Hazard mitigation and strengthening of unreinforced masonry walls using composites

Composite Structures 73 (2006) 458–477

www.elsevier.com/locate/compstruct

Hazard mitigation and strengthening of unreinforced masonrywalls using composites

W.W. El-Dakhakhni a,*, A.A. Hamid a, Z.H.R. Hakam b, M. Elgaaly c

a Department of Civil Engineering, McMaster University Centre for Effective Design of Structures, Hamilton, ON, Canada L8S 4L7b Bechtel Power Corporation, Frederick, MD 21703, USA

c Civil, Architectural and Environmental Engineering Department, Drexel University, Philadelphia, PA 19104, USA

Available online 9 April 2005

Abstract

An experimental investigation was conducted to study the behavior of unreinforced masonry (URM) walls retrofitted with com-

posite laminates. The first testing phase included testing 24 URM assemblages under different stress conditions present in masonry

walls. Tests included prisms loaded in compression normal and parallel to bed joints, diagonal tension specimens, and specimens

loaded under joint shear. In the second testing phase, five masonry-infilled steel frames were tested with and without retrofit.

The composite laminates increased the stiffness and strength and enhanced the post-peak behavior by stabilizing the masonry walls

and preventing their out-of-plane spalling. Tests reported in this paper demonstrate the efficiency of composite laminates in improv-

ing the deformation capacity of URM, containing the hazardous URM damage, preventing catastrophic failure and maintaining the

wall integrity even after significant structural damage.

� 2005 Elsevier Ltd. All rights reserved.

Keywords: Composite masonry; Concrete masonry; Fiber reinforced plastics; Retrofitting; Seismic hazard; Seismic loads; Steel frames

1. Introduction

Earthquakes have long been recognized as one of the

most damaging natural hazards, along with hurricanes,

tornadoes, floods and fire. No other force in nature

has the potential to wreak so much havoc in such a short

time. Earthquakes typically strike without warning and,

after only a few seconds, leave casualties and damage in

their wake. Although earthquakes cannot be prevented,

the current state-of-the-art in science and engineeringprovides new tools that can be used to reduce their dam-

aging effects. Through prudent action, the loss of life,

serious injury, and property damage as well as social

and economic disruptions resulting from earthquakes

can be reduced. The principal threat to human life and

0263-8223/$ - see front matter � 2005 Elsevier Ltd. All rights reserved.

doi:10.1016/j.compstruct.2005.02.017

* Corresponding author. Tel.: +1 905 525 9140x26109; fax: +1 905

529 9688.

E-mail address: [email protected] (W.W. El-Dakhakhni).

safety is the shaking damage and the collapse of build-

ings and other structures that have been inadequatelydesigned or poorly constructed. Major earthquakes

can severely disrupt regional or national economic activ-

ity by damaging social lifelines such as roads, railways,

water, power and communications infrastructures and

office and residential buildings.

A common type of construction in urban centers for

low-rise and mid-rise buildings is unreinforced masonry

(URM) walls filling the space bounded by the structuralframing members. Although considered non-structural

elements, yet under seismic excitation, infill walls tend

to interact with the surrounding frame and may result

in different undesirable failure modes both to the frame

and to the infill wall [1]. In general, URM infill walls

have demonstrated poor performance record even in

moderate earthquakes. Their behavior is usually brittle

with little or no ductility and they, typically, suffervarious types of damage ranging from invisible cracking

mailto:[email protected]

Nomenclature

DH horizontal extension

DV vertical shorteningc shear strain

AFRP cross-sectional area of the FRP per unit

length of wall

An net area of the diagonal tension specimen

As cross-sectional area of the steel reinforcement

per unit length of wall

EFRP modulus of elasticity of the FRP

Es modulus of elasticity of steel

g vertical gage length

h diagonal tension specimen heightn percentage of solid in the masonry unit

P applied load on the diagonal tension speci-

mens

Ss shear stress based on the net area

t diagonal tension specimen thickness

w diagonal tension specimen width

W.W. El-Dakhakhni et al. / Composite Structures 73 (2006) 458–477 459

to crushing and, eventually, disintegration and total col-

lapse. This behavior constitutes a major source of haz-

ard during seismic events and creates a major seismic

performance problem facing designers today.

Seismic upgrading by adding new structural frames

or shear walls, have been proven to be impractical, they

have been either too costly or restricted in use to certain

types of structures. Other strengthening methods such asgrout injection, insertion of reinforcing steel, prestress-

ing, jacketing and different surface treatments were sum-

marized elsewhere [2] and specified by the Federal

Emergency Management Agency documents [3,4]. Each

of these methods adds considerable mass and stiffness

leading to higher seismic loads. They also involve the

use of skilled labor and disrupt the normal function of

the building. The use of fiber reinforced polymer(FRP) laminates for retrofitting and strengthening is a

valid alternative because of their small thickness, high

strength-to-weight ratio, high stiffness, and relative ease

of application.

A strong earthquake introduces severe in-plane and

out-of-plane forces to masonry walls which may lead

to catastrophic collapse as seen in Fig. 1 during the

1999 Turkey earthquake. However, the majority ofwork conducted to date [5–11] has been concentrating

on the out-of-plane behavior of URM walls strength-

ened with externally applied FRPs. Infill panels (or large

portions of wall) may fall out of the surrounding frame

due to inadequate out-of-plane restraint at the frame–in-

fill interface, or due to out-of-plane flexural or shear fail-

ure of the infill panel. In undamaged infills, these failures

may result from out-of-plane inertial forces, especiallyfor infills at higher story levels and with large slender-

ness ratios. However, it is more likely for out-of-plane

failure to occur after the masonry units become dis-

lodged due to damage from in-plane loading [4].

The work presented herein investigates the effects of

applying FRP laminates on the in-plane behavior of

URM assemblages subjected to different stress condi-

tions present in masonry infill walls (Fig. 2c). One ofthe objectives of the present experimental study is to

investigate the effects of FRP laminates on altering the

failure modes and strength and deformation characteris-

tics of different assemblages. Another objective is to

demonstrate the potential of the FRP on enhancing

the shear and compressive strength of URM infill walls

and preventing brittle collapse by means of stabilizing

the face shell even after excessive damage. This would

also maintain the wall�s structural integrity and wouldreduce the possibility of URM walls collapsing and

spalling, which, in itself, is a major source of hazard dur-

ing earthquakes, even if the whole structure remained

safe and functioning.

2. Behavior of infill masonry walls

Masonry infill walls in frame structures have been

long known to affect the strength and stiffness of the in-

filled-frame structures. In seismic areas, ignoring the

frame–wall interaction is not always on the safe side,

since, under lateral loads, the infill walls dramatically in-

crease the stiffness by acting as a diagonal strut as seen

in Fig. 2a, thus resulting in possible change in the seis-

mic demand due to significant reduction in the naturalperiod of the composite structural system [1]. Also, the

composite action of the frame–wall system changes the

magnitude and the distribution of straining actions in

the frame members, i.e. critical sections in the infilled-

frame differ from those in the bare frame, which may

lead to unconservative or poorly detailed designs. More-

over, these designs may be uneconomical since an

important source of structural strength, which is partic-ularly beneficial in regions of low and, sometimes, mod-

erate seismic demand, is wasted. However, URM infill

walls exhibit poor seismic performance under moderate

and high seismic demand. This behavior is due to a ra-

pid degradation of stiffness, strength and low energy dis-

sipation capacity, resulting from the brittle and sudden

damage of the URM infill walls.

The rationale behind neglecting infill walls in thedesign process is partly attributed to incomplete

Fig. 1. Failure of masonry walls during Turkey earthquake, 1999: (a) out-of-plane failure, (b) in-plane failure and (c) combined in- and out-of-plane

failures.

Fig. 2. Behavior of masonry infill walls.

460 W.W. El-Dakhakhni et al. / Composite Structures 73 (2006) 458–477

knowledge of the behavior of quasi-brittle materials

such as URM and to the lack of conclusive experimentaland analytical results to substantiate a reliable design

procedure for this type of structures. On the other hand,

and because of the large number of interacting parame-

ters, if the infill wall is to be considered in the analysis

and design stages, a modeling problem arises because

of the many possible failure modes (Fig. 2b) that need

to be evaluated with a high degree of uncertainty. This

is why it is not surprising that no consensus has emergedleading to a unified approach for the design of infilled-

frame systems, despite more than five decades of

research. It is, however, generally accepted that, under

lateral loads, the infill wall acts as a diagonal strut

connecting the two loaded corners. However, this is only

applicable to the case of solid infill walls (i.e. with no

openings) failing in corner-crushing mode [1].

3. Experimental program

The experimental program consisted of two phases.

In Phase I, four different types of assemblages (Fig.

2c) were tested under different types of loading condi-

tions representing critical regions in an infill masonrywall.

• Axial compression: in-plane, concentric, compressive

loads were applied at 90� (normal to the bed joints)

and 0� (parallel to the bed joints).

Table 1

FRP composite and dry fibers properties

Composite laminate properties Dry fibers properties

Ultimate tensile strength in primary fibers direction (MPa) 309.0 Tensile strength (GPa) 3.24

Elongation at break (%) 1.6 Tensile modulus (GPa) 72.4

Tensile modulus (GPa) 19.3 Ultimate elongation (%) 4.5

Ultimate tensile strength 90� to primary fibers direction (MPa) 309.0 Density (g/cm3) 2.55

Laminate thickness (mm) 0.25 Weight (g/m2) 295.0


• Diagonal tension: this is a standard testing procedure

used to evaluate the diagonal tensile (or shear)

strength of URM and creates a state of stress similar

to that occurring in infill walls.

• Joint shear: this enabled evaluating the strengthening

effect of the FRP laminates against the traditionally

weak and brittle horizontal shear slip failure mode.

In Phase II, five single-story/single-bay, one-third

scale, moment-resisting, structural steel frames infilled

with unretrofitted and retrofitted hollow block ma-

sonry walls were tested under displacement controlled

diagonal loading to evaluate the behavior of the

composite frame–wall system. Two series were

considered:

• Weak frames series: including three identical steel

frames tested as a bare (i.e. no infill) frame and two

frames were infilled with URM, one of which was ret-

rofitted with FRP.

• Strong frames series: including two identical steel

frames tested as both were infilled with URM and

one of which was retrofitted with FRP.

3.1. FRP selection

In order to select an FRP laminate for the URM

assemblages, an equivalent-stiffness-based approach

[12] was employed. The laminate required was equated

to the minimum steel reinforcement ratio of 0.2% (based

on the gross cross-sectional area of the wall) accordingto the requirement of the Masonry Standards Joint

Committee [13]. This minimum steel ratio is required

in high seismic zones to be distributed between the ver-

tical and horizontal directions in masonry walls. In

other words, the required thickness of FRP laminate

was determined based on the premise that, the stiffness

of the FRP laminate must be at least equal to or greater

than the axial stiffness of the reinforcement in the walls.The required thickness of the FRP laminate was there-

fore calculated by direct scaling of the reinforcement

area by the ratio of the elastic moduli of the steel and

FRP material as follows,

AFRP ¼Es

EFRP

As ð1Þ

where AFRP is the cross-sectional area of the FRP lam-

inate per unit length of wall, Es is Young�s modulus of

steel, EFRP is the modulus of elasticity of the FRP lam-

inate, and As is the cross-sectional area of the steel rein-

forcement per unit length of the wall. The numerical

procedure required to select the FRP laminate is given

elsewhere [14].

3.2. Material properties

The one-third-scale true-model blocks [15] used in

this investigation were replicas of the standard, full-scale

150 mm wide hollow concrete masonry units [16]. The

average net-area-based compressive strength of the

blocks was 27.87 MPa. The masonry assemblages were

constructed using scaled down mortar joints with anominal thickness of 3.2 mm. To simulate actual con-

struction practice, the mortar mix was designed as Type

S mortar [17] and all mortar joints were tooled to a con-

cave profile. The selected FRP was a bi-directional 0�/

90� Glass-FRP with 0.295 kg/m2 of E-glass fibers. The

properties of the GFRP composites, given in Table 1,

were determined according to ASTM D-3039 specifica-

tion [18] and were supplied by the manufacturer. How-ever, an average strength of 260 MPa (84% of the

specified strength in the GFRP data sheet) with 8.0%

C.O.V. was determined by testing five specimens accord-

ing to the ASTM D-3039 [18]. The steel used for the steel

frame sections was of A 36 grade (yield strength

243 MPa) for the Weak, W-Series, frames, and A572-

50 (yield strength 379 MPa) for the Strong, S-Series,

of frames.

4. Phase I testing: assemblages

4.1. Test setup and instrumentation

The test specimens were chosen to represent typical

loading cases of masonry infill walls as shown in Fig.

2c. A total of 24 1/3-scale specimens were constructed

in the laboratory and tested to failure under displace-

ment controlled loading. The overall displacement was

measured using Linear Variable Differential Transduc-ers (LVDTs) connected to a PC data acquisition system,

which also recorded the applied load on the specimens.

Fig. 3. Assemblages test setup and LVDTs configurations: (a) joint

shear and (b) diagonal tension.

Table 2

Phase I test results

Specimen Results

Test

number

Strength

Individual

(MPa)

Average

(MPa)

C.O.V.

(%)

90U 1 8.45 7.55 18.5

2 8.25

3 5.94

90R 1 13.10

2 11.75 12.17 6.6

3 11.67

00U 1 5.56

2 8.12 7.22 19.9

3 7.97

00R 1 11.49

2 11.27 11.76 5.6

3 12.51

DTU 1 0.85 0.88 12.5

2 0.79

3 1.00

DTR 1 3.73 4.01 6.3

2 4.23

3 4.07

JSU 1 0.97 0.82 18.9

2 0.83

3 0.66

JSR 1 6.28 6.53 6.4

2 7.02

3 6.30

Examples: DTU2 is the second, unretrofitted assemblage tested under

diagonal tension; 90R3 is the third, retrofitted assemblage tested under

axial compression normal to the bed joint.


Typical LVDTs configurations are shown in Fig. 3 for

the diagonal tension and the joint shear specimens.

4.2. Preparation of test specimens

The two axial compression assemblages were of sim-

ilar dimensions in order to permit direct comparison of

their failure loads. Since it was not feasible to cut the 0�assemblages from a built URM walls, the individual

blocks for each assemblage were initially cut to shape

using a diamond saw. The head mortar joint between

the two middle blocks in the joint shear specimens was

left unfilled to allow for the specimen to fail in shear.

All specimens were constructed with face shell mortar

bedding. After air curing for at least 28 days, half of

the constructed specimens were retrofitted using two lay-ers of FRP laminates, one on each surface of the

specimens.

Before applying the FRP laminate, specimen surfaces

were first cleaned from mortar protrusions and dust

using a wire brush and air blasting, respectively. The

epoxy mixture was then applied using a paint roller to

both surfaces of the specimen. The pre-cut fabrics were

then placed on the wet surfaces and more epoxy was ap-plied to insure complete fabric saturation. The assem-

blages were tested in accordance with the ASTM E

477-92 [19] and ASTM E 519-81 [20] specifications.

5. Phase I test results

The test results are summarized in Table 2, and dis-cussed in the following sections in terms of failure

modes, strengths, and deformation characteristics. Each

specimen series was assigned a name according to the

notation shown with the examples at the bottom of

Table 2. The first two characters refer to the axial com-

pression (90� or 00�), diagonal tension (DT), or joint

shear (JS) assemblages. The third character is assigned

one of two letters, either ‘‘U’’ or ‘‘R,’’ indicating

whether the assemblage was Unretrofitted or Retrofit-

ted, respectively.

5.1. Axial compression

5.1.1. Failure modes

The unretrofitted axial compression assemblages 90U

(see Fig. 4a) and 00U (see Fig. 4b) assemblages exhibited

typical compression failure modes characterized by ver-

tical splitting along the webs of the two middle blocks

[21]. The splitting cracks left the two face shells to

deform individually, as shown in Fig. 4c, with a high

slenderness ratio. Finally, the specimens totally disinte-

grated as a result of the out-of-plane buckling and/orspalling of the face shells or a combination of both.

Noticeably, all failure modes were brittle and the assem-

blages disintegrated almost immediately after reaching

their respective maximum loads.

In contrast, all the retrofitted assemblages exhibited

one failure mode initiated by vertical splitting of the

Fig. 4. Failure modes of the unretrofitted axial compression assemblages: (a) Series 90U, (b) Series 00U and (c) web splitting mechanism of face shell

mortar bedded masonry [21].

Fig. 5. Failure modes of the retrofitted axial compression assemblages:

(a) Series 90R and (b) Series 00R.

Strain(a)

Stre

ss (M

Pa)


interior webs followed by a gradual increase in the load

up to the peak load. After reaching the peak load, a sud-

den bang was heard as a result of the blocks webs com-

pletely breaking off the face shells. The specimens

continued to carry more load under the displacement

controlled loading with a gradual decrease in capacity.All the retrofitted assemblages were reduced to two in-

tact face shells with all interior webs damaged (see

Fig. 5).

00U2

0.00250.0

0.0000 0.0005 0.00150.0010 0.0020Strain

(b)

4.0

6.0

8.0

10.0

12.0

2.0

Stre

ss (M

Pa)

00R2

00R3

00U1

00U3

00R1

0.00350.0030

Fig. 6. Stress–strain relationships for the axial compression assem-

blages: (a) Series 90U and 90R, and (b) Series 00U and 00R.

5.1.2. Strength characteristics

Table 2 gives the variation of the compressive

strengths of the unretrofitted versus retrofitted assem-

blages. To facilitate comparison, the compressivestrengths of the assemblages were calculated as the max-

imum load-carrying capacity divided by the gross

assemblage area perpendicular to the direction of the

applied load (6431.0 mm2). It is clear that the values

of the coefficients of variation for the retrofitted assem-

blages are generally lower than those of the unretrofitted

specimens (see Table 2). This demonstrates the lami-

nate�s role in reducing the inevitable variations inURM construction.

In general, the retrofitted assemblages did not lose

all their strength nor disintegrated upon reaching the

maximum strength. In fact, in the majority of the

tests, a plateau region was attained during which

the compressive stress almost stabilized and began togradually decrease with increased displacement. Such

plateau can be regarded as residual strength after fail-

ure of the assemblages, an absent feature in the case

of URM.


5.1.3. Deformation characteristics

The measured displacement and applied load were

used to generate the stress–strain relationships shown

in Fig. 6. In general, a good agreement can be observed

between the initial slopes for the prisms tested with

h = 0� and 90� in both the retrofitted and the unretrofit-ted series.

5.2. Diagonal tension


All the unretrofitted diagonal tension assemblages

exhibited shear slip failure along the mortar bed joints.

This is attributed to the mortar joint weak bond strengthcompared to the tensile strength of the concrete blocks.

The observed shear slip failure mode was highly brittle

and, as soon as shear slip along a bed joint was initiated,

the assemblages split into two parts and subsequently

Fig. 7. Failure modes of the diagonal tension assemblages: (a) specimen DTU

DTR web splitting.

disintegrated. In two out of the three assemblages

(DTU1 and DTU2) the shear slip occurred along the

middle bed joints (see Fig. 7a), while the third specimen

(DTU3) failed along the first bed joint (see Fig. 7b). No

signs of cracking or distress were observed prior to fail-

ure by shear slip.For the retrofitting assemblages, using the FRP lam-

inates effectively prevented any tension or shear failure

modes. All three retrofitted assemblages failed by local

crushing of their corners contained within the steel load-

ing shoes. Fig. 7c illustrates the local crushing failures.

Due to the compressive stress buildup at the toes, verti-

cal cracking through the webs was observed and ex-

tended into the first two courses as shown in Fig. 7d.Other than the local crushing and the minor delamina-

tion at the vicinity of the loading shoes, no other

signs of distress or cracking were observed in the

assemblages.

2, (b) specimen DTU3, (c) Series DTR corner crushing and (d) Series



The maximum stresses sustained by the assemblages

are given in Table 2. In accordance with ASTM E

519-81 [20] standard specification, the diagonal tensile

or shear strength is calculated from,

Ss ¼0:707P

An

ð2Þ

where Ss is the shear stress based on the net area, P is theapplied load, and An is net area of the specimen calcu-

lated as follows:

An ¼wþ h

2

� �tn ð3Þ

where w, h and t are the specimen width, height and

thickness respectively, and n is the percentage of solid

in the unit, expressed as a decimal.

The average strength of the retrofitted diagonal ten-

sion specimens was 4.58 times that of the unretrofittedones and, as expected, the relatively low coefficient of

variation is indicative of the role of the laminates in

reducing the anisotropy and variability of URM.


In accordance with ASTM E 519-81 [20] specifica-

tion, the shear strain was calculated using the vertical

shortening along the compression diagonal and the hor-izontal extension along the tension diagonal as follows:

c ¼ DV þ DHg

ð4Þ

where c is the shear strain, DV is the vertical shortening,

DH is the horizontal extension, and g is vertical gage.

Fig. 8 illustrates the shear stress versus shear strain

relationship for the diagonal tension assemblages. Due

to the sudden brittle failure mode, obtaining post-peak

behavior for the unretrofitted assemblages was not

feasible.

For the retrofitted specimens, the LVDTs installed atthe center of the assemblages at the 102 mm gage length,

did not record any appreciable deformations in any of

the tests. Furthermore, as the load approached its peak,

the long compression LVDT brackets on assemblages

0.0025

DTU3

2.0

Shea

r Str

ess

(MPa

)

0.00050.0000

0.50.0

1.01.5

DTU1

4.5

3.02.5

3.54.0

0.00150.0010

DTU2

0.0020Shear Strain

DTR3DTR2

DTR1

0.00350.0030

Fig. 8. The shear stress versus shear strain relationship for the

diagonal tension assemblages.

DTR2 and DTR3 detached due to local delamination

and crushing at the vicinity of the bottom loading shoe.

The same occurred in assemblage DTR1 towards the

end of the test, yet its shear stress versus strain curve

showed a load plateau with a slight increase in load.

In fact, all the assemblages were able to sustain residualloads under increasing displacement (even though the

resulting load plateaus could not be plotted for DTR2

and DTR3).

5.3. Joint shear


The unretrofitted specimen failed in a brittle shearslip debonding mode at very low load and displacement

levels. This is a result of the weak mortar joint bond

strength and the absence of friction resistance due to

the lack of compressive stresses normal to the mortar

bed joints. The failure was in the form of complete sep-

aration in the top and/or bottom mortar joints vicinity

(Fig. 9a). This failure mode is highly brittle and oc-

curred without much time elapse between the crack ini-tiation at the block–mortar interface and the

consequential debonding of the blocks.

In a manner similar to the retrofitted axial compres-

sion assemblages, none of the retrofitted assemblages

failed by shear slip along the block–mortar interface.

At most, minor signs of delamination and stretching

of the laminate were observed along a portion of the

bed joints. All assemblages ultimately failed after oneof the middle blocks cracked through the webs and split

open, though remaining attached to the assemblage as

shown in Fig. 9b. The splitting of the middle blocks is

attributed to the induced lateral tensile stresses devel-

oped in the laminates, which resisted the closing of the

head joint gap between the top and bottom blocks.

The tensile stresses induced in the middle block�s face

shells eventually resulted in cracking through the webs.This failure might also be attributed to the fact that,

with the presence of the laminates and their ability to

transfer shear stresses to the middle blocks, the blocks

were subjected to a state of stress similar to that occur-

ring in the specimens under compression parallel to the

bed joint, and thus the observed failure mode was devel-

oped. Nevertheless, this failure mode implies that shear

failure can be eliminated and the wall strength would begoverned by the compressive strength of the composite

prisms.


The strengths of the three unretrofitted joint shear

assemblages are presented in Table 2. In order to deter-

mine the joint shear strengths shown in the table, the

failure load for each assemblage was divided by therespective net-mortared area. Since face shell mortar

bedding was employed, the actual lengths of the

Fig. 9. Failure modes of the joint shear assemblages: (a) Series JSU and (b) Series JSR.

Average Slip (mm)0.015

JSU1

0.005

0.2

She

ar S

tres

s (M

Pa)

0.6

1.0

0.8

0.4

0.0000.0

JSU2

0.010

JSU3

0.020

Average Slip (mm)

Sh

ear

Str

ess

(MP

a)

0.00

9.0

8.0

7.0

6.0

5.0

4.0

3.0

2.0

1.0

0.00.10 0.20 0.400.30 0.50 0.60 0.70

Fig. 10. The shear stress versus slip relationship for the joint shear

assemblages.


mortared bed joints less the head joint gap were mea-

sured using a caliper and multiplied by the average min-

imum bottom face shell thickness of 8.7 mm to

determine the net joint shear area.

The FRP laminates were cut precisely and adhered to

cover only the lengths of the mortared bed joints ensur-ing that the head joint gap between the middle blocks

was not obstructed. Therefore, similar to the unretrofit-

ted specimens, the shear area of the retrofitted assem-

blages is the net-mortared area of the bed joints

(excluding the head joint gap area). The increase in the

joint shear strength for the retrofitted assemblages was

8.2 times that of the unretrofitted ones.


Fig. 10 illustrates the shear stress versus shear slip

behavior of the joint shear assemblages. The unretrofit-
ted assemblages exhibited dramatic load drop after
reaching their respective maximum stress at an average


slip of 0.011 mm, with a coefficient of variation of

29.0%.

In general, as soon as the retrofitted assemblages

reached their respective maximum loads, a sudden load

drop occurred which signified the web splitting of one of

the middle blocks as discussed above. This was followedby a plateau with a gradually descending strength; the

assemblage at this stage was still able to carry more load

than the maximum load reached by the unretrofitted

specimens. Examination of the failed assemblages re-

vealed that the mortar bond between the top/bottom

and middle blocks was damaged and that the laminates

were entirely transferring the vertical applied load from

the top to the bottom block through the middle blocks.In all the retrofitted assemblages, the laminate was not

entirely torn and the assemblage could have resisted fur-

ther loads.

The average slip at the maximum shear stress was

0.379 mm within a coefficient of variation of 17.8%. This

is in excess of 34 times that of the unretrofitted joint

shear assemblages, thus indicative of the significant

deformation capability gained by using the FRPlaminate.

Beam W6x15

Beam W6x15

Column W6x

158 blocks wide

16co

urses

high

L6x6x3/8 Angle

L6x6x3/8 Angle

Column W

6x15

Reaction “Strong” Floor

Out-of-plane bracingplates held via C-clampsto the support angles

Load Cell

Top Loading Shoe

Rea

ctio

n C

olum

nW

14x9

0

AMSLERLoading Jack

Spherically-seatedActuator Head

AdjustableStilt

30.0

4"

Rea

ctio

nFr

ame

Brac

ing

Actuator and Reaction Frame Bracing connected to Top Girder

T ST SETUP FORCM & CR FRAMESDrawing Scale 1:20

6.00"

78.49"1.99 m

(a)

Fig. 11. Masonry infilled steel frame specime

6. Phase II testing: infilled frames

This phase focused on testing of one-third scale, mo-

ment-resisting, structural steel frames infilled with unret-

rofitted and retrofitted hollow block masonry walls.

Single-story/single-bay infilled frame subjected to diago-nal-compressive loading (as shown in Fig. 11a) were

used to evaluate the behavior of the composite frame–

wall system.

The interaction between the infill wall and the sur-

rounding columns and beams result in unequal contact

lengths along the boundaries of the infill with each of

the weak and strong frame members. This, in turn, re-

sults in different infill contribution with different frames.Table 3 lists the structural properties of the weak and

strong frame sections [22] and outlines the five frames

tested in Phase II.

One-third scaling of the typical clear floor height and

column span of the prototype structure was used to ob-

tain the dimensions of the model infilled frame [15]. A

clear height between the beams and columns of

1100 mm was used. This is equivalent to a full-scaledimension of 3.3 m. The masonry infill walls were built

LOCATIONS OF STRAIN GAGES AND LVDTs

Beam

BeamColu

mn

Column

MC

MC

MTMT LTLT ST ST

LC

LC

SC

SC

TEN

SIO

ND

IAG

ON

AL

TEN

SIO

ND

IAG

ON

AL

COMPRESSIONDIAGONAL

COMPRESSIONDIAGONAL

KEYMC = Main Compression LVDTMT = Main Tension LVDTLC = Long Compression LVDTLT = Long Tension LVDTSC = Short Compression LVDTST = Short Tension LVDT

= Strain gages (adhered tothe top of the upper andlower flanges)

Infill h

eight=

Infill l

ength

==

43.50

0"

hl

= Sectio

n

Depth

= d

d

l/2

6.000

"

Applied Load

(b)

ns: (a) test setup, (b) instrumentation.

Table 3

Structural properties and test matrix of the steel frames

Structural property Weak (W)

frame S3 · 5.7

Strong (S)

frame W6 · 15

Area 1077 mm2 2858 mm2

Depth 76 mm 152 mm

Web thickness 4 mm 6 mm

Flange width 59 mm 152 mm

Flange thickness 7 mm 7 mm

Strong axis (X-axis)

Elastic moment of inertia 1,049,000 mm4 12,112,000 mm4

Elastic section modulus 27,500 mm3 159,300 mm3

Plastic section modulus 31,950 mm3 176,980 mm3

Weak axis (Y-axis)

Elastic moment of inertia 189,400 mm4 3,879,300 mm4

Elastic section modulus 6390 mm3 50,960 mm3

Plastic section modulus 10,700 mm3 77,840 mm3

Infill type Specimen

Bare WB –

Unretrofitted WU SU

Retrofitted WR SR

Steel frame section: ‘‘W’’ weak frame S3 · 5.7 or ‘‘S’’ strong frame

W6 · 15.

Infill type: ‘‘B’’ bare (no infill), ‘‘U’’ unretrofitted-masonry infill, or

‘‘R’’ retrofitted-masonry infill.

Fig. 12. Load–deflection relationship for frame WB.


eight blocks wide by 16 courses high. To ensure symme-

try in the construction, mortar was packed along all the

boundaries between the infill and the confining steel

frame with a nominal mortar joint thickness of 3 mm.

To identify the different frames tested in Phase II,

each specimen was assigned a name according to the

notation in the bottom of Table 3. The first character

is used to identify the steel section of the boundingframe whether S3 · 5.7 (Weak frame) or W6 · 15

(Strong frame). The second character describes the type

of infill, if any, ‘‘B’’ refers to no wall (Bare frame), ‘‘U’’

or ‘‘R,’’ indicating whether the wall was Unretrofitted or

Retrofitted, respectively.

For the stronger S-frames, the clear height between

the beams and the clear span between the columns

were similar to those of the W-frames. Moreover, thebeam-column connections were also designed as full-

moment-resisting and fabricated using complete-joint-

penetration groove welds to weld the beam flanges to

the column flanges while 3.0 mm fillet welds were used

to weld the beam webs to the column flanges. However,

due to the expected high diagonal-compressive loading

force, 10 mm thick stiffener plates were welded using

3.0 mm fillet welds between the column flanges in orderto prevent premature web buckling at the loaded ends.

To maintain symmetry, similar plates were also welded

at the four corners of the S-frames.

6.1. Test setup and instrumentation

The general test setup and loading assembly are

shown in Fig. 11a. The compressive top load was

applied using an AMSLER hydraulic jack with a load

capacity of 490 kN and a maximum stroke of 125 mm.

A lateral bracing system consisting of four L 6 · 6 · 3/

8 angles were used to prevent accidental out-of-plane

deformations during the in-plane loading of the frames.

Fig. 11b illustrates the typical instrumentation in-stalled on the infill wall and bounding frame, as well

as the locations of the critical sections along the steel

frame where strain gages were placed (an infilled

W6 · 15 steel frame is shown in the figure for illustra-

tion). All LVDTs, strain gages, and the load cell used

to measure the applied compressive load were all con-

nected to a PC data acquisition system.

7. Phase II test results

7.1. WB frame

As expected, the frame joints underwent severe rota-

tion and distortion. In addition, both joints along the

tension diagonal experienced tearing of the webs asshown in Fig. 12, although no signs of cracking were

noted in the welds between the columns and the beams

at these moment-resisting connections (thus indicative

of the strength of the weld). It was evident from the dis-

torted shape of the frame and the bent columns and

beams, that the permanent (i.e., plastic) deformations

propagated from the joints inwards.

The initial stiffness obtained from the load–deflectioncurve shown in Fig. 12 was determined to be 2.2 kN/mm.

A linear behavior was observed up to a load of a 25.8 kN

corresponding to an average top displacement of

11.9 mm. Subsequently, a load plateau of 27.0 kN was

attained at a top displacement of 14.0 mm. Attributed

to the strain hardening effects, a slight increase in load

occurred resulting in an ultimate load of 28.3 kN. The

actual load–deflection behavior closely resembles the ex-pected behavior in which an initial linear response occurs

until initiation of yielding followed by a plateau then a

slight load increase to attain the ultimate failure load.

7.2. WU frame

The maximum load-carrying capacity attained by the

infilled frame was 104.0 kN. A stepped diagonal crack

Deflection (mm)

00.0 20.0

Load

(kN

)

100

WB40

20

80

60

WU

40.0 60.0 80.0

Applied Load

Applied Load

Fig. 13. Load–deflection relationship for frames WU and WB.


was observed at the center of the infill panel along the

compression diagonal as shown in Fig. 13, was an-

nounced by an audible bang and occurred at a load of

7.8 kN corresponding to 7.5% of the maximum attained

load. The load–deflection relation obtained for frame

WU is shown in Fig. 13. The load versus deflection

curve of the bare frame tested earlier is reproduced on

the same plot for comparison.Initially, the load–deflection curve was characterized

by a steady rise as shown in Fig. 13. Prior to attaining

the first peak at 97.7 kN at a corresponding deflection

of 1.9 mm and shortly afterwards, cracking noises were

continuously heard although no visible cracks were ob-

served, thus indicating possible damage in the interior

webs of the masonry infill panel resulting in a series of

quick load descents and ascents. Shortly afterwards, amajor off-diagonal crack parallel to the initial toothed

crack at the center of the infill panel was observed. This

crack occurred at a trough in the load–deflection curve

at approximately 5.1 mm. Unlike the initial toothed

crack, this crack propagated vertically through the

Fig. 14. Damage of frame WU: (a) at the left column, (b) at the wa

blocks and mortar joints. As the infill panel further

readjusted indicating redistribution of the transferred

load, the load steadily rose to reach the maximum load

carried by this infilled frame in spite of the occurrence of

local cracking due to crushing of the infill in the vicinity

of the bottom loading shoe at approximately 7.6 mm. At12.7 mm, crushing and cracking was observed in the in-

fill panel near the top loading shoe. Subsequently, a

steady decline in the load-carrying capacity of the frame

occurred as more off-diagonal cracks started appearing

on both sides of the first toothed crack in addition to

spalling of the block face shells near these cracks. At

approximately 15.0 mm, cracks along the infill bed

joints were observed and propagated at mid-length ofthe frame columns until a significant portion of the infill

panel face shell near the mid-length of the left frame col-

umn separated (approximately at 28.0 mm) as shown in

Fig. 14a. Face shell spalling continued rapidly (at

33.0 mm) resulting in severe deterioration and cracking

in the infill panel until, finally (at 56.0 mm), was attained

signifying the beginning of a load plateau. In turn, this

marks the point after which the infill panel ceased tocontribute in resisting the applied load. Extensive local

crushing and cracking at the toe of the infill panel near

the bottom loading shoe is shown in Fig. 14b. The sever-

ity of the damage at the center of the masonry infill is

illustrated in Fig. 14c. From 56.0 mm onwards, the steel

frame was entirely carrying the applied load and was de-

formed severely as shown in Fig. 14d. Similar to the WB

frame, the joints of the steel frame underwent severeplastic rotation at the joints along the tension diagonal,

ll toe, (c) at the infill center and (d) at the beam-column joint.


in addition to local web buckling in the columns and

beams at the loaded joints.

The infilled WU frame attained a maximum load-car-

rying capacity of 104.0 kN which represents an increase

of 267.5% compared to the capacity of the bare WB

frame. The initial stiffness of the WU frame measuredas the secant stiffness at 50% of the maximum load-car-

rying capacity was determined as 55.7 kN/mm which is

25 times that of the bare frame. The significant increase

in WU frame�s load-carrying capacity and initial stiff-

ness compared to the WB frame was expected, particu-

larly in view of the fact that the infill panel is

relatively stiff compared to the frame. Ultimately, upon

attaining a load plateau at a displacement of 57.4 mmsignifying the end of the infill�s participation in in-plane

load resistance, the plastic load-carrying capacity of the

WU frame was determined from the test as 28.2 kN.

This is comparable to the bare frame�s ultimate load

capacity obtained previously from the WB frame test.

7.3. WR frame

The load versus deflection curves for the WR frame is

shown in Fig. 15a. The maximum load attained by the

frame was 218.9 kN which represents increases of 7.7

times and 2.2 times the maximum loads attained by

the WB and WU frames respectively. At an applied load

of 182.4 kN corresponding to 83.3% of the maximum

Fig. 15. Frame WR: (a) load–deflection relationship, (b)

load-carrying capacity, some hairline cracks were ob-

served in the blocks near the vicinity of the top loading

shoe. These cracks were visible underneath the clear

laminate adhered on the exterior of the masonry infill

panel. As the load began to decrease, cracking noises

and clicks were heard until suddenly at a load of175.1 kN, corresponding to 80.0% of the ultimate load

on the descending branch of the load curve, the interior

webs near the top portion of the infill panel were dam-

aged causing the separated retrofitted face shells near

the top loading shoe to snap outwards and moved out-

side the flanges of the frame members (Fig. 16a). Unfor-

tunately, shortly before and after this outward ‘‘burst,’’

the buckled face shell brushed against the main and infillcompression LVDTs, thus preventing further recording

of displacement.

Minor signs of delamination along the second bed

joint on the backside of the infill panel were evident.

There were no signs of distress in the steel frame, clearly

indicating that the applied load was primarily endured

by the retrofitted infill panel with minimal contribution

from the steel frame.A thorough understanding of the behavior and re-

sponse of the retrofitted infill panel was further facili-

tated upon its removal from the bounding steel frame

thereby enabling a closer inspection. Fig. 16b shows

the wall separation in the left side of the frame. The ret-

rofit technique using FRP laminates was very successful

delamination zone and (c) damaged webs region.


in preserving the integrity of the highly brittle masonry.

The fact that the panel, simulating a story-high wall, was

removed in one intact piece (in spite of some damage to

the interior webs) is testimony to the beneficial effect of

retrofit with FRP overlay. At the toes of the infill panel

within the vicinity of the loaded corners of the frame, allinterior webs were damaged. As shown in Figs. 15c and

16c, the web damage extended inwards towards the cen-

ter of the panel to approximately one-quarter of the

diagonal length. Web damage was also evident along

the perimeter of the panel which was in contact with

the frame members. The web damage was minimal

near the corners of the tension diagonal.

However, the separated face shells were held intactby the strong laminate and, in general, there was min-

imal (if any) delamination between the overlay and

the block face shells (except in the location shown

in Fig. 15b). No web damage was evident at the cen-

ter of the infill. Furthermore, unlike in the unretrofit-

ted-masonry infill wall of the WU frame, the laminate

successfully prevented the occurrence of any diagonal

Fig. 16. Damage of frame WR: (a) out-of-plane wall burst, (

tension cracks or shear slip along the bed joints in

the infill. As shown in the various assemblage tests

discussed in Phase I of the experimental program,

the FRP laminate basically suppressed any tension

and shear failure modes in the masonry by reinforcing

the weak mortar joints. Thus, the resulting retrofitted-masonry infill wall has been transformed into two

very strong face shells connected by masonry webs

which are considered as the weak elements in the

assembly.

The secant stiffness at 50% of the maximum load, the

initial stiffness of the BR frame is 131.4 kN/mm which

represents increases of 58.7 times and 2.4 times the ini-

tial stiffness values of the WB and WU frames respec-tively. The peak load was reached at a compressive

deflection of 5.6 mm. Similar to the WU frame, the re-

corded deflections along the compression diagonal in

the direction of the applied load were greater than those

along the tension diagonal. This is attributed to the local

cracking at the infill�s loaded toes resulting in a reduced

stiffness along the loading direction.

b) infill wall separation and (c) extent of web splitting.


Ultimately, as the infill panel was no longer in any

effective contact with the bounding steel frame, a load

plateau was attained which represents the bare frame�splastic load capacity. The load stabilized at a value of

28.9 kN which is comparable to the WB frame capacity.

7.4. SU frame

Plot of the applied top load versus deflections for the

SU frame is shown in Fig. 17. The initial secant stiffness

of the SU frame was 91.4 kN/mm. The ultimate load-

carrying capacity of the SU frame determined in the sec-

ond test was 284.4 kN. The unretrofitted-masonry infill

panel remained crack-free up to an applied diagonalload of 122.4 kN corresponding to 43.0% of the ultimate

load-carrying capacity of the SU frame. Thereafter, a

longitudinal crack at the middle of the panel occurred

similar to the WU frame. However, unlike the toothed

crack in the WU frame, which propagated along the

head and bed joints around the masonry units, the mid-

dle crack in the SU frame�s panel extended through both

the units and the mortar joints. As loading progressed,the middle crack extended further in addition to the for-

mation of some off-diagonal hairline cracks and a short

bed joint crack above the first masonry course. Near the

peak load, signs of crushing of the boundary mortar

joint between the steel frame and the infill panel in the

vicinity of the loaded corners were observed. Moreover,

a hairline separation crack between the panel and the

frame at the tension corners was observed to extendapproximately three courses long.

The cracking pattern of the masonry infill wall resem-

bles that encountered in the WU frame test in which a

central crack is first initiated along the loaded diagonal

of the wall followed by the formation of some off-diag-

onal cracks. In the second test attempt, the existing but

closed hairline cracks resulting from the first test wid-

ened as the frame reached a first peak load at266.9 kN at a corresponding deflection of 6.8 mm.

Shortly before reaching the peak load, minor cracking

was observed in the infill�s toe near the bottom loading

shoe. As the frame was pushed further in spite of the de-

creased load resistance, small off-diagonal cracks began

forming on the left and right sides of the central crack.

These cracks assisted in the redistribution of the load

0.00 20.00 80.0040.00 60.00Deflection (mm)

150.0

0.050.0

100.0

200.0250.0300.0

Load

(kN

)

Fig. 17. Frame SU load–deflection relationship and diagonal cracking.

within the infill panel as it adjusted to bear against the

deforming shape of the steel frame. Suddenly, face shell

spalling occurred in the tension corner regions (to the

left and right of the central crack where extensive off-

diagonal cracks were occurring) as shown in Fig. 18a.

This occurred at an approximate SU deflection of9.0 mm. In fact, this served as a major indicator of the

shift in load resistance between the infill panel and the

steel frame. As the frame was still being pushed diago-

nally, due to incompatible deformations between the

steel frame and the masonry wall, the infill wall was

quickly losing structural integrity accompanied with ra-

pid face shell spalling and collapse of massive ‘‘chunks’’

of the upper region of the wall (Fig. 18b).At the end, only the lower three courses of the ma-

sonry wall remained standing on the lower beam and

column of the frame as shown in Fig. 18c. As the load

increased, the SU frame was simply behaving as a bare

W6 · 15 moment-resisting frame and yielding com-

menced at the beam-column joints.

The formation of plastic hinges, eventually leading to

a plastic collapse mechanism, characterizes the failuremode of the CM frame in which the masonry infill panel

did not increase its load-carrying capacity. However,

failure of the masonry infill panel is attributed to a com-

bination of toe crushing (characterized by local com-

pressive crushing of the masonry at the vicinity of the

loaded corners) and diagonal-compression failure at

the center of the panel (characterized by the formation

of extensive diagonal and off-diagonal longitudinalcracks). The frame was unloaded after the full stroke

of the loading actuator was consumed.

7.5. SR frame

Fig. 19a shows the load–deflection relationship of the

SR frame which had a maximum load capacity of

343.0 kN and an initial stiffness of 262.7 kN/mm.At 0.8 mm, an audible bang was heard although no

crack was detected visually. Characterized by a small

shift in the load–deflection curve, the bang suggested

the occurrence of a crack in the interior webs, possibly

at one of the loaded toes of the infill wall. At 2.3 mm,

the crack location most probably occurred in the top

toe. At a deflection of 1.98 mm, the load increased fur-

ther until it reached the first peak of 270.1 kN. At thispoint, a greater load bang was heard and crushing at

the top loaded toe of the infill panel was observed. As

in the retrofitted assemblages, crushing at the top toe

of the retrofitted-masonry infill panel involved damage

of the interior webs leading to the laminated-face shells

snapping outwards. The damage extended along the in-

fill-frame boundary for a length of approximately six

courses (three block-lengths) and only one course wide(one half a block-length) as shown in Fig. 19b. A conse-

quential loss in load capacity occurred but gradually

Fig. 18. Damage of frame SU: (a) face shell spalling in the diagonally cracked region, (b) collapse of the upper infill region and (c) infill wall

remnants.

Fig. 19. Frame SR: (a) load–deflection relationship, (b) damage progress and (c) final damaged zones.



increased as the frame was further loaded. Similar crush-

ing also occurred at the bottom loaded toe of the infill

panel (Fig. 19b).

Separation between the infill panel and the steel

frame at the tension diagonal corners occurred as load-

ing progressed. At approximately 22.0 mm, the separa-tion gap was clearly visible as shown in Fig. 20a. As

the frame was pushed further beyond 22.0 mm, the sep-

aration between the frame and the infill increased. The

extent of toe crushing, which is defined as splitting of

the face shell and at times accompanied by minor delam-

ination between the overlay and the face shell itself, also

increased (Fig. 20b and c). The steel frame was consid-

erably deformed with significant plastic rotation at thetension joints as shown in Fig. 20d. Examination of

the infill panel at the end of the test (Fig. 20e) indicated

Fig. 20. Frame SR damage: (a) at left side, (b) at top loaded corner, (c) at b

configuration.

that, other than minor delamination at the infill-frame

boundaries and toe crushing and in spite of the separa-

tion between the infill and the frame at the tension diag-

onal corners, the central region of the wall was intact

without any cracking or damage. A schematic diagram

illustrating the state of the retrofitted infill wall at theend of the test is shown in Fig. 19c.

Even though the load–deflection curves indicated that

a load plateau was reached, it was decided to further

load the SR frame. This decision was triggered by the

fact that the plateau occurred at load of 339.0 kN which

is 21.0% greater than the expected plastic load-carrying

capacity of the bare W6 · 15 frame of 280.4 kN as deter-

mined experimentally from the prior SU frame test, thussuggesting that the retrofitted infill wall still contributed

to load resistance.

ottom loaded corner, (d) at beam-column joint and (e) final damaged


8. Summary of Phase II test results

Beside local toe crushing, secondary signs of distress

resulting from severe face shell splitting such as tearing

of the laminate or minor delamination between the

block face shells and the laminates were the only ob-served damages as the FRP retrofitted infilled-frames

were pushed to severely deformed configuration. The

frames with the retrofitted infill walls depicted similar

behavior throughout the entire loading history. In both

the WR and SR tests, as soon as local crushing occurred

at the wall�s corners, clearly visible and wide separation

gaps between the panel and the frame constantly in-

creased unlike in the WU and SU frames. Unlike theunretrofitted-masonry infill walls which disintegrated

soon after the infilled-frame system reached its peak

load, the retrofitted infill walls remained supported with-

in the bounding steel frame until the end of the loading

and even after attainment of load plateau which signaled

that the frames reached the plastic load-carrying capac-

ity of the bare steel frame. This behavior demonstrates

the superior contribution of FRP laminates in alteringthe brittle hazardous behavior of URM infill walls to

a ductile and damage-tolerant wall with apparent post-

peak capacity and energy dissipation capabilities.

Table 4 summarizes the maximum diagonal-compres-

sive load sustained by the five tested frames. In addition,

the plastic load capacity of the bare W6 · 15 which was

experimentally determined through testing the SU frame

after the remains of the infill panel were removed (thethird test of the SU frame), is also included in the table.

The percentage and the corresponding multiple in-

creases in the load-carrying capacity compared to that

of the bare frame and the unretrofitted-masonry infilled

frame for each of the two steel frame types are also cal-

culated and presented in Table 4.

Table 4 also compares the initial secant stiffness of the

five tested frames in this study. The stiffness values werecalculated as the slope of line joining the origin and the

Table 4

Phase II test results

Frame Maximum

load (kN)

% Increase compared to X�s increase c

Bare frame Unretrofitted

infilled frame

Bare frame

W-Frames

(S3 · 5.7)

WB 28.3

WU 104.0 267.5% 3.67

WR 219.0 673.9% 110.6% 7.74

S-Frames

(W6 · 15)

SB 280.4b

SU 284.4 1.4% 1.01

SR 343.1 22.4% 20.6% 1.22

a Initial secant stiffness determined from the load–deflection curve of the te

50% of the maximum load-capacity.b Plastic load-carrying capacity of the bare W6 · 15 frame was experimenta

the infill wall were removed entirely.c Initial secant stiffness of the bare W6 · 15 frame was experimentally dete

point at 50% of the ultimate load using the applied diag-

onal load versus in-line compressive displacement

curves. Although a bare W6 · 15 steel frame was not

tested as a separate frame, its stiffness which is shown

in Table 4 was experimentally determined from the sec-

ond test of the SR frame whose infill panel was sepa-rated from the frame along the majority of its

perimeter and sustained local damage at its loaded toes

whereas the steel frame did not experience any distress in

the prior test. For each frame within the two main steel

frame types, the increases in stiffness compared to the

bare frame and the first test of the unretrofitted-masonry

infilled frame are computed and presented in the table.

9. Conclusions

This paper presents an experimental investigation on

the retrofitting of concrete masonry infill walls using

FRP laminates, which provides a strengthening alterna-

tive for URM infill walls. The relative ease with which

FRP laminates can be installed on the walls makes thisform of strengthening attractive to the owner, consider-

ing both reduced installation cost and down time of the

occupied structure. Another reason is to comply with

new seismic codes requirements without the need to

demolish the whole wall and rebuild it. The following

conclusions resulted from Phase I of the investigation:

1. The laminates significantly increased the load-carry-ing capacity of the masonry assemblages exhibiting

shear failures along the mortar joints (joint shear

and diagonal tension). The average joint shear

strength of the retrofitted specimens was equal to

eight times that of their unretrofitted counterparts.

2. The unretrofitted axial compression assemblages

failed suddenly and disintegrated totally upon reach-

ing peak stress. However, the FRP supplied the ten-sile strength required to stabilize the out-of-plane

ompared to Initial secant

stiffnessa (kN/mm)

X�s Increase Compared to

Unretrofitted

infilled frame

Bare frame Unretrofitted-

masonry

infilled frame

2.2

55.7 24.92

2.11 131.4 58.74 2.36

20.9c

91.4 4.38

1.21 175.1 8.40 1.92

sted frames as the slope of the line joining the origin and the point at

lly determined from the third test of the SU frame after the remains of

rmined from retesting the SR frame.


buckling of the individual face shells, thus preventing

brittle failure after webs splitting and allowing the

specimen to carry more loads. This prevented cata-

strophic failure of the masonry–FRP composite

assemblages compared to their URM counterparts.

3. The FRP laminates resulted in a gradual prolongedfailure, and a stable wall with noticeable structurally

integrity and residual strength even after failure.

Thus, the long known hazard problem associated

with URM can be eliminated using the proposed ret-

rofit technique. In seismic zones, the prevention of

the brittle failure mode is highly desirable since it pro-

vides a means for energy dissipation and conse-

quently reduces the seismic forces on the framestructure.

The following conclusions resulted from Phase II of

the investigation:

4. Retrofitting the infill panel with externally, epoxy-

bonded FRP laminates resulted in an increase in

load-carrying capacity of 2.1 and 1.2 times that ofthe corresponding unretrofitted-masonry infilled

frames for the W-frames and the S-frames

respectively.

5. Even in the S-frames whose load capacity was not sig-

nificantly increased due to retrofit of the infill panel,

the laminates were able to completely alter the defor-

mation characteristics and behavior of the wall itself.

In the unretrofitted-masonry infilled frames, the wallswere completely destroyed and the blocks fell out-of-

plane which in real life poses a hazard to buildings�occupants. The failure mode of the two unretrofitted

frames was characterized as corner-crushing and

diagonal-compression respectively. In the retrofit-

ted-masonry infilled frames, no signs of diagonal

cracking were observed and both frame types failed

due to local crushing at the loaded corners. Examina-tion of the retrofitted panel indicated that the central

region remained intact and that the majority of the

damage occurred at the outermost perimeter and at

the loaded corners where the inner webs of the blocks

cracked resulting in the formation of separate lami-

nated-face shells.

6. The retrofitting technique maintained the walls struc-

tural integrity and prevented collapse and debris fall-out. The FRP laminates contained and localized the

damage of the URM walls even after ultimate failure

and no signs of distress were evident throughout the

wall except at the vicinity of the corners and around

the openings. In contrast to the URM walls, the

strengthened walls were stable after failure. In a real

building, this can reduce the seismic hazard associ-

ated with the wall tipping off or falling out of theframe, and eliminate injuries or loss of lives and

properties due to the wall collapse. This would also

maintain the wall�s structural integrity and would

reduce the possibility of URM walls collapsing and

spalling, which, in itself, is a major source of hazard

during earthquakes, even if the whole structure

remained safe and functioning.

7. The masonry–FRP composite walls do not fail cata-strophically as their URM counterparts. The FRP

laminates resulted in a gradual prolonged failure

and a stronger wall with more energy dissipation

and apparent post-peak strength. This should result

in a higher response modification factor than that

typically selected for the analysis of URM structures.

8. By supplying the shear strength at the mortar joints,

FRP laminates can serve as external reinforcementfor unreinforced or under-reinforced masonry walls,

thus providing a quick and cost-effective solution to

conform to the more restrict emerging seismic codes

requirements.

Acknowledgments

The work presented herein was supported under

Grant No. CMS-9730646 from the National Science

Foundation (NSF). The results, opinions, and conclu-

sions expressed in this paper are solely those of the

authors and do not necessarily reflect those of the

NSF. The authors would like to gratefully acknowledgeassistance of Edward Fyfe, Peter Milligan and Sarah

Cruickshank, Fyfe Co. LLC, California, for providing

the FRP, and John Sabia, D.M. Sabia Co., Pennsylva-

nia for providing the mason. The authors would also

like to thank Mr. Omar El-Dakhakhni for his

assistance.

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Documents

Hazard mitigation and strengthening of unreinforced masonry walls using composites