-linson, “Foundation Design and Construction,”Halsted Press, New York.)

7.33 Needles and Grillages

Needles are beams installed horizontally totransfer the load of a wall or column to either orboth sides of its foundation, to permit digging ofunderpinning pits (Fig. 7.53b). These beams aremore expensive than shores, which transmit theload directly into the ground. Needles usually aresteel wide-flange beams, sometimes plate girders,used in pairs, with bolts and pipe spreadersbetween the beams. This arrangement providesresistance to lateral buckling and torsion. Theneedles may be prestressed with jacks to eliminatesettlement when the load is applied.

The load from steel columns may be transmittedthrough brackets to the needles. For masonry walls,the needles may be inserted through niches. The

Geotechnical Engineering n 7.99

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load should be transferred from the masonry to theneedles through thin wood fillers that crush whenthe needles deflect and maintain nearly uniformbearing.

Wedges may be placed under the ends of theneedles to shift the load from the member to besupported to those beams. The beam ends may becarried on timber pads, which distribute the loadover the soil.

Grillages, which have considerably more bear-ing on the ground than needles, often are used asan alternative to needles and shores for closelyspaced columns. A grillage may be installedhorizontally on soil at foundation level to supportand tie together two or more column footings, or itmay rest on a cellar floor (Fig. 7.53c). Thesepreliminary supports may consist of two or moresteel beams, tied together with bolts and pipespreaders, or of a steel-concrete composite. Also,grillages sometimes are used to strengthen or

repair existing footings by reinforcing them andincreasing their bearing area. The grillages maytake the form of dowels or of encircling concrete orsteel-concrete beams. They should be adequatelycross-braced against buckling and torsion. Holesshould be made in steel beams to be embedded inconcrete, to improve bond.

7.34 Pit Underpinning

After preliminary supports have been installed andweak construction strengthened or repaired, under-pinning may start. The most common method ofunderpinning a foundation is to construct concretepiers down to deeper levels with adequate bearingcapacity and to transfer the load to the piers bywedging up with dry packing. To build the piers,pits must be dug under the foundation. Because ofthe danger of loss of ground and consequent settle-

Fig. 7.53 Temporary supports used in underpinning: (a) shores; (b) needle beams; (c) grillage.

7.100 n Section Seven

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ment where soils are saturated, the method usuallyis restricted to dry subsoil.

When piers have to be placed close together, acontinuous wall may be constructed instead. Butthe underpinning wall should be built in shortsections, usually about 5 ft long, to avoid under-mining the existing foundation. Alternate sectionsare built first, and then the gaps are filled in.

Underpinning pits rarely are larger than about5 ft square in cross section. Minimum size foradequate working room is 3 � 4 ft. Access to the pit

is provided by an approach pit started alongsidethe foundation and extending down about 6 ft. Thepits must be carefully sheeted and adequatelybraced to prevent loss of ground, which can causesettlement of the structure.

In soils other than soft clay, 2-in-thick woodplanks installed horizontally may be used to sheetpits up to 5 ft square, regardless of depth. Sides ofthe pits should be trimmed back no more thanabsolutely necessary. The boards, usually 2� 80s, areinstalled one at a time with 2-in spaces between

Fig. 7.54 Vertical section through White House, Washington, D.C., during restoration. Pit underpinningwas used for the walls. (Spencer, White & Prentis, Inc.)

Geotechnical Engineering n 7.101

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them vertically. Soil is repacked through these lou-vers to fill the voids behind the boards. In runningsands, haymay be stuffed behind the boards to blockthe flow. Corners of the sheeting often are nailedwith vertical 2 � 4 in wood cleats.

In soft clay, sheeting must be tight and bracedagainst earth pressure. Chicago caissons, or a simi-lar type, may be used (Art. 7.23).

In water-bearing soil with a depth not exceed-ing about 5 ft, vertical sheeting can sometimes bedriven to cut off the water. For the purpose, lightsteel or tongue-and-groove wood sheeting may beused. The sheeting should be driven below thebottom of the pit a sufficient distance to preventboiling of the bottom due to hydrostatic pressure.With water cut off, the pit can be pumped dry andexcavation continued.

After a pit has been dug to the desired level, it isfilled with concrete to within 3 in of the foundationto be supported. The gap is dry-packed, usually byramming stiff mortar in with a 2 � 4 pounded byan 8-lb hammer. The completed piers should belaterally braced if soil is excavated on one side to adepth of more than about 6 ft. An example of pitunderpinning is the work done in the restoration ofthe White House, in which a cellar and subcellarwere created (Fig. 7.54).

7.35 Pile Underpinning

If water-bearing soil more than about 5 ft deepunderlies a foundation, the structure may have tobe underpinned with piles. Driven piles generallyare preferred to jacked piles because of lower cost.The feasibility of driven piles, however, dependson availability of at least 12 ft of headroom andspace alongside the foundations. Thus, driven pilesoften can be used to underpin interior buildingcolumns when headroom is available. But they arehard to install for exterior walls unless there isample space alongside the walls. For very lightlyloaded structures, brackets may be attached tounderpinning piles to support the structure. Butsuch construction puts bending into the piles,reducing their load-carrying capacity.

Driven piles usually are 12- to 14-in diametersteel pipe, 3⁄8 in thick. They are driven open-ended,to reduce vibration, and in lengths determined byavailable headroom. Joints may be made with cast-steel sleeves. After soil has been removed from thepipe interior, it is filled with concrete.

Jacked piles require less headroom and may beplaced under a footing. Also made of steel pipeinstalled open-ended, these piles are forced downby hydraulic jacks reacting against the footing. Theoperation requires an approach pit under thefooting to obtain about 6 ft of headroom.

Pretest piles, originally patented by Spencer,White & Prentis, New York City, are used to prev-ent the rebound of piles when jacking stops andsubsequent settlement when the load of the struc-ture is transferred to the piles. A pipe pile is jackeddown, in 4-ft lengths, to the desired depth. Thehydraulic jack reacts against a steel platemortared tothe underside of the footing to be supported. Afterthe pile has been driven to the required depth andcleaned out, it is filled with concrete and cappedwith a steel bearing plate. Then two hydraulic jacksatop the pile overload it 50%. As the load is applied,a bulb of pressure builds up in the soil at the pilebottom. This pressure stops downward movementof the pile. While the jacks maintain the load, a shortlength of beam is wedged between the pile top andthe steel plate under the footing. Then, the jacks areunloaded and removed. The load, thus, is trans-ferred without further settlement. Later the spaceunder the footing is concreted. Figure 7.55 showshow pretest piles were used for underpinningexisting structures during construction of a subwayin New York City.

7.36 MiscellaneousUnderpinning Methods

Spread footings may be pretested in much the sameway as piles. The weight of the structure is used tojack down the footings, which then are wedged inplace, and the gap is concreted. Themethodmay beresorted to for unconsolidated soils where a highwater table makes digging under a footing unsafeor where a firm stratum is deep down.

A form of underpinning may be used for slabson ground. When concrete slabs settle, they maybe restored to the proper elevation by mud jacking.In this method, which will not prevent futuresettlement, a fluid grout is pumped under the slabthrough holes in it, raising it. Pressure is main-tained until the grout sets. The method also may beused to fill voids under a slab.

Chemical or thermal stabilization (Arts. 7.38 and7.39) sometimes may be used as underpinning.

7.102 n Section Seven

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Ground Improvement

Soil for foundations can be altered to conform todesired characteristics. Whether this should bedone depends on the cost of alternatives.

Investigations of soil and groundwater con-ditions on a site should indicate whether soil im-provement, or stabilization, is needed. Tests may benecessary to determine which of several applicabletechniques may be feasible and economical. Table7.19 lists some conditions for which soil improve-ment should be considered and the methods thatmay be used.

As indicated in the table, ground improvementmay increase strength, increase or decrease per-meability, reduce compressibility, improve stability,or decrease heave due to frost or swelling. The maintechniques used are constructed fills, replacement of

unsuitable soils, surcharges, reinforcement, mech-anical stabilization, thermal stabilization, and chemi-cal stabilization. (Federal Highway Administration“Ground Improvement Technical Summaries” Pub-lication No. FHWA-SA-98-086, December 1999).

7.37 MechanicalStabilization of Soils

This comprises a variety of techniques forrearranging, adding, or removing soil particles.The objective usually is to increase soil density,decrease water content, or improve gradation.Particles may be rearranged by blending the layersof a stratified soil, remolding an undisturbed soil,or densifying a soil. Sometimes, the desired im-provement can be obtained by drainage alone.

Fig. 7.55 Pretest piles support existing buildings, old elevated railway, and a tunnel during con-struction of a subway. (Spencer, White & Prentis, Inc.)

Geotechnical Engineering n 7.103

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Often, however, compactive effort plus watercontrol is needed.

7.37.1 Embankments

Earth often has to be placed over the existingground surface to level or raise it. Such constructedfills may create undesirable conditions because ofimproper compaction, volume changes, and unex-pected settlement under the weight of the fill. Toprevent such conditions, fill materials and their

gradation, placement, degree of compaction, andthickness should be suitable for properly support-ing the expected loads.

Fills may be either placed dry with conventionalearthmoving equipment and techniques or wet byhydraulic dredges. Wet fills are used mainly forfilling behind bulkheads or for large fills.

A variety of soils and grain sizes are suitablefor topping fills for most purposes. Inclusion oforganic matter or refuse should, however, be pro-hibited. Economics usually require that the source

Table 7.19 Where Soil Improvement May Be Economical

SoilDeficiency

Probable Typeof Failure

ProbableCause

PossibleRemedies

Slope instability Slides on slope Pore-waterpressure

Drain; flatten slope; freeze

Loose granularsoil

Compact

Weak soil Mix or replace with select materialMud flow Excessive water Exclude waterSlides—movementat toe

Toe instability Place toe fill, and drain

Low bearingcapacity

Excessive settle-ment

Saturated clay Consolidate with surcharge, and drain

Loose granularsoil

Compact; drain; increase footing depth;mix with chemicals

Weak soil Superimpose thick fill; mix or replace withselect material; inject or mix withchemicals; freeze (if saturated); fuse withheat (if unsaturated)

Heave Excessive rise Frost For buildings: place foundations below frostline; insulate refrigeration-room floors;refrigerate to keep ground frozen

For roads: Remove fines from gravel; replacewith nonsusceptible soil

Expansion of clay Exclude water; replace with granular soil

ExcessivePermeability

Seepage Pervious soil orfissured rock

Mix or replace soil with select material;inject or mix soil with chemicals;construct cutoff wall with grout; enclosewith sheetpiles and drain

“Quick” bottom Loss of strength Flow under Add berm against cofferdam inner face;increase width of cofferdam betweenlines of sheeting; drain with wellpointsoutside the cofferdam

7.104 n Section Seven

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of fill material be as close as possible to the site. Formost fills, soil particles in the 18 in belowfoundations, slabs, or the ground surface shouldnot be larger than 3 in in any dimension.

For determining the suitability of a soil as filland for providing a standard for compaction, themoisture-density relationship test, or Proctor test(ASTM D698 and D1557), is often used. Several ofthese laboratory tests should be performed on theborrowed material, to establish moisture-densitycurves. The peak of a curve indicates the maximumdensity achievable in the laboratory by the testmethod as well as the optimum moisture content.ASTM D1557 should be used as the standard whenhigh bearing capacity and low compressibility arerequired; ASTM D698 should be used when re-quirements are lower, for example, for fills underparking lots.

The two ASTM tests represent different levels ofcompactive effort. But much higher compactiveeffort may be employed in the field than that usedin the laboratory. Thus, a differentmoisture-densityrelationship may be produced on the site. Proctortest results, therefore, should not be considered aninherent property of the soil. Nevertheless, the testresults indicate the proposed fill material’s sensi-tivity to moisture content and the degree of fieldcontrol that may be required to obtain the specifieddensity.

See also Art. 7.40.

7.37.2 Fill Compaction

The degree of compaction required for a fill isusually specified as a minimum percentage of themaximum dry density obtained in the laboratorytests. This compaction is required to be accom-plished within a specific moisture range. Minimumdensities of 90 to 95% of the maximum density aresuitable for most fills. Under roadways, footings, orother highly loaded areas, however, 100% compac-tion is often required. In addition, moisture contentwithin 2 to 4% of the optimum moisture contentusually is specified.

Field densities can be greater than 100% of themaximum density obtained in the laboratory test.Also, with greater compactive effort, such densitiescan be achieved with moisture contents that do notlie on the curves plotted from laboratory results.(Fine-grained soils should not be overcompactedon the dry side of optimum because when they getwet, they may swell and soften significantly.)

For most projects, lift thickness should berestricted to 8 to 12 in, each lift being compactedbefore the next lift is placed. On large projectswhere heavy compaction equipment is used, a liftthickness of 18 to 24 in is appropriate.

Compaction achieved in the field should bedetermined by performing field density tests oneach lift. For that purpose, wet density andmoisture content should be measured and the drydensity computed. Field densities may be ascer-tained by the sand-cone (ASTM D1556) or balloonvolume-meter (ASTM D2167) method, from anundisturbed sample, or with a nuclear moisture-density meter. Generally, one field density test foreach 4000 to 10,000 ft2 of lift surface is adequate.

Hydraulically placed fills composed of dredgedsoils normally need not be compacted duringplacement. Although segregation of the silt andclay fractions of the soils may occur, it usually is notharmful. But accumulation of the fine-grainedmaterial in pockets at bulkheads or under struc-tures should be prevented. For the purpose, inter-nal dikes, weirs, or decanting techniques may beused.

7.37.3 Soil Replacement or Blending

When materials at or near grade are unsuitable, itmay be economical to remove them and substitutea fill of suitable soil, as described in Art. 7.37.1.When this is not economical, consideration shouldbe given to improving the soil by other methods,such as densification or addition or removal of soilparticles.

Mixing an existing soil with select materials orremoving selected sizes of particles from anexisting soil can change its properties considerably.Adding clay to a cohesionless soil in a nonfrostregion, for example, maymake the soil suitable as abase course for a road (if drainage is not too greatlyimpaired). Adding clay to a pervious soil mayreduce its permeability sufficiently to permit its useas a reservoir bottom. Washing particles finer than0.02 mm from gravel makes the soil less susceptibleto frost heave (desirable upper limit for thisfraction is 3%).

7.37.4 Surcharges

Where good soils are underlain by soft, compres-sible clays that would permit unacceptable settle-ment, the site often can be made usable by

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surcharging, or preloading, the surface. Theobjective is to use the weight of the surcharge toconsolidate the underlying clays. This offsets thesettlement of the completed structure that wouldotherwise occur. A concurrent objective may be toincrease the strength of the underlying clays.

If the soft clay is overlain by soils with adequatebearing capacity, the area to be improved may beloaded with loose, dumped earth, until the weightof the surcharge is equivalent to the load that laterwill be imposed by the completed structure. (Ifhighly plastic clays or thick layers with littleinternal drainage are present, it may be necessaryto insert vertical drains to achieve consolidationwithin a reasonable time.) During and afterplacement of the surcharge, settlement of theoriginal ground surface and the clay layer shouldbe closely monitored. The surcharge may beremoved after little or no settlement is observed.If surcharging has been properly executed, thecompleted structure should experience no furthersettlement due to primary consolidation. Potentialsettlement due to secondary compression, how-ever, should be evaluated, especially if the soft soilsare highly organic.

7.37.5 Densification

Any of a variety of techniques, most involvingsome form of vibration, may be used for soildensification. The density achieved with a specifictechnique, however, depends on the grain size ofthe soil. Consequently, grain-size distribution mustbe taken into account when selecting a densifica-tion method.

Compaction of clean sands to depths of about6 ft usually can be achieved by rolling the surfacewith a heavy, vibratory, steel-drum roller. Al-though the vibration frequency is to some extentadjustable, the frequencies most effective are inthe range of 25 to 30 Hz. Bear in mind, however,that little densification will be achieved below adepth of 6 ft, and the soil within about 1 ft of thesurface may actually be loosened. Compactiveeffort in the field may be measured by the num-ber of passes made with a specific machine ofgiven weight and at a given speed. For a givencompactive effort, density varies with moisturecontent. For a given moisture content, increasingthe compactive effort increases the soil densityand reduces the permeability.

Compaction piles also may be used to densifysands. For that purpose, the piles usually are madeof wood. Densification of the surrounding soilsresults from soil displacement during driving ofthe pile or shell and from the vibration producedduring pile driving. The foundations to be con-structed need not bear directly on the compactionpiles but may be seated anywhere on the densifiedmass.

Vibroflotation and Terra-Probe are alternativemethods that increase sand density by multipleinsertions of vibrating probes. These form cylind-rical voids, which are then filled with offsite sand,stone, or blast-furnace slag. The probes usually areinserted in clusters, with typical spacing of about41⁄2 ft, where footings will be placed. Relative densi-ties of 85% or more can be achieved throughout thedepth of insertion, which may exceed 40 ft. Use ofvibrating probes may not be effective, however, ifthe fines content of the soil exceeds about 15% orif organic matter is present in colloidal form inquantities exceeding about 5% by weight.

Another technique for densification is dynamiccompaction, which in effect subjects the site tonumerous mini-earthquakes. In saturated soils,densification by this method also results frompartial liquefaction, and the elevated pore press-ures produced must be dissipated between eachapplication of compactive energy if the followingapplication is to be effective. As developed byTechniques Louis Menard, dynamic compaction isachieved by dropping weights ranging from 10 to40 tons from heights up to 100 ft onto the groundsurface. Impact spacings range up to 60 ft. Multipledrops are made at each location to be densified.This technique is applicable to densification oflarge areas and a wide range of grain sizes andmaterials.

7.37.6 Drainage

This is effective in soil stabilization becausestrength of a soil generally decreases with anincrease in amount and pressure of pore water.Drainage may be accomplished by gravity, pum-ping, compression with an external load on the soil,electro-osmosis, heating, or freezing.

Pumping often is used for draining the bottomof excavations (Art. 7.30). For slopes, however,advantage must be taken of gravity flow toattain permanent stabilization. Vertical wells maybe used to relieve artesian pressures. Usually,

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intercepting drains, laid approximately along con-tours, suffice.

Where mud flows may occur, water must beexcluded from the area. Surface and subsurfaceflow must be intercepted and conducted away atthe top of the area. Also, cover, such as heavymulching and planting, should be placed over theentire surface to prevent water from percolatingdownward into the soil. See also Art. 7.40.

Electrical drainage adapts the principle thatwater flows to the cathode when a direct currentpasses through saturated soil. The water may bepumped out at the cathode. Electro-osmosis is rela-tively expensive and therefore usually is limited tospecial conditions, such as drainage of silts, whichordinarily are hard to drain by other methods.

Vertical drains, or piles, may be used to com-pact loose, saturated cohesionless soils or to con-solidate saturated cohesive soils. They provide anescape channel for water squeezed out of the soilby an external load. A surcharge of perviousmaterial placed over the ground surface also servesas part of the drainage system as well as part ofthe fill, or external load. Usually, the surcharge isplaced before the vertical drains are installed, tosupport equipment, such as pile drivers, over thesoft soil. Fill should be placed in thin layers toavoid formation of mud flows, which might shearthe sand drains and cause mud waves. Analysesshould be made of embankment stability at variousstages of construction.

7.38 Thermal Stabilizationof Soils

Thermal stabilization generally is costly and isrestricted to conditions for which other methodsare not suitable. Heat may be used to strengthennonsaturated loess and to decrease the compressi-bility of cohesive soils. One technique is to burnliquid or gas fuel in a borehole.

Freezing a wet soil converts it into a rigidmaterial with considerable strength, but it must bekept frozen. The method is excellent for a limitedexcavation area, for example, freezing the groundto sink a shaft. For the purpose, a network of pipesis placed in the ground and a liquid, usually brine,at low temperature is circulated through the pipes.Care must be taken that the freezing does notspread beyond the area to be stabilized and causeheaving damage.

7.39 Chemical Stabilizationof Soils

Utilizing, portland cement, bitumens, or othercementitous materials, chemical stabilization meetsmany needs. In surface treatments, it supplementsmechanical stabilization to make the effects morelasting. In subsurface treatments, chemicals may beused to improve bearing capacity or decreasepermeability.

Soil-cement, a mixture of portland cement andsoil, is suitable for subgrades, base courses, andpavements of roads not carrying heavy traffic(“Essentials of Soil-Cement Construction,” Port-land Cement Association). Bitumen-soil mixturesare extensively used in road and airfield construc-tion and sometimes as a seal for earth dikes(“Guide Specifications for Highway Construction,”American Association of State Highway andTransportation Officials, 444 North Capitol St.,N.W., Washington, DC 20001 (www.ashto.org)).Hydrated, or slaked, lime may be used alone as asoil stabilizer, or with fly ash, portland cement, orbitumen (“Lime Stabilization of Roads,” NationalLime Association, 200 North Globe Road, Suite 800Arlington, VA 22203 (www.lime.org)). Calcium orsodium chloride is used as a dust palliative and anadditive in construction of granular base andwearing courses for roads (“Calcium Chloride forStabilization of Bases and Wearing Courses,”Calcium Chloride Institute).

Grouting, with portland cement or other chem-icals, often is used to fill rock fissures, decrease soilpermeability, form underground cutoff walls toeliminate seepage, and stabilize soils at consider-able depth. The chemicals may be used to fill thevoids in the soil, to cement the particles, or to forma rocklike material.

(K. Terzaghi and R. B. Peck, “Soil Mechanics inEngineering Practice,” John Wiley & Sons, Inc.,New York (www.wiley.com); G. P. Tschebotarioff,“Soil Mechanics, Foundations, and Earth Struc-tures,” McGraw-Hill Book Company, New York(books.mcgraw-hill.com); H. Y. Fang, “FoundationEngineering Handbook,” 2nd ed., Van NostrandReinhold Company, New York.)

7.40 Geosynthetics

In the past, many different materials have beenused for soil separation or reinforcement, including

Geotechnical Engineering n 7.107

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grasses, rushes, wood logs, wood boards, metalmats, cotton, and jute. Because they deterioratedin a relatively short time, required maintenancefrequently, or were costly, however, use of moreefficient, more permanent materials was desirable.Synthetic fabrics, grids, nets, and other structuresare now used as an alternative.

Types of geosynthetics, polymer compositionsgenerally used, and properties important forspecifying materials to achieve desired perform-ance are described in Art. 5.29. Principal appli-cations of geosynthetics, functions of geosyntheticsin those applications, recommended structures foreach case, and design methods are discussed in thefollowing. Table 7.20 summarizes the primaryfunctions of geosynthetics in applications oftenused and indicates the type of geosyntheticgenerally recommended by the manufacturers ofthese materials for the applications. (“Geosynthe-tic Design and Construction Guidelines” FederalHighway Administration Publication No. FHWA-HI-95-038, May, 1995).

7.40.1 Design Methods forGeosynthetics

The most commonly used design methods forgeosynthetics in geotechnical applications are theempirical (design by experience), specification, andrational (design by function) methods.

The empirical design process employs a selec-tion process based on the experience of the projectgeo-technical engineer or of others, such asdesigners of projects reported in engineering liter-ature, manufacturers of geosynthetics, and profes-sional associations.

Design by specification is often used for routineapplications. Standard specifications for specificapplications may be available from geosyntheticsmanufacturers or developed by an engineeringorganization or a government agency for its ownuse or by an association or group of associations,such as the joint committee established by theAmerican Association of State Highway andTransportation Officials, Associated General Con-tractors, and American Road and TransportationBuilders Association (Art. 5.29).

When using the rational design method, design-ers evaluate the performance, construction methodsrequired, and durability under service conditions ofgeosynthetics that are suitable for the planned ap-plication. This method can be used for all site con-ditions to augment the preceding methods. It isnecessary for applications not covered by standardspecifications. It also is required for projects of suchnature that large property damage or personal injurywould result if a failure should occur. The methodrequires the following:

A decision as to the primary function of a geo-synthetic in the application under consideration

Table 7.20 Primary Function of Geosynthetics in Geotechnical Applications

Application Function Geosynthetic

Subgrade stabilization Reinforcement, separation, filtration Geotextile or geogridRailway trackbed stabilization Drainage, separation, filtration GeotextileSedimentation-control silt fence Sediment retention, filtration, separation GeotextileAsphalt overlays Stress-relieving layer and waterproofing GeotextileSoil reinforcement:Embankments Reinforcement Geotextile or geogridSteep slopes Reinforcement Geotextile or geogridRetaining walls Reinforcement Geotextile or geogrid

Erosion control:Reinforcement Reinforcement, separation GeocompositeRiprap Filtration and separation GeotextileMats Filtration and separation Geotextile

Subsurface drainage filter Filtration GeotextileGeomembrane protection Protection and cushion GeotextileSubsurface drainage Fluid transmission and filtration Prefabricated

drainage composite

7.108 n Section Seven

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Estimates or calculations to establish the requiredproperties (design values) of the material for theprimary function

Determination of the allowable properties, such asminimum tensile or tear strength or permittivity, ofthe material by tests or other reliable means

Calculation of the safety factor as the ratio of allow-able to design values

Assessment of this result to ascertain that it issufficiently high for site conditions

(“A Design Primer: Geotextiles and RelatedMaterials,” Industrial Fabric Association Inter-national, 345 Cedar St., Suite 800, St. Paul, MN55101; “Geotextile Testing and the Design Engin-eer,” STP 952, ASTM; R. M. Koerner, “Designingwith Geosynthetics,” 2nd ed., Prentice-Hall, Engle-wood Cliffs, N. J.)

7.40.2 Geosynthetics Nomenclature

Following are some of the terms generally used indesign and construction with geosynthetics:

Apparent Opening Size (AOS). A property desig-nated O95 applicable to a specific geotextile thatindicates the approximate diameter of the largestparticle that would pass through the geotextile. Aminimum of 95% of the openings are the same sizeor smaller than that particle, as measured by thedry sieve test specified in ASTM D4751.

Blinding. Blocking by soil particles of openingsin a geotextile, as a result of which its hydraulicconductivity is reduced.

Chemical Stability. Resistance of a geosyntheticto degradation from chemicals and chemical reac-tions, including those catalyzed by light.

Clogging. Retention of soil particles in the voidsof a geotextile, with consequent reduction in thehydraulic conductivity of the fabric.

Cross-Machine Direction. The direction withinthe plane of a fabric perpendicular to the directionof manufacture. Generally, tensile strength of thefabric is lower in this direction than in the machinedirection.

Denier. Mass, g, of a 9000-m length of yarn.

Fabric. Polymerfibers or yarn formed into a sheetwith thickness so small relative to dimensions in

the plane of the sheet that it cannot resist com-pressive forces acting in the plane. A needle-punched fabric has staple fibers or filamentsmechanically bonded with the use of barbedneedles to form a compact structure. A spun-bonded fabric is formed with continuous filamentsthat have been spun (extruded), drawn, laid into aweb, and bonded together in a continuous process,chemically, mechanically, or thermally. A wovenfabric is produced by interlacing orthogonally twoor more sets of elements, such as yarns, fibers,rovings, or filaments, with one set of elements in themachine direction. Amonofilament woven fabric ismade with single continuous filaments, whereas amultifilament woven fabric is composed of bundlesof continuous filaments. A split-film woven fabricis constructed of yarns formed by splitting longi-tudinally a polymeric film to forma slit-tape yarn.Anonwoven fabric is produced by bonding orinterlocking of fibers, or both.

Fiber. Basic element of a woven or knitted fabricwith a length-diameter or length-width ratio of atleast 100 and that can be spun into yarn orotherwise made into a fabric.

Filament. Variety of fiber of extreme length, notreadily measured.

Filtration. Removal of particles from a fluid orretention of soil particles in place by a geosynthetic,which allows water or other fluids to pass through.

Geocomposite. Manufactured laminated or com-posite material composed of geotextiles, geomem-branes, or geogrids, and sometimes also naturalmaterials, or a combination.

Geogrid. Orthogonally arranged fibers, strands,or rods connected at intersections, intended for useprimarily as tensile reinforcement of soil or rock.

Geomembrane. Geosynthetic, impermeable ornearly so, intended for use in geotechnical appli-cations.

Geosynthetics. Materials composed of polymersused in geotechnical applications.

Geotextile. Fabric composed of a polymer andused in geotechnical applications.

Grab Tensile Strength. Tensile strength deter-mined in accordance with ASTM D4632 and typi-cally found from a test on a 4-in-wide strip offabric, with the tensile load applied at the midpointof the fabric width through 1-in-wide jaw faces.

Geotechnical Engineering n 7.109

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Gradient Ratio. As measured in a constant-headpermittivity test on a geotextile, the ratio of theaverage hydraulic gradient across the fabric plus1 in of soil adjoining the fabric to the averagehydraulic gradient of the 2 in of soil between 1 and3 in above the fabric.

Machine Direction. The direction in the plane ofthe fabric parallel to the direction of manufacture.Generally, the tensile strength of the fabric is largestin this direction.

Monofilament. Single filament, usually of adenier higher than 15.

Mullen Burst Strength. Hydraulic burstingstrength of a geotextile as determined in accor-dance with ASTM D3786.

Permeability (Hydraulic Conductivity). A measureof the capacity of a geosynthetic to allow a fluid tomove through its voids or interstices, as representedby the amount of fluid that passes throughthe material in a unit time per unit surface areaunder a unit pressure gradient. Accordingly, perm-eability is directly proportional to thickness of thegeosynthetic.

Permittivity. Like permeability, a measure of thecapacity of a geosynthetic to allow a fluid to movethrough its voids or interstices, as represented bythe amount of fluid that passes through a unitsurface area of the material in a unit time per unitthickness under a unit pressure gradient, withlaminar flow in the direction of the thickness ofthe material. For evaluation of geotextiles, use ofpermittivity, being independent of thickness, ispreferred to permeability.

Puncture Strength. Ability of a geotextile to resistpuncture as measured in accordance with ASTMD3787.

Separation. Function of a geosynthetic to preventmixing of two adjoining materials.

Soil-Fabric Friction. Resistance of soil by frictionto sliding of a fabric embedded in it, exclusive ofresistance from cohesion. It is usually expressed asa friction angle.

Staple Fibers. As usually used in geotextiles,very short fibers, typically 1 to 3 in long.

Survivability. Ability of geosynthetics to performintended functions without impairment.

Tearing Strength. Force required either tostart or continue propagation of a tear in afabric as determined in accordance with ASTMD4533.

Tenacity. Fiber strength, grams per denier.

Tex. Denier divided by 9.

Transmissivity. Amount of fluid that passes inunit time under unit pressure gradient with lami-nar flow per unit thickness through a geosyntheticin the in-plane direction.

Yarn. Continuous strand composed of textilefibers, filaments, or material in a form suitable forknitting, weaving, or otherwise intertwining toform a geotextile.

7.40.3 Geosynthetic Reinforcementof Steep Slopes

Geotextiles and geogrids are used to reinforce soilsto permit slopes much steeper than the shearingresistance of the soils will permit. (Angle of repose,the angle between the horizontal and the maxi-mum slope that a soil assumes through naturalprocesses, is sometimes used as a measure of thelimiting slopes for unconfined or unreinforced cutsand fills, but it is not always relevant. For dry,cohesionless soils, the effect of height of slope onthis angle is negligible. For cohesive soils, incontrast, the height effect is so large that angle ofrepose is meaningless.) When geosynthetic rein-forcement is used, it is placed in the fill in horizon-tal layers. Vertical spacing, embedment length, andtensile strength of the geosynthetic are critical inestablishing a stable soil mass.

For evaluation of slope stability, potentialfailure surfaces are assumed, usually circular orwedge-shaped but other shapes also are possible.Figure 7.56a shows a slope for which a circularfailure surface starting at the bottom of the slopeand extending to the ground surface at the topis assumed. An additional circular failure sur-face is indicated in Fig. 7.56b. Fig. 7.56c shows awedge-shaped failure surface. An infinite num-ber of such failure surfaces are possible. Fordesign of the reinforcement, the surfaces are as-sumed to pass through a layer of reinforcement atvarious levels and apply tensile forces to the rein-forcement, which must have sufficient tensilestrength to resist them. Sufficient reinforcement

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embedment lengths extending into stable soilbehind the surfaces must be provided to ensurethat the geosynthetic will not pull out at designloads.

Pull-out is resisted by geotextiles mainly byfriction or adhesion—and by geogrids, which havesignificant open areas, also by soil-particle strike-through. The soil-fabric interaction is determined inlaboratory pull-out tests on site-specific soils andthe geosynthetic to be used, but long-term load-transfer effects may have to be estimated. Designof the reinforcement requires calculation of the

embedment required to develop the reinforcementfully and of the total resisting force (numberof layers and design strength) to be provided by thereinforcement. The design should be based onsafety factors equal to or greater than those re-quired by local design codes. In the absence of localcode requirements, the values given in Table 7.21may be used. A stability analysis should be per-formed to investigate, at a minimum, circular andwedge-shaped failure surfaces through the toe (Fig.7.56a), face (Fig. 7.56c), and deep seated below thetoe (Fig. 7.56b). The total resisting moment for a

Fig. 7.56 Stabilization of a steep slope with horizontal layers of geosynthetic reinforcement. (a)Primary reinforcement for a circular failure surface. (b) Embedment lengths of reinforcement extend fromcritical failure surfaces into the backfill. (c) Intermediate reinforcement for shallow failure surfaces.

Geotechnical Engineering n 7.111

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circular slip surface can be determined fromFig. 7.56b as

MR ¼ RFr þXi¼n

i¼1

RiTi (7:90)

where R ¼ radius of failure circle

FR ¼ soil shearing resistance along the slipsurface ¼ tfLsp

tf ¼ soil shear strength

Lsp ¼ length of slip surface

Ri ¼ radius of slip surface at layer i

Ti ¼ strength of the reinforcing required forlayer i

The driving moment, or moment of the forcescausing slip, is

MD ¼ Wrþ Sd (7:91)

where W ¼ weight of soil included in assumedfailure zone (Fig. 7.56a)

r ¼ moment arm of W with respect to thecenter of rotation (Fig. 7.56a)

S ¼ surcharge

d ¼ moment arm of S with respect to thecenter of rotation (Fig. 7.56a)

The safety factor for the assumed circular failuresurface then is

KD ¼ MR

MD(7:92)

A safety factor should be computed for eachpotential failure surface. If a safety factor is lessthan the required minimum safety factor for pre-vention of failure of the soil unreinforced, either astronger reinforcement is required or the numberof layers of reinforcement should be increased.This procedure may also be used to determine thereinforcement needed at any level to prevent fail-ure above that layer.

The next step is calculation of length Le ofreinforcement required for anchorage to preventpullout.

Le ¼ KFD2so tanfsr

(7:93)

where FD ¼ required pull-out strength

K ¼ minimum safety factor: 1.5 for cohe-sionless soils; 2 for cohesive soils

so ¼ overburden pressure above the rein-forcing level ¼ wh

w ¼ density of the soil

h ¼ depth of overburden

fsr ¼ soil-reinforcement interaction angle,determined from pull-out tests

Table 7.21 Minimum Safety Factors K for Slope Reinforcement

External Stability Internal Stability

Condition K Condition K

Sliding 1.5 Slope stability 1.3Deep seated (overall stability) 1.3 Design tensile strength Td *Dynamic loading 1.1 Allowable geosynthetic

strength Ta†

Creep 4Construction 1.1 to 1.3Durability 1.1 to 1.2Pull-out resistanceCohesionless soils 1.5‡

Cohesive soils 2

* Td at 5% strain should be less than Ta.† Ta ¼ TL/KcKd, where TL is the creep limit strength, Kc is the safety factor for construction, and Kd is the safety factor for durability. In

the absence of creep tests or other pertinent data, the following may be used: Ta ¼ Tu/10.4 or Tu � 10.4Td, where Tu is the ultimate tensilestrength of the geosynthetic.

‡ For a 3-ft minimum embedment.

7.112 n Section Seven

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Embedment length Le should be at least 3 ft. Thetotal length of a reinforcement layer then is Le plusthe distance from the face of the slope to the failurecircle (Fig. 7.56b). The total length of the reinforce-ment at the toe should be checked to ascertain thatit is sufficient to resist sliding of the soil mass abovethe base of the slope.

Among the family of potential failure surfacesthat should be investigated is the wedge shape,such as the one shown in Fig. 7.56c. To reinforce thefailure zones close to the face of the slope, layers ofreinforcement are required in addition to thoseprovided for the deep failure zones, as indicated inFig. 7.56c. Such face reinforcement should have amaximum vertical spacing of 18 in and a minimumlength of 4 ft. Inasmuch as the tension in this rein-forcement is limited by the short embedment,a geosynthetic with a smaller design allowabletension than that required for deep-failurereinforcement may be used. In construction of thereinforced slope, fill materials should be placed sothat at least 4 in of cover will be between thegeosynthetic reinforcement and vehicles or equip-ment operating on a lift. Backfill should notincorporate particles larger than 3 in. Turning ofvehicles on the first lift above the geosyntheticshould not be permitted. Also, end dumping offill directly on the geosynthetic should not beallowed.

7.40.4 Geosynthetics inRetaining-WallConstruction

Geotextiles and geogrids are used to form retainingwalls (Fig. 7.57a) or to reinforce the backfill of aretaining wall to create a stable soil mass (Fig.7.57b). In the latter application, the reinforcementreduces the potential for lateral displacement of thewall under the horizontal pressure of the backfill.

As in the reinforcement of steep slopes discussedin Art. 7.40.3, the reinforcement layers must inter-sect all critical failure surfaces. For cohesionlessbackfills, the failure surface may be assumed to bewedge shaped, as indicated in Fig. 7.56c, with thesloping plane of the wedge at an angle of 458 þ f/2with the horizontal. If the backfill is not homo-geneous, a general stability analysis should becarried out as described in Art. 7.40.3.

The design process for cohesionless soils can besimplified by use of a constant vertical spacing svfor the reinforcement layers. This spacing would beapproximately

Sv ¼ Ta

KKawH(7:94)

where Ta ¼ allowable tension in the reinforcement

K ¼ safety factor as specified in a local codeor as given in Table 7.21

Fig. 7.57 Geosynthetic applications with retaining walls: (a) Reinforced earth forms a retaining wall.(b) Retaining wall anchored into backfill.

Geotechnical Engineering n 7.113

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Ka ¼ coefficient of active earth pressure (Art.7.27)

w ¼ density of backfill

H ¼ average height of embankment

If Eq. (7.94) yields a value for sv less than theminimum thickness of a lift in placement of thebackfill, a stronger geosynthetic should be chosen.The minimum embedment length Le may be com-puted from Eq. (7.93). Although the total reinforce-ment length thus computed may vary from layer tolayer, a constant reinforcement length would beconvenient in construction.

When the earth adjoining the backfill is arandom soil with lower strength than that of thebackfill, the random soil exerts a horizontal pres-sure on the backfill that is transmitted to the wall(Fig. 7.58). This may lead to a sliding failure of thereinforced zone. The reinforcement at the baseshould be sufficiently long to prevent this type offailure. The total horizontal sliding force on thebase is, from Fig. 7.58,

P ¼ Pb þ Ps þ Pv (7:95)

where Pb ¼ KawbH2/2

wb ¼ density of soil adjoining the reinforce-ment zone

Ps ¼ KawshH

wsh ¼ weight of uniform surcharge

Pv ¼ force due to live load V as determinedby the Boussinesq method (Art. 7.11)

The horizontal resisting force is

FH ¼ [(wshþ wrH) tanfsr þ c]L (7:96)

where wrH ¼ weight of soil in the reinforcementzone

fsr ¼ soil-reinforcement interaction angle

c ¼ undrained shear strength of thebackfill

L ¼ length of the reinforcement zone base

The safety factor for sliding resistance then is

Ksl ¼ FHP

(7:97)

and should be 1.5 or larger. A reinforcement lengthabout 0.8H generally will provide base resistancesufficient to provide a safety factor of about 1.5.

The most economical retaining wall is one inwhich the reinforcement is turned upward andbackward at the face of the wall and also serves asthe face (Fig. 7.58a) The backward embedmentshould be at least 4 ft. If desired for esthetic reasons

Fig. 7.58 Retaining wall anchored with geosynthetic reinforcement is subjected to pressure fromrandom-soil backfill, sand backfill, surcharge, and live load. Assumed pressure distribution diagramsare rectangular and triangular.

7.114 n Section Seven

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or to protect the geosynthetic from damage ordeterioration from exposure to ultraviolet light,sprayed concrete may be applied to the wall face.

As an alternative, the wall may be composed ofconcrete block or precast concrete panels that areanchored to soil reinforcement. The reinforcementshould be installed taut to limit lateral movementof the wall during construction.

See Art. 7.40.3 for other precautions to be takenduring construction.

7.40.5 Geosynthetic Reinforcementfor Embankments

Geosynthetics placed in horizontal layers may beused to reinforce embankments in a manner simi-lar to that used to reinforce steep slopes (Art.7.40.3). The reinforcement may permit greaterembankment height and a larger safety factor inembankment design than would an unreinforcedembankment. Also, displacements during con-struction may be smaller, thus reducing fill re-quirements. Furthermore, reinforcement properlydesigned and installed can prevent excessive hori-zontal displacements along the base that can causean embankment failure when the underlying soil isweak. Moreover, reinforcement may decrease hori-zontal and vertical displacements of the underlyingsoil and thus limit differential settlement. Rein-forcement, however, will reduce neither long-termconsolidation of the underlying weak soil norsecondary settlement.

Either geotextiles or geogrids may be used asreinforcement. If the soils have very low bearingcapacity, it may be necessary to use a geotextileseparator with geogrids for filtration purposes andto prevent the movement of the underlying soilinto the embankment fill.

Figure 7.59 illustrates reinforcement of anembankment completely underlain by a weak soil.Without reinforcement, horizontal earth pressurewithin the embankment would cause it to spreadlaterally and lead to embankment failure, in theabsence of sufficient resistance from the soil. Rein-forcement is usually laid horizontally in the direc-tion of major stress; that is, with strong axis normalto the longitudinal axis of the embankment. Rein-forcement with strong axis placed parallel to thelongitudinal axis of the embankment may also berequired at the ends of the embankment. Seamsshould be avoided in the high-stress direction.

Design of the reinforcement is similar to thatrequired for steep slopes (Art. 7.40.3).

For an embankment underlain by locally weakareas of soil or voids, reinforcement may be incor-porated at the base of the embankment to bridgethem.

7.40.6 Soil Stabilization withGeosynthetics

Woven or nonwoven geotextiles are used toimprove the load-carrying capacity of roads overweak soils and to reduce rutting. Acting primarilyas a separation barrier, the geosynthetic preventsthe subgrade and aggregate base from mixing. Thegeosynthetic may also serve secondary functions.Acting as a filter, it prevents fines from migratinginto the aggregate due to highwater pressure. Also,the geotextile may facilitate drainage by allowingpore water to pass through and dissipate into theunderlying soil. In addition, acting as reinforce-ment, the geotextile can serve as a membranesupport of wheel loads and provide lateral restraintof the base and subgrade through friction betweenthe fabric, aggregate, and soil.

Installation techniques to be used depend on theapplication. Usually, geosynthetics are laid directlyon the subgrade (Fig. 7.60a). Aggregate then isplaced on top to desired depth and compacted.

Design of permanent roads and highwaysconsists of the following steps: If the CBR � 3,need for a geotextile is indicated. The pavement isdesigned by usual methods with no allowance forstructural support from the geotextile. If a thickersubbase than that required for structural supportwould have to be specified because of the sus-

Fig. 7.59 Geosynthetic reinforcement for anembankment on weak soil is placed directly onthe subgrade.

Geotechnical Engineering n 7.115

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ceptibility of the underlying soil to pumping andsubbase intrusion, the subbase may be reduced50% and a geotextile selected for installation at thesubbase-subgrade interface. For stabilization of thesubgrade during construction, an additional deter-mination of thickness of subbase assisted by a geo-textile is made by conventional methods (bearingcapacityNc about 3.0 without geotextiles and about5.5 with) to limit rutting to a maximum of 3 inunder construction vehicle loads. The thickersubbase thus computed is selected. Then, thegeotextile strength requirements for survivabilityand filtration characteristics are checked. (Detailsfor this are given in B. R. Christopher and R. D.Holtz, ”Geotextile Design and Construction Guide-lines,“ FHWADTFH61-86-C-00102, National High-way Institute, Federal Highway Administration,Washington, DC 20590 (www.fhwa.dot.gov).)

Geosynthetics are also used under railwaytracks for separation of subgrade and subballastor subballast and ballast (Fig. 7.60b). They also areused for roadbed filtration, lateral permeability,and strength and modulus improvement.

7.40.7 Geosynthetics in ErosionControl

For purposes of erosion control, geosynthetics areused as turf reinforcement, as separators and filters

under riprap, or armor stone, and as an alternativeto riprap. Different types of geosynthetics are usedfor each of these applications.

Turf Control n To establish a reinforced turf inditches and water channels and on slopes, three-dimensional erosion-control mats often are used.Entangling with the root and stem network ofvegetation, they greatly increase resistance to flowof water down slopes and thus retard erosion.

Mats used for turf reinforcement should have astrong, stable structure. They should be capable ofretaining the underlying soil but have sufficientporosity to allow roots and stem to grow throughthem. Installation generally requires pinning themat to the ground and burying mat edges andends. Topsoil cover may be used to reduce erosioneven more and promote rapid growth of vege-tation.

When a geosynthetic is placed on a slope, itshould be rolled in the direction of the slope.Horizontal joints should not be permitted. Verticaljoints should be shingled downstream. Ditch andchannel bottoms should be lined by rolling thegeosynthetic longitudinally. Joints transverse towater flow should have a 3-ft overlap and beshingled downstream. Roll edges should be over-lapped 2 to 4 in. They should be staked at intervalsnot exceeding 5 ft to prevent relative movement.

Fig. 7.60 Geosynthetic is used (a) to reinforce a road, (b) to reinforce a railway roadbed.

7.116 n Section Seven

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Where highly erodible soils are encountered, ageotextile filter should be installed under the turfreinforcement and staked or otherwise bonded tothe mats. For stability and seeding purposes, woodchips may be used to infill the turf reinforcement.

Use of Geosynthetics with Riprap n

Large armor stones are often used to protect soilagainst erosion and wave attack. Graded-aggre-gate filter generally is placed between the soil andthe riprap to prevent erosion of the soil through thearmoring layer. As a more economical alternative,geotextiles may be used instead of aggregate. Theyalso offer greater control during construction, es-pecially in underwater applications. Geosyntheticstypically used are nonwoven fabrics, monofilamentnonwoven geotextiles, and multifilament or fibril-lated woven fabrics.

The geosynthetics should have sufficient per-meability to permit passage of water to relievehydrostatic pressure behind the riprap. Also, thegeosynthetic should be capable of retaining theunderlying soil. Conventional filter criteria can beused for design of.the geosynthetic, although somemodifications may be required to compensate forproperties of the geosynthetic.

Installation precautions that should be observedinclude the following: Riprap should be installedwith care to avoid tearing the geosynthetic inas-much as holes would decrease its strength. Stoneplacement, including drop heights, should betested in field trials to develop techniques that willnot damage the geosynthetic. As a general guide,for material protected by a sand cushion andmaterial with properties exceeding that requiredfor unprotected applications, drop height forstones weighing less than 250 lb should not exceed3 ft; without a cushion, 1 ft. Stone weighing morethan 250 lb should be placed without free fall,unless field tests determine a safe drop height.Stone weighing more than 100 lb should not bepermitted to roll along the geosynthetic. Installa-tion of the armor layer should begin at the base ofslopes and at the center of the zone covered by thegeosynthetic. After the stones have been placed,they should not be graded.

Special construction procedures are requiredfor slopes greater than 2.5 :1. These includeincrease in overlap, slope benching, eliminationof pins at overlaps, toe berms for reaction againstslippage, and laying of the geosynthetic suffi-

ciently loose to allow for downstream movement,but folds and wrinkles should not be permitted.

The geosynthetic should be rolled out with itsstrong direction (machine direction for geotextiles)up and down the slope. Adjoining rolls should beseamed or shingle overlapped in the downslope ordownstream direction. Joints should be stapled orpinned to the ground. Recommended pin spacingis 2 ft for slopes up to 3 :1, 3 ft for slopes between3:1 and 4:1, 5 ft for 4 :1 slopes, and 6 ft for slopessteeper than 4:1. For streambanks and slopesexposed to wave action, the geosynthetic should beanchored at the base of the slope by burial aroundthe perimeter of a stone-filled key trench. It shouldalso be keyed at the top of the slope if the armor-geosynthetic system does not extend several feetabove high water.

Riprap Replacement n Instead of the riprapgenerally used for erosion control, concrete matsmay be used. For this purpose, the concreteconventionally has been cast in wood or steelforms. Use of expandable fabric forms, however,may be more economical. Such forms are made byjoining two fabric sheets at discrete points. Afterthe sheets are placed over the area to be protected,grout is pumped into the space between the sheetsto form a mattress that initially will conform to theshape of the ground and later harden. Thickness ofthe mattress is controlled by internal spacerthreads. Filter points and bands are formed in themattress to dissipate pore water from the subsoil.The fabric forms may be grouted underwater, evenin flowing water, and in hazardous-liquid con-ditions. The fabric usually used is a wovengeotextile.

7.40.8 Uses of Geosynthetics inSubsurface Drainage

Subsurface drainage is required for many con-struction projects and geotextiles find many usesin such applications. Their primary function is toserve, with graded granular filter media, as apermeable separator to exclude soil from thedrainage media but to permit water to pass freely.Nonwoven geotextiles are usually used for thispurpose because of their high flow capacity andsmall pore size. Generally, fabric strength is not a

Geotechnical Engineering n 7.117

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primary consideration for subsurface drainageapplications, except during installation.

Following are brief descriptions of typicalapplications of geotextiles in subsurface drainage:

Permeable separators wrapped around trench oredge drains

Drains for retaining walls and bridge abutmentswith the geotextile enclosing the backfill

Encirclement of slotted or jointed drains and wallpipes to prevent filter particles from entering thedrains while permitting passage of water

Wraps for interceptor, toe, and surface drains onslopes to assist stabilization by dissipating excesspore-water pressures and retarding erosion

Seepage control with chimney and toe drains forearth dams and levees with the geotextile laidalong the upstream face and anchored by a berm.

7.40.9 Geosynthetics as PondLiners

Geomembranes, being impermeable, appear to bean ideal material for lining the bottom of a pond toretain water or other fluid. Used alone, however,they have some disadvantages. In particular, theyare susceptible to damage from many sources andrequire a protective soil cover of at least 12 in. Also,for several reasons, it is advisable to lay a geotextileunder the geomembrane. The geotextile provides aclean working area for making seams. It addspuncture resistance to the liner. It increases frictionresistance at the interface with the soil, thuspermitting steeper side slopes. And it permitslateral and upward escape of gases emitted fromthe soil. For this purpose, needle-punched non-woven geotextiles, geonets, or drainage compositeswith adequate transmissivity for passing the gasesare needed. In addition, it is advantageous to coverthe top surface of the geomembrane with anothergeotextile. Its purpose is to maintain stability ofcover soil on side slopes and to prevent sharpstones that may be present in the cover soil frompuncturing the liner. This type of construction isalso applicable to secondary containment ofunderground storage tanks for prevention ofleakage into groundwater.

In selection of a geosynthetic for use as a pondliner, consideration should be given to its chemical

resistance with regard to the fluid to be containedand reactive chemicals in the soil. For determi-nation of liner thickness, assumptions have to bemade as to loads from equipment during installa-tion and basin cleaning as well as to pressure fromfluid to be contained.

7.40.10 Geosynthetics as LandfillLiners

Liners are used along the bottom and sides oflandfills to prevent leachate formed by reaction ofmoisture with landfill materials from contami-nating adjacent property and groundwater. Clayliners have been traditional for this purpose(Fig. 7.61a). They have the disadvantage of beingthick, often in the range of 2 to 6 ft, and beingsubject to piping under certain circumstances,permitting leakage of leachate. Geomembranes,geotextiles, geonets, and geocomposites offer analternative that prevents rather than just minimizesleachate migration from landfills.

The U. S. Environmental Protection Agency(EPA) requires that all new hazardous-waste land-fills, surface impoundments, and waste piles havetwo ormore liners with a leachate-collection systembetween the liners. This requirement may besatisfied by installation of a top liner constructedof materials that prevent migration of any constitu-ent into the liner during the period such facilityremains inoperationanda lower linerwith the sameproperties. In addition, primary leachate-collectionand leak-detection systems must be installed withthe double liners to satisfy the following criteria:

The primary leachate-collection system should becapable of keeping the leachate head from excee-ding 12 in.

Collection and leak-detection systems should in-corporate granular drainage layers at least 12 inthick that are chemically resistant to the waste andleachate. Hydraulic conductivity should be at least0.02 ft/min. An equivalent drainage geosynthetic,such as a geonet, may be used instead of granularlayers. Bottom slope should be at least 2%.

A granular filter or a geotextile filter should beinstalled in the primary system above the drainagelayer to prevent clogging.

When gravel is used as a filter, pipe drains resistantto chemicals should be installed to collect leachateefficiently (Fig. 7.61a).

7.118 n Section Seven

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Figure 7.61b illustrates a lining system thatmeets these criteria. Immediately underlying thewastes is a geotextile that functions as a filter. Itoverlies the geocomposite primary leachate drain.Below is the primary liner consisting of a geo-membrane above a clay blanket. Next comes ageotextile filter and separator, followed underneathby a geonet that functions as a leak-detection drain.These are underlain by the secondary liner con-sisting of another geomembrane and clay blanket,which rests on the subsoil.

The EPA requires the thickness of a geomem-brane liner for containment of hazardous materialsto be at least 30 mils (0.75 mm) with timely cover or45 mils (1.2 mm) without such cover. The second-ary geomembrane liner should be the samethickness as the primary liner. Actual thicknessrequired depends on pressures from the landfilland loads from construction equipment duringinstallation of the liner system.

Terminals of the geosynthetics atop the sideslopes generally consist of a short runout and adrop into an anchor trench, which, after insertion ofthe geosynthetics, is backfilled with soil andcompacted. Side-slope stability of liner system andwastes needs special attention in design.

7.40.11 GeosyntheticsBibliography

“A Design Primer: Geotextiles and RelatedMaterials,” Industrial Fabrics Association Inter-national, 1801 County Road BW Roseville, MN55113, (www.ifai.com).

“Construction over Soft Soils,” The TensarCorporation, 1210 Citizens Parkway, Morrow, GA30260 (www.tensarcorp.com).

“Guidelines for Selection and Installation ofExxon Geotextiles” and “Exxon Geotextile DesignManual for Paved and Unpaved Roads,” ExxonChemical Americas, 380 Interstate North, Suite 375,Atlanta, GA 30339.

“Specifiers Guide,”(annual) Geotechnical FabricsReport, Industrial Fabrics Association International,St. Paul, Minn. (www.ifai.com)

“Woven Geotextiles” and “Nonwoven Geo-textiles,” Amoco Fabrics and Fibers Company, 900Circle 75 Parkway, Suite 300, Atlanta GA 30339.

Ahlrich, R. C., “Evaluation of Asphalt Rubberand Engineering Fabrics as Pavement Interlayers,”GL-86-34, U.S. Army Corps of Engineers, Vicks-burg, Miss.

Fig. 7.61 Landfill liner systems: (a) with filter soil, gravel, drain pipe, and clay liner; (b) with geotextileseparator-filter, geocomposite leachate drain, primary geomembrane and clay liner, geotextile filter,geonet for leak detection, and secondary geomembrane and clay liner.

Geotechnical Engineering n 7.119

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Berg, R. R., “Guidelines for Design, Specifica-tion, and Contracting of Geosynthetic Mechani-cally Stabilized Earth Slopes on Firm Foundations,”FHWA-SA-93-025, Federal Highway Adminis-tration, Washington, D. C.

Bonaparte, R., and B. Christopher, “Design andConstruction of Reinforced Embankments overWeak Foundations,” Record 1153, TransportationResearch Board, Washington, D. C.

Cedergren, H. R., “Seepage, Drainage, and FlowNets,” 3rd ed., John Wiley & Sons, Inc., New York(www.wiley.com).

Christopher, B. B., and R. D. Holtz, “GeotextileDesign and Construction Guidelines,” FHWADTFH61-86-C-00102, Federal Highway Admin-istration, Washington, D. C (www.fhwa.dot.gov).

Ingold, T. S., and and K. S. Miller, “GeotextilesHandbook,” Thomas Telford, London.

Kays, W. B., ”Construction of Linings forReservoirs, Tanks, and Pollution Control Facilities,“John Wiley & Sons, Inc., New York (www.wiley.com).

Koerner, R. M., “Designing with Geosyn-thetics,” 2nd ed., Prentice-Hall, Englewood Cliffs,N. J (www.prenhall.com).

Matrecon, Inc., “Lining of Waste ImpoundmentandDisposal Facilities,” EPA SW-870, U.S. Environ-mental Protection Agency, Washington, D.C.

Mitchell, J. K., et al., “Reinforcement of EarthSlopes and Embankments,” Report 290, Transpor-tation Research Board, Washington, D. C.

Richardson, G. N., and R. M. Koerner, “Geosyn-thetic Design Guidance for Hazardous WasteLandfill Cells and Surface Impoundments,” EPA68-03-3338, U.S. Environmental Protection Agency,Washington, D. C.

Rixner, J. J., et al., “Prefabricated VerticalDrains,” FHWA/RD-86/168, Federal HighwayAdministration, Washington, D. C (www.fhwa.dot.gov).

7.41 GeotechnicalEngineering Sites on theWorld Wide Web

American Association of State Highway andTransportation Officials (AASHTO), www.aashto.org

American Geological Institute, www.agiweb.org

American Geophysical Union (AGU), www.agu.org

American Rock Mechanics Association, www.armarocks.org

American Society of Civil Engineers, www.asce.org

American Society for Testing and Materials(ASTM), www.astm.org

Association of Engineering Geologists, www.aegweb.org

Association of Geotechnical & Geoenviron-mental Specialists (AGS), www.ags.org.uk

British Geotechnical Society (BGS), www.geo.org.uk

Canadian Geotechnical Society, www.cgs.ca

Deep Foundations Institute, www.dfi.org

Federal Highway Administration (FHWA),Bridge Technology, Geotechnical Section, www.fhwa.dot.gov/bridge/geo.htm

Geo-Institute of the American Society of CivilEngineers, www.geoinstitute.org

Geological Society of America, www.geosociety.org

Geosynthetic Institute (GSI), www.geosyn-thetic-institute.org

Geosynthetic Materials Association (GMA),www.gmanow.com

Geotechnical & Geoenvironmental SoftwareDirectory, www.ggsd.com

International Association of Foundation Drilling(ADSC), www.adsc-iafd.com

International Geosynthetic Society (IGS), http://mergrb.rmc.ca/

International Society for Soil Mechanics andGeotechnical Engineering (ISSMGE), www.issmge.org

Mexican Society of Soil Mechanics, www.smms.org.mx

National Council for Geo-Engineering andConstruction (Geo Council), www.geocouncil.org

National Geophysical Data Center (NGDC),www.ngdc.noaa.gov

7.120 n Section Seven

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National Geotechnical Experimentation Sites(NGES) www.geocouncil.org/nges/nges.html

National Science Foundation Directorate forEngineering (NSF), www.eng.nsf.gov

Pile Driving Contractors Association (PDCA),www.piledrivers.org

Professional Firms Practicing in the Geosciences(ASFE), www.asfe.org

United States Universities Council for Geotech-nical Engineering Research (USUCGER), www.usucger.org

U.S. Corps of Engineers Waterways ExperimentStation (WES), www.wes.army.mil

U.S. Geological Survey, www.usgs.gov

World Wide Web Virtual Library of Geotechni-cal Engineering, http://geotech.civen.okstate.edu/wwwvl/

Geotechnical Engineering n 7.121

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