1

DUCTILE DETAILING OF STEEL MOMENT FRAMES:

BASIC CONCEPTS, RECENT DEVELOPMENTS

AND UNRESOLVED ISSUES

Michael D. Engelhardt1

SUMMARY

This paper provides a discussion of key issues related to ductile detailing of seismic-

resistant steel moment frames. Basic concepts of ductile detailing are discussed, followed

by a detailed discussion of beam-to-column connections in steel moment frames, and a

brief discussion of stability issues. The paper concludes by looking at several unresolved

issues in ductile detailing of steel moment resisting frames.

INTRODUCTION

Moment resisting frames (MRFs) are among of the most commonly used lateral force

resisting systems for steel buildings. Steel MRFs typically consist of a simple rectangular

arrangement of beams and columns, with the beams rigidly connected to the columns to

permit transfer of moment. MRFs resist lateral loads through rigid frame action, i.e., by

the development of bending and shear in the beams and columns. Figure 1 shows a photo

of a recently constructed steel MRF located in California.

MRFs have a number of advantages, which often make them a popular choice for a

lateral force resisting system. A key advantage of MRFs is their architectural versatility.

The openness of this framing system eliminates architectural obstructions caused by

braces or walls, permitting maximum flexibility in space usage. A second reason why

MRFs are frequently chosen as a lateral system is their reputation for providing high

levels of ductility. Despite recent problems with these systems in the Northridge and

Kobe Earthquakes, MRFs still enjoy a reputation as providing excellent ductility and

safety under earthquake loading. Building codes in the US and elsewhere typically assign

MRFs the highest force reduction factors among all structural systems, reflecting the high

level of ductility and safety anticipated for this system. On the other hand, a disadvantage

of MRFs in design can be their flexibility. Compared to braced frames, MRFs have

relatively low stiffness, and therefore often require larger members to satisfy code

specified drift limits.

Steel MRFs have a history of good performance is past earthquakes, with few reported

collapses and little loss of life attributed to these systems. Nonetheless, experience with

steel MRFs in strong earthquakes has not been entirely positive. The collapse of a 22

story building of the Pino Suarez complex in the 1985 Mexico City Earthquake (Ger et al

1 Professor, Department of Civil Engineering, University of Texas at Austin, Austin, Texas 78712

2

1993), followed by extensive damage observed at welded moment connections in both

the 1994 Northridge Earthquake (Youssef et al 1995) and the 1995 Kobe Earthquake

(Steel Committee 1995) demonstrated vulnerabilities in steel MRF systems. The lessons

of these earthquakes indicate that care in the design, detailing, and construction of steel

MRFs is needed to assure satisfactory performance in strong earthquakes. Fortunately,

steel MRF problems observed in recent earthquakes have been extensively investigated,

and lessons learned from these failures have led to improved design and construction

practices, as well as improved building code regulations. In the US, building code

provisions for earthquake resistant steel buildings, including steel MRFs, are contained in

the Seismic Provisions for Structural Steel Buildings published by the American Institute

of Steel Construction (American 2001). These provisions have been updated frequently in

recent years to reflect lessons learned from Northridge and Kobe.

Figure 1 - Typical Steel Moment Resisting Frame

This paper provides a brief overview of ductile detailing requirements for seismic-

resistant steel MRFs. Emphasis is placed on the behavior and design of moment resisting

beam-to-column connections. Proper design and detailing of connections is critical for

ductile response of an MRF, and has been the focus of attention in research since

Northridge and Kobe. This is followed by a brief discussion of stability issues in steel

MRFs.

3

BASIC CONCEPTS

Under the action of a strong earthquake, an MRF will be driven well into the inelastic

range of response. The fundamental goal of the designer is to understand and control this

inelastic response to permit the development of a plastic mechanism under lateral load

that is capable of sustaining large inelastic deformations, i.e., possessing large ductility.

Effective seismic-resistant design requires an understanding of inelastic frame behavior.

To this end, an understanding of classical methods of plastic analysis and design is highly

beneficial.

Ductile response of a steel MRF is achieved by yielding of steel. Nonductile response of

steel systems is the result of fracture or instability. Consequently, a key design objective

is to maximize the yielding of steel frame elements, while at the same time delaying the

onset of fracture or instability until large inelastic deformations are achieved. To achieve

this objective, the designer must first choose the frame locations where yielding is

intended to occur, i.e., the locations of plastic hinges. In a steel MRF, the designer

generally has the choice of three possible hinge locations: 1) plastic flexural hinges at the

beam ends; 2) plastic shear hinges in the joint panel zones; or 3) plastic flexural hinges at

the clear span ends of columns. Most building code regulations discourage the third

option, i.e., hinge formation in the columns, as this may lead to the formation of a soft

story. The prohibition of column hinges is generally implemented through strong column

– weak girder design requirements. Consequently, plastic hinge formation in steel MRFs

is typically restricted to the beam ends (flexural yielding), or the joint panel zone (shear

yielding), or to a combination of these two.

The designer can control the location of plastic hinges in the MRF by assuring that the

element where the hinge is intended to form is the weakest element in the frame.

Conversely, this can be achieved by assuring that all other frame elements (i.e. elements

where inelastic action should not occur) are stronger, through the application of capacity

design concepts. For example, the most common design philosophy for MRFs is to

develop a plastic mechanism with plastic hinges at the beam ends. To assure that the

hinges actually form in the beam ends, all other frame elements (columns, joints and

connections) are designed to develop the capacity of the beams. In this approach, the

maximum moment that can be generated at the end of a fully yielded and strain hardened

beam is first estimated. This can be taken as the plastic moment of the beam, computed

using an estimate of the actual yield stress of the beam (as opposed to the minimum

specified yield stress), with an additional allowance for strain hardening. This provides a

reasonable upper bound estimate of the maximum moment that can be developed at the

beam ends regardless of the intensity of the earthquake loading on the frame. Next, the

shear at the beam ends that is in equilibrium with the maximum end moments is then

computed. These maximum beam end moments and shear forces are then used as the

loads for which the beam-to-column connections are designed, for which the joint panel

is designed, and for which the columns are designed. Thus, the connections, joints and

columns are not designed for code specified lateral forces, but rather are designed to

resist the maximum forces that can be imparted on them by the yielded and strain

hardened beams, i.e., they are designed to develop the capacity of the beams. This design

4

approach assures that the desired mechanism, i.e., a mechanism with hinges in the beam

ends, will actually form.

The concept of capacity design can also be understood through an analogy to electrical

wiring in a building. Electrical wiring is protected by a fuse, so that in case of an

electrical overload, the fuse burns out before the wiring is damaged, thereby protecting

the wiring. This protection can only be achieved if the fuse is weaker than the wiring. In a

seismic-resistant frame, the plastic hinge locations serve as a fuse to protect against

overload. In an earthquake, the plastic hinge “fuses” limit the forces that can be

transferred to the remainder of the frame, thereby protecting the remainder of the frame.

However, this can only be achieved if the plastic hinge locations are the weakest

elements of the frame, to assure ductile yielding occurs in the beam ends, before say

fracture of a connection or buckling of a column.

The final key element in the design of a steel MRF is to properly detail the plastic hinge

regions to permit the development of large levels of ductility at the hinges prior to the

occurrence of fracture or instability. For plastic hinges at the beam ends, the goal is to

develop the beam’s fully plastic moment, and maintain this moment through large plastic

rotations. Current US building code requirements for steel MRFs with the highest levels

of ductility (known as Special Moment Frames) require that plastic hinge regions be

capable of developing at least 0.03 radian of plastic rotation without significant loss of

strength. The actual code requirement calls for the development of a total (elastic +

inelastic) interstory drift angle of at least 0.04 radian without loss of strength. This

requirement is based on the assumption that a typical steel MRF yields at an interstory

drift angle of about 0.01 radian, thereby leaving 0.03 radian available for inelastic

rotation.

Proper detailing of the hinge regions therefore requires that the beam ends maintain their

plastic strength for 0.03 radian of plastic rotation without the occurrence of instability or

fracture. Instability can cause premature strength loss, i.e. loss of ductility, through the

occurrence of local buckling in the beam (flange and/or web buckling) or through the

occurrence of lateral torsional buckling of the beam. Premature strength loss due to local

buckling can be prevented by adhering to code specified (American 2000) width-

thickness limits for the beam flanges and web. Similarly, premature strength loss due to

lateral torsional buckling can be prevented by adhering to code specified lateral bracing

requirements. These width-thickness limits and lateral bracing requirements can normally

be easily satisfied for typical steel MRFs with members made of standard hot-rolled wide

flange shapes. However, when beams are made of built-up wide flange shapes or of

trusses, particular care is needed to assure that stability limits are satisfied.

Once stability concerns are addressed through width-thickness limits and lateral bracing,

the other key detailing issue is to prevent premature loss of strength at the beam ends due

to the occurrence of fracture. If a fracture is to occur, it will normally occur at the beam-

to-column connection. Thus, a critical issue in the design of a seismic-resistant steel MRF

is the proper design, detailing and construction of the beam-to-column connections to

permit the development of the full plastic strength of the beam and to maintain that

5

strength through large plastic rotations, without the occurrence of fracture. The

widespread occurrence of fractures at beam-to-column connections in the Northridge

Earthquake clearly indicated that this design goal was not achieved. The experience of

the 1994 Northridge Earthquake, followed by a similar experience in the 1995 Kobe

Earthquake, indicated a fundamental deficiency in beam-to-column connections in steel

MRFs. The following sections of this paper provide a detailed discussion of issues related

to beam-to-column connections.

BEAM-TO-COLUMN CONNECTIONS

The 1994 Northridge Earthquake caused unprecedented damage at welded beam-to-

column connections in many steel moment frame buildings in the Los Angeles area. This

damage motivated a large number of investigations and research studies aimed at

understanding the causes of the damage and at developing improved design and

construction practices for better performance of welded connections in future

earthquakes. A number of key factors contributing to the damage have been identified,

and a number of improved moment connection details have been developed. Further,

many of the important advances in moment connection design and welding have been

rapidly adopted for the construction of new steel buildings in high seismic zones of the

US.

The following sections provide a brief review of typical US design and welding practices

for steel moment connections prior to the Northridge Earthquake. Some of the most

common types of damage observed at these connections after the Earthquake are then

illustrated. This is followed by a discussion of some of the major factors that have been

identified as contributing causes of the damage. Some of the modifications and

improvements for steel moment connections implemented in US practice since the

Northridge Earthquake are then presented.

Northridge Connection Damage

The primary damage to steel moment frames in the Northridge Earthquake occurred at

the typical welded flange - bolted web moment connection detail widely used in west

coast US construction since the early 1970's. Many thousands of steel moment frame

buildings have been constructed in high seismic zones of the US using this connection

detail. In recent years, the welded flange - bolted web moment connection detail has also

become quite popular in lower seismic zones in the U.S. for resisting lower seismic loads

and wind loads.

Figures 2 and 7a show the welded flange - bolted web moment connection detail as

commonly constructed prior to the Northridge Earthquake. The beam flanges were field

welded to the column using single bevel complete penetration groove welds, and the

beam web was field bolted to a single plate shear tab. Field welding of the beam flange

groove welds was most commonly accomplished using the self shielded flux cored arc

6

welding (SS-FCAW) process. The electrode most commonly used for these welds was

classified as E70T-4 according to the classification system of the American Welding

Society. This electrode provides a specified minimum tensile strength of 480 MPa but has

no minimum specified notch toughness. The backing bars and weld runoff tabs used to

make the groove welds were normally left in place after completion of the weld.

Column

Beam

Column Flange

Beam Bottom Flange

Weld Access Hole

Shear Tab

Backing Bar

Detail at Bottom Flange Weld

Figure 2 - Typical Pre-Northridge Welded Flange - Bolted Web Moment Connection

Continuity plates were sometimes provided in the column as shown in Figure 7a.

Similarly, doubler plates were sometimes provided in the column to increase the column

panel zone shear strength. Code provisions and typical design practices for establishing

the need, size and welding details for continuity plates and doubler plates have varied

over the past 30 years. In some cases, additional welds were provided between the shear

tab and the beam web, as shown in Figure 7a. These welds were first required in US

building code provisions released in 1988 (International 1988). These welds were

required for any beam in which the plastic modulus of the beam web was in excess of 30

percent of the total plastic section modulus of the entire cross-section. The additional web

welds were intended to increase the amount of bending moment transferred by the web

connection, thereby relieving somewhat the moment transferred by the flange welds

(Engelhardt and Husain 1993).

The total number of steel moment frame buildings damaged in the Northridge Earthquake

has not been documented, although the number has been speculated to be well in excess

of 100. The number of damaged moment connections appears to be in the thousands.

The damaged steel buildings cover a wide spectrum of ages, heights, and configurations

(Youssef et al 1995). Many of the damaged buildings were designed and constructed to

the most recent building codes in effect at the time of the earthquake. The widespread

7

nature of the damage suggests fundamental problems in design and construction practices

for these buildings.

The most common forms of damage observed at these connections after the Northridge

Earthquake were a variety of fractures in the vicinity of the beam flange groove welds

(Youssef et al 1995, American 1994, Bertero et al 1994, Engelhardt and Sabol 1995). The

majority of fractures were observed at beam bottom flange welds, although some damage

also occurred at top flange connections. Interestingly, many of the buildings with

damaged connections showed no outward signs of distress, with virtually no

nonstructural damage. Many of these buildings were believed to be completely

undamaged following the earthquake, with the normal function of the building continuing

uninterrupted. Connection damage was revealed only upon detailed visual and ultrasonic

inspection of the connections.

Figure 3 schematically illustrates some typical fractures observed in the vicinity of the

beam bottom flange groove welds. Examples of several of these fractures are shown in

the photographs in Figs. 4 to 6.

Figure 3 - Typical Fractures

Probably the most common type of fracture observed at steel moment connections was

through the weld metal, typically near the face of the column (Fig. 3a). Figure 4

illustrates examples of these fractures. Note the business card shown in the photo in Fig.

4a, indicating that the fracture extends the full depth of the weld. In some cases, the

fractures did not extend through the full depth of the weld and could only be detected

through ultrasonic examination of the weld. Subsequent investigations have suggested

that many small weld fractures found near the root of the bottom flange weld were not in

(a) (b)

(c) (d)

8

fact caused by the earthquake, but were likely welding defects that occurred during

construction of the buildings (Paret and Freeman 1997).

(a) (b)

Figure 4 - Fractures at Interface of Groove Weld and Column Flange

Figure 5 - "Divot" Type Fracture in Figure 6 - Fracture Through Column

Column Flange Flange and Web

Another type of damage is illustrated in Figs. 3b and 3c. In this case, a fracture initiating

at or near the groove weld runs up through the column flange. In some cases the fracture

stops within the column flange, as in Fig. 3b. In other cases, the fracture emerges from

the column flange at a distance of several centimeters above the top of the groove weld.

For this type of failure, a "divot" of column flange material is pulled away from the

column. The photo in Fig. 5 illustrates such a “divot” type fracture.

Figure 3d illustrates a fracture that runs across the column. In some cases, only the

column flange fractured. In other cases the fracture continued into the column web (Fig.

6). In a small number of cases, fractures were reported to run across the full width of the

column, passing through both flanges and the web.

9

As described earlier, the design intent for a seismic-resistant steel MRF is that in the

event of a strong earthquake, ductile flexural yielding develop at the beam ends without

the occurrence of fracture at the connections. The non-ductile damage observed after the

Northridge Earthquake was clearly contrary to the design intent of a ductile moment

frame. Fortunately, none of the connection damage resulted in collapse of a steel moment

frame building nor did it result in loss of life. Such damage, however, may represent a

more serious safety concern for ground motions that differ in intensity, duration, or

frequency content from that experienced in Northridge. While the safety implications of

the Northridge moment connection damage have been the subject of debate, there appears

to be broad consensus that fractures at moment connections represent an undesirable type

of response that should be prevented in future earthquakes.

Causes of Connection Damage

A great deal of discussion and debate regarding the causes of the Northridge moment

connection damage ensued since the first discovery of fractures shortly following the

earthquake. This discussion has not yet led to a complete consensus among the structural

engineering profession in the US regarding the causes of the damage, and disagreement

still remains on several issues. Nonetheless, a number of field, laboratory, and analytical

studies have been completed investigating moment connection behavior and damage. The

results of these and other studies have provided considerable insight into some of the key

factors that may have contributed to the damage.

A number of laboratory tests on “Pre-Northridge” connections were conducted at several

universities following the earthquake to better understand the behavior and controlling

failure mechanisms of this connection (Popov et al 1995, Shuey et al 1995, Whittaker et

al 1995, Uang and Bondad 1995). These tests were conducted on specimens constructed

using design details and welding practices typical of west coast U.S. construction prior to

the Northridge Earthquake. These test programs confirmed the poor behavior of this

connection detail, with a number of test specimens experiencing connection failures with

little or no ductility. All of the failure modes observed in the field (Fig. 3) were

successfully reproduced in the laboratory tests, including weld fractures, “divot” type

column fractures, and fractures running across the column flange and web.

An example of a tested connection detail is illustrated in Fig. 7a, and its experimental

response, plotted as bending moment at the face of the column versus plastic rotation, is

shown in Fig. 7b (Shuey et al 1995). Typical of many Pre-Northridge specimens, this

connection did not develop the plastic moment of the beam, and exhibited a premature

failure with essentially no ductility. This particular specimen exhibited a divot type

fracture in the column flange material near the bottom flange groove weld, similar to Fig.

3c. Recent building code provisions in the US now require that steel moment connections

be capable of developing at least 0.03 radian of plastic rotation without failure. The test

result shown in Fig. 7b clearly falls very short of this requirement. The very poor

performance of this specimen also calls into question the suitability of the Pre-Northridge

type connection for wind load applications.

10

10

30o

T & B FLANGESE70T-4Backing Bar and Weld TabsTo Remain In Place

8

8

BOLTS: 10 - 22mm A325-SCHOLES: 24 mm

P 16 x 127 x 760L

8 100T & B of PL

8T & B of PL

75

W36x150 (A36)

W14x257(A572 Gr. 50)

P 12 x 150L(Both Sides)

CJP - 3 SidesTYP

(a) Connection Detail for Test Specimen

-5000

-4000

-3000

-2000

-1000

0

1000

2000

3000

4000

5000

-0.03 -0.02 -0.01 0 0.01 0.02 0.03

Plastic Rotation (rad)

Ben

din

g M

om

en

t (k

N-m

)

X

Brittle Fracture at Bottom

Flange Weld

Mp

Mp

(b) Response of Test Specimen

Figure 7 - Example of Experimental Response of Pre-Northridge Connection

In a study conducted at Lehigh University, detailed metallurgical and fractograhic

examinations were made on several failed “Pre-Northridge” laboratory test specimens,

including the specimen shown in Fig. 7 (Kaufmann and Fisher 1995a, Kaufmann et al

1996a). This study also examined portions of damaged connections removed from actual

buildings in the Los Angeles area. The study concluded that the majority of connection

failures occurred by a brittle fracture mechanism initiating at the root of bottom flange

weld. The point of fracture initiation was typically at a weld defect located at mid-width

of the weld, i.e. in the area of the weld access hole. Thus, even fractures within the

column, of the type illustrated in Figs. 3b to 3d, appeared to initiate at the weld root in

many cases. The Lehigh study also concluded that low toughness of the weld metal

played a very important role in triggering the brittle fractures. Analytical studies of

building response in the Northridge Earthquake suggest that many connections fractured

11

while the attached beam was still elastic, further confirming the nonductile behavior of

the connection (SAC 1995).

Numerous studies conducted after the Northridge Earthquake, including those discussed

above, as well as studies conducted prior to the Northridge Earthquake, provide

considerable insight into the problems associated with the “Pre-Northridge” connection

and the causes of the damage observed after the Northridge Earthquake. Following is a

brief discussion of these factors.

Welding Factors

Since the majority of fractures observed after the Northridge Earthquake were at beam

flange groove welds, welding related issues are clearly of importance in understanding

the connection failures. Several issues related to welding have been identified as

potentially contributing to the observed damage, including low weld metal toughness,

poor workmanship, and detrimental effects of backing bars and weld tabs.

Requirements for structural welding in the US are specified in the Structural Welding

Code - Steel, AWS D1.1, published by the American Welding Society (American 2000).

This document includes requirements for qualification of welding procedures and of

welders, as well as requirements for inspection. Prior to the Northridge Earthquake, this

document did not contain specific requirements for earthquake resistant construction. It

did, however, differentiate between statically loaded versus dynamically loaded

structures. The welding requirements for dynamically loaded structures were largely

intended for high cycle fatigue conditions and thus were not usually specified for

earthquake resistant building construction. Welding of the beam flanges therefore was

typically controlled by the static loading requirements of AWS D1.1 in most project

specifications. These requirements did not require removal of backing bars and weld tabs,

and did not require minimum levels of notch toughness for the weld metal.

Nondestructive evaluation of the beam flange groove welds was most commonly

accomplished by ultrasonic testing. Depending on the code jurisdiction and project

specifications, normally between 25 percent and 100 percent of the beam flange groove

welds were ultrasonically tested. Ultrasonic acceptance criteria were typically those

specified for statically loaded structures in AWS D1.1.

Among the most important welding factors contributing to the moment connection

damage observed after the Northridge Earthquake appears to be low weld metal

toughness. Beam flange groove welds prior to the Northridge Earthquake were

commonly made using the E70T-4 electrode using the flux-cored arc welding process.

Charpy V-Notch (CVN) tests on weld metal from damaged building connections, as well

as from laboratory specimens, sometimes showed values under 10 to 15J at room

temperature. This low toughness made the connections highly vulnerable to brittle

fracture and was likely a leading factor responsible for the poor performance of the

welded flange – bolted web connection in the Northridge Earthquake and in laboratory

testing of this connection (Kaufmann and Fisher 1995a 1995b, Kaufmann et al 1996b,

12

Tide 1994, Tide 1998). Prior to the Northridge Earthquake, it was not common practice

to specify toughness requirements for weld metal in building connections, and no such

requirements were contained in US codes for seismic resistant construction. Further, it

appears that little, if any, US research on moment connections prior to the Northridge

Earthquake identified the importance of weld metal toughness for seismic applications.

In addition to low weld metal toughness, inadequate welding workmanship and

inspection have also been identified as potentially important contributors to the observed

damage (Tide 1995, American 1994). Upon discovery of the moment connection damage

after the Northridge Earthquake, poor quality welding combined with inadequate

inspection was widely speculated to be a primary cause. While field investigations clearly

revealed welding defects and poor welding practices in some instances, the precise role

that these defects played in the failures is not clear. For example, large-scale connection

tests conducted shortly after the Northridge Earthquake evaluated welded flange - bolted

web connections that were constructed under close scrutiny with very high quality

welding workmanship (Engelhardt et al 1995). In addition to the very close attention to

workmanship, the backing bars and weld tabs were removed from the groove welds to

eliminate any potential notch condition introduced by these items. The welds passed two

independent ultrasonic tests. Despite these measures to assure the highest levels of

workmanship, the connections still experienced brittle fractures at the beam flange

groove welds at very low levels of ductility. These failed specimens were welded using

the low toughness E70T-4 electrode, suggesting the importance of weld metal toughness.

These test results imply that even with high quality welding workmanship, failures might

still have been expected in the field. In many instances, it appears that low weld metal

toughness, combined with the presence of weld defects, made the connections

particularly vulnerable to brittle fracture.

A final welding related factor implicated in the Northridge damage is the presence of

backing bars and weld tabs (American 1994, Kaufmann and Fisher 1995b, Tide 1995). In

typical practice prior to the Northridge Earthquake, the beam flange groove welds were

made using backing bars and weld tabs that were usually left in place after completion of

the weld. Several investigators have noted that a backing bar may act as an artificial edge

crack, and can initiate a brittle fracture. Similarly, the weld runoff regions contained

within the weld tabs have been speculated to be potential fracture initiation sites. In

addition to introducing a notch, the left-in-place backing bar may also inhibit inspection

of the weld. The backing bar may increase the difficulty in properly interpreting

ultrasonic test results, and also prevent visual inspection of the weld root.

Design Factors

Tests conducted on the welded flange – bolted web connection both before and after the

Northridge Earthquake indicate that even if weld fracture is prevented, connection

performance is often still unsatisfactory (Engelhardt and Husain 1993, Tsai and Popov

1988, Stojadinovic et al 2000). Rather than fractures within the weld, fractures are

frequently observed in the base metal region immediately adjacent to the weld. Such base

13

metal fractures often occur after only limited ductility is developed. Such evidence

suggests fundamental design deficiencies with the welded flange-bolted web detail.

Both laboratory and analytical studies indicate that very high levels of stress and strain

are developed in the vicinity of the beam flange groove welds (El-Tawil et al 2000, Yang

and Popov 1995, Mao et al 2001). These high demands have been attributed to a number

of causes. Inadequate participation of the bolted beam web connection in transferring

moment and shear has been identified as a cause of high beam flange stresses. It appears

that much of the moment and shear normally carried in the beam web, as predicted by

simple beam theory, is actually transferred through the flanges at the connection, which

serves to significantly increase the stress at the beam flange groove welds. Very high

localized stresses and strains are also possible due to local bending of the column and

beam flanges near the beam flange groove welds. The weld access hole also introduces a

significant stress concentration (El-Tawil etal 2000, Mao et al 2001). The severity of this

stress concentration depends on the size, placement and geometry of the access hole.

Deformations of the column panel zone also appear to have an important effect on the

state of stress in the region of the beam flange groove welds, as does the presence and

thickness of continuity plates. Finally, analyses have also suggested that a high degree of

restraint can exist in the beam flange groove welds near the face of the column, resulting

in the development of complex triaxial states of stress. The presence of such triaxial

states of stress has also been conjectured to contribute to poor ductility in this region

(Blodgett 1995, Yang and Popov 1995).

Finite element analyses of welded flange-bolted web connections indicate that very high

localized stresses can occur at the root of the bottom flange groove weld, particularly

near the center portion of the flange in the vicinity of the weld access hole. As noted

earlier, this region also frequently contains weld defects due to the difficulty of welding

and inspection in this area. Consequently, it appears that very high stresses occurred in a

region of the connection that had a high likelihood of defects and very low toughness

material. The combined effects of high stresses, large defects and low toughness seem to

have virtually assured poor performance of this connection.

Base Metal Factors

In typical US practice prior to the Northridge Earthquake, beams in steel moment frames

were specified to be of ASTM A36 steel (minimum specified Fy = 250 MPa). Columns

were normally specified to be either of ASTM A36 steel or of ASTM A572 Grade 50

steel (minimum specified Fy = 345 MPa). There are several factors related to the

properties of the steel in the beams and columns that have been conjectured to have

played a role in the observed connection damage. One of these factors is the high level of

actual yield strength in ASTM A36 beams. A statistical study for A36 steel sold in the

US showed a mean yield stress of 340 Mpa for rolled shapes, in comparison with the

minimum specified value of 250 Mpa (SAC 2000b). These high values of yield stress

result in high stresses on the beam flanges, the beam flange groove welds, and in the

column, prior to the development of a plastic hinge in the beam. These higher stress

14

levels further increase the likelihood of fracture. In effect, the beam-to-column

connection is intended to be stronger than the beam. As beam strength increases due to

elevated yield stress values, the connection may in fact become weaker than the beam,

promoting non-ductile connection failure prior to ductile flexural yielding of the beam.

Elevated yield stress values are not generally accompanied by an equal increase in tensile

strength. Consequently, the yield ratio (Fy /Fu) of the steel is also increased. A statistical

study showed that the mean value of yield ratio for A36 steel was 0.72, compared to a

value of 0.62 based on the minimum specified yield and tensile strength (SAC 2000b).

High values of yield ratio may further increase the likelihood of beam flange fracture.

Questions have also been raised in regard to through-thickness properties for column

flanges. These questions were motivated largely by the discovery of “divot” type

fractures within columns, as illustrated in Figs. 3b and 3c. As noted earlier, forensic

studies on failed connections suggest that most column fractures initiated at the weld root

and then propagated into the column (Kaufmann and Fisher 1995a). Further, recent tests

evaluating through-thickness loading on column flanges have shown excellent properties

(Dexter 2000). Consequently, currently available research suggests that inadequate

through-thickness properties of column flanges was not likely a significant contributor to

the Northridge connection damage.

Methods to Achieve Improved Connection Performance

US building code provisions for seismic resistant steel construction have been

extensively revised to reflect the experiences of the Northridge Earthquake (American

2000, SAC 2000a). These provisions require that connections for special moment frames

be capable of developing at least 0.04 radian interstory drift angle without failure. This

level of deformation capacity must be verified by cyclic loading tests on full scale or

nearly full-scale specimens using standard loading protocols. As described earlier, this

requirement presumes that approximately 0.01 radian drift angle will be accounted for by

elastic deformations. Consequently, these code provisions imply the need to develop 0.03

radian of plastic rotation. Developing this level of plastic rotation has therefore been the

primary goal in the development of new moment connections for use in US practice. To

achieve this high level of ductility has required improved practices with respect to

welding, connection design and detailing, and steel material characterization.

Major changes to welding practices have been implemented since the Northridge

Earthquake. The most significant change has been recognition of the importance of weld

metal toughness. It has become common practice to specify minimum notch toughness

requirements for weld metal in groove welds. Much of the successful research on

improved connection details has used weld metal with a minimum specified CVN value

of 27 J at -29o C. This level of notch toughness is currently required in building code

provisions for seismic resistant steel moment frames (American 2000). In the near future,

this requirement will likely be refined to call for CVN values of 27 J at -18o C and 54 J at

room temperature (SAC 2000a).

15

Practices with respect to groove weld backing bars and weld tabs have also changed.

Prior to the earthquake, it was common practice to leave backing bars and weld tabs in

place. Since the Northridge Earthquake, it has become common practice to remove the

bottom flange backing bar in an attempt to eliminate the notch effect of the bar and to

permit better inspection of the weld root. At the top flange groove weld, the backing bar

is typically left in place, but is welded to the column face in an attempt to mitigate the

bar’s notch effect. Weld runoff tabs are typically removed at both the top and bottom

flange groove welds.

A number of tests have been conducted on the welded flange – bolted web connection

detail where the above welding improvements have been implemented (Stojadinovic et al

2000). That is, the connections were welded using electrodes with a CVN rating of of 27

J at -29o C, the bottom flange backing bars were removed, the top flange backing bars

were welded to the face of the column, and weld tabs were removed at the top an bottom

flanges. However, other than these welding modifications, the basic connection design

was not changed. In these tests, no weld fractures were observed, indicating that the

welding modifications were effective in preventing weld failure. However, the

connections still failed by fracture in the base metal or heat affected zone adjoining the

beam flange groove welds, typically initiating near the weld access hole. These

connections developed plastic rotations on the order of about 0.01 to 0.02 radian. These

tests suggest that welding improvements alone cannot assure the development of 0.03

radian of plastic rotation.

In addition to changes in welding practices, improvements have also been made in

structural steel properties. A new grade of structural steel has recently been introduced in

the US. This new grade, designated ASTM A992 provides better control of yield stress

and yield ratio. ASTM A992 has a minimum specified Fy of 345 MPa and a minimum

specified Fu of 450 MPa, the same as A572 Grade 50. However, unlike A572 Grade 50,

the new A992 steel places an upper limit on Fy of 450 MPa, and an upper limit on yield

ratio (Fy / Fu) of 0.85. This closer control on yield and yield ratio should permit more

reliable connection design.

US building code provisions (American 2000) have also been updated to reflect the fact

that steel often has a yield stress higher than the minimum specified value. This has been

accomplished by introducing into building code provisions the expected yield stress of

steel, designated as Fye. The value of Fye is intended to provide an estimate of the actual

expected yield stress of a particular grade of steel. For ASTM A36, A572 Grade 50, and

A992, the value of Fye is specified to be 380 MPa. This indicates that the actual yield

stress of each of these three grades of steel is expected, on average, to be 380 MPa. The

expected yield stress is used for all capacity design calculations. For example, the beam-

to-column connection must be designed to develop the capacity of the beam. When

computing the maximum moment expected at the end of a yielded beam, the designer

assumes the yield stress of the beam is equal to the expected yield stress.

16

In addition to changes in welding practices and material characterization, a wide variety

of connection design and detailing modifications have been investigated in an attempt to

develop improved ductility. Many of these modifications have been implemented in

construction practice. An example of a successful design that is currently seeing

widespread usage in US building construction practice is the Reduced Beam Section

(RBS) connection. In the RBS connection, portions of the beam flanges are trimmed near

the beam-to-column connection. Plastic hinge formation occurs within this reduced

section, rather than at the face of the column, thereby reducing stress and strain demands

in the critical region of the beam flange groove welds. When the RBS is combined with

the welding improvements described above, excellent performance has been achieved in

laboratory tests.

An example of test results on an RBS connection (Engelhardt 1998) is shown in Fig. 8.

Figure 8(a) shows the connection detail, Fig. 8(b) shows response under cyclic loading,

and Fig. 8(c) shows a photo of the specimen at the end of testing. This specimen

developed well above 0.03 radian of plastic rotation without significant strength loss. The

yielding pattern seen in Fig. 8(c) shows the formation of a plastic hinge within the

reduced section. Compared to the response of the Pre-Northridge connection (Fig. 7b),

the response of the RBS shows that excellent connection performance can be achieved.

A number of beam-to-column connection details, including the RBS, have been

investigated in a major research program coordinated by the SAC Joint Venture. These

include both welded and bolted connection options. Through this work, detailed design

guidelines have been developed, and have been published (FEMA 2000a). Consequently,

designers can now have a variety of connection details available for use in steel MRFs

that have been tested and for which design guidance is available.

STABILITY ISSUES

As discussed earlier, the ductility of a steel MRF will ultimately be limited by the

occurrence of fracture or instability. Thus, in addition to addressing fracture concerns at

connections, design of a ductile MRF also requires consideration of stability issues. The

deformation capacity of a plastic hinge in a beam can be limited by the occurrence of

local flange buckling, local web buckling or lateral torsional buckling at the hinge region.

Frequently, a combination of all of three of these types of buckling is observed in

laboratory tests. Building code regulations (American 2000) provide limits on width-

thickness ratios and lateral bracing requirements that are intended to permit a steel

member to develop its full plastic strength, and maintain that plastic strength through

large plastic rotations before local and/or lateral torsional buckling ultimately cause

failure.

17

For wide flange beams in steel MRFs, current codes (American 2000, SAC 2000a) limit

the flange slenderness ratio bf/2tf to a maximum value of yF52/ , and the web

slenderness ratio h/tw to a maximum value of yF418/ , where Fy is in ksi. The spacing

of beam lateral bracing is limited to a maximum value of 2500ry /Fy. The web slenderness

limit was recently reduced from yF520/ to yF418/ based on studies (Uang and Fang

2001) that indicate web buckling plays a particularly important role in triggering strength

degradation and that stricter web slenderness limits are needed.

30o

L

45o

30o

Grind Parallel to Flang eGrind Smooth

10

6

658

Bolts: 25 mm A325 225 mm c-cHoles: 27mm D IA.P 9.5 x 152 x 762

10

8

8

8

1015 mm Radius

230 685

60

60

190

B.U . Bar to R emain

Weld B.U . Bar to Column

After Root isC leaned and Inspected

Remove B.U . Bar

W36x194 (A36)

W14x426

A572 Gr. 50

LP 25 x 125

Groove welds:E71T-8Remove Weld Tabs

-5000

-4000

-3000

-2000

-1000

0

1000

2000

3000

4000

5000

-0.04 -0.03 -0.02 -0.01 0 0.01 0.02 0.03 0.04

Plastic Rotation (rad)

Ben

din

g M

om

en

t (k

N-m

)

Mp

Mp

(a) RBS Connection Test Specimen Detail (b) Response of Test Specimen

(c) Photo of RBS Specimen at Completion of Test

Figure 8 – Example of Experimental Response of an RBS Connection

18

The RBS specimen shown in Fig. 8(c) clearly shows evidence of local buckling. The

W36x194 beam section used for this specimen fully satisfies the code specified width-

thickness limits and lateral bracing requirements noted above. Thus, although local and

lateral buckling occurred within this beam at large plastic rotation levels, the rate of

strength degradation was quite gradual, as seen in the moment-rotation response in Fig.

8(b) and did not prevent the beam from achieving 0.03 radian of plastic rotation.

When members are used with high width-thickness ratios, i.e., very slender elements, or

adequate lateral bracing is not provided, the ductility of a plastic hinge can be greatly

reduced due to the premature occurrence of local or lateral torisonal buckling. In these

cases, the rate of strength degradation can be very rapid, similar to the rate of strength

degradation due to fracture. Thus, adherence to code specified width-thickness limits and

lateral bracing limits is essential for ductile response.

In a typical MRF, in addition to developing plastic hinges at the beam ends, plastic

hinges can also be expected to form at the base of columns. Consequently, in this region,

width-thickness limits and lateral bracing requirements similar to those used for beams

should also be applied to the colunms. However, even for portions of columns that are

expected to remain elastic, the use of slender cross-section elements or large unbraced

lengths is inadvisable. Even with sophisticated inelastic analysis methods, the response of

a steel MRF to strong earthquake motions cannot be predicted with certainty, and the

possibility of unanticipated inelastic action in the columns must be considered. If slender

elements are used in the column, such unanticipated inelastic action can result in failure

of the column and can endanger the gravity load resisting system of the building. In

general, the use of slender elements should be avoided in any member of an MRF to

provide a robust structural system capable of ductile response under a wide range of

conditions.

In addition to local buckling, more global forms of instability must also be prevented.

These can include the overall buckling of a column, or sidesway instability of a story or

of the entire frame. The prediction of these types of instability under dynamic loading is

not yet completely understood. However, building codes contain a number of simplified

provisions that are intended to prevent global instability of a column, story or the entire

frame. The buckling of an isolated column can be addressed by designing that column

using capacity design concepts. That is, the maximum forces that can be transferred to the

columns by the yielded beams can be estimated, and the columns can then be designed

for these forces. Alternatively, building codes (American 2000) provide a simplified

approach that requires designing the column for a specified amplified force. The

prevention of a story sidesway collapse is typically addressed by strong column – weak

girder requirements. Code imposed drift limits are also intended to contribute to the

prevention of overall frame instability.

Although the prevention of global modes of instability in an MRF is addressed through

simplified rules and approaches, experience suggests that these simple approaches have

been effective in providing safe designs. In the long run however, an improved

understanding of frame stability under dynamic loading would be desirable. In the mean

19

time, however, it is essential to adhere to building code design rules that are intended to

address global instability, particularly strong column – weak girder design provisions as

well as code specified drift limits.

UNRESOLVED ISSUES

The extensive body of research on steel MRFs combined with many years of experience

with these systems in the field has provided the basic technology needed to design steel

MRF systems capable of providing a high degree of safety under strong earthquake

ground motions. Intensive research efforts since the Northridge and Kobe earthquakes

has significantly extended the understanding of steel MRF behavior, and important

results of this research have been rapidly implemented in building code regulations

(American 2000, SAC 200a) and in practice.

Nonetheless, there are still a number of issues affecting the performance or economy of

steel MRFs are not adequately understood. One of these unresolved issues is the proper

design of the column panel zone region. This region is subject to high shear forces, and

can form a plastic shear hinge in an MRF subject to lateral loading. Test results indicate

that very high levels of ductility can be achieved by permitting shear yielding within the

panel zone. However, tests and analysis also show that very large inelastic shear

deformations in the panel zone can result in fracture at the corners of the panel zone, in

the vicinity of the beam flange groove welds. The degree of shear yielding that be safely

accommodated within the column panel zone is not yet clearly understood. Recent

revisions to US building code (American 2000) provisions limit shear yielding of panel

zones, with the consequent need for relatively thick and expensive web doubler plates in

many designs. Further research is needed to better understand panel zone response,

particularly as it relates to the development of fracture.

Additional research is also needed on the proper design of column base plates in steel

MRFs. Very large forces and deformations can be generated at the base of the columns

under earthquake loading. However, very little research has been conducted on column

baseplates, and very little design guidance is available to designers.

Finally, as noted earlier, issues related to the overall stability of steel MRF systems under

dynamic earthquake loading are not well understood. Stability is currently addressed

through simplified rules and empirical procedures. Further research is needed to better

understand the stability of these systems.

ACKNOWLEDGEMENTS

The National Science Foundation (Award Nos. CMS-9416287 and CMS-9358186), the

Federal Emergency Management Agency, the National Institute of Standards and

Technology, and the American Institute of Steel Construction are gratefully

acknowledged for their support of the writer’s research on seismic resistant steel moment

frames.

20

REFERENCES

American Institute of Steel Construction (1994), Proceedings of the AISC Special Task

Committee on the Northridge Earthquake, Los Angeles, 14-15 March 1994.

American Institute of Steel Construction (2000). “Seismic Provisions for Structural Steel

Buildings – Supplement No. 2,” Chicago.

American Welding Society (2000), “Structural welding code – steel”, AWS Standard

D1.1-98, Miami.

Bertero V.V., Anderson, J.C. and Krawinkler, H. (1994), “Performance of steel building

structures during the Northridge earthquake,” Report No. UCB/EERC - 94/09,

Earthquake Engineering Research Center, University of California at Berkeley.

Blodgett, O. (1995), “Notes On Beam To Column Connections,” Steel Moment Frame

Connection Advisory No. 3. Report No. SAC 95-01. SAC joint Venture. Sacramento.

Dexter, R. and Melendrez, M. (2000), “Through-Thickness Properties of Column Flanges

in Welded Moment Connections,” Journal of Structural Engineering, ASCE, 126(1): 24-

31.

El-Tawil, S., Mikesell, T., and Kunnath, S.K. (2000), “Effect of Local Details and Yield

Ratio on Behavior of FR Steel Connections,” Journal of Structural Engineering, ASCE,

126(1): 79-87.

Engelhardt, M.D. and Husain, A.S. (1993), “Cyclic loading performance of welded

flange bolted web connections,” Journal of Structural Engineering, ASCE, 119(12):

3537-3550.

Engelhardt, M.D. and Sabol, T.A. (1995) "Lessons Learned from the Northridge

Earthquake: Steel Moment Frame Performance," Proceedings: Symposium on A New

Direction in Seismic Design, Tokyo, Architectural Institute of Japan, pp. 1-14.

Engelhardt, M.D., Sabol, T.A., Aboutaha, R.S. and Frank, K.H. (1995), “An Overview

Of The AISC Northridge Moment Connection Test Program”, Proceedings of the 1995

National Steel Construction Conference, San Antonio, 17-19 May 1995. American

Institute of Steel Construction, Inc. Chicago.

Engelhardt, M.D., Winneberger, T., Zekany, A.J. ,and Potyraj, T. (1998), “Experimental

Investigation Of Dogbone Moment Connections,” Engineering Journal, American

Institute of Steel Construction , 35(4), 128-139.

21

FEMA (2000a), “Recommended Seismic Design Criteria for New Steel Moment-Frame

Buildings,” Document FEMA-350, Federal Emergency Management Agency,

Washington, D.C.

FEMA (2000b), “State of the Art Report on Base Metals and Fracture,” Document

FEMA-355A, Federal Emergency Management Agency, Washington, D.C.

Ger, J.F., Cheng, F.Y. and Lu, L.W. (1993). “Collapse Behavior of Pino Suarez Building

During 1985 Mexico City Earthquake,” Journal of Structural Engineering, ASCE, 119

(3), 852-870.

International Conference of Building Officials (1988), “Uniform Building Code”,

Whittier, California.

Kaufmann, E.J. and Fisher J.W (1995a), “Fracture Analysis Of Failed Moment Frame

Weld Joints Produced In Full Scale Laboratory Tests And Buildings Damaged In The

Northridge Earthquake”, Technical Report: Experimental Investigations of Materials,

Weldments and Nondestructive Examination Techniques. Report SAC 95-08. SAC Joint

Venture. Sacramento.

Kaufmann, E.J. and Fisher, J.W. (1995b), “A Study Of The Effects Of Material And

Welding Factors On Moment Frame Weld Joint Performance Using A Small-Scale

Tension Specimen”, Technical Report: Experimental Investigations of Materials,

Weldments and Nondestructive Examination Techniques. Report SAC 95-08. SAC Joint

Venture. Sacramento.

Kaufmann, E.J., Fisher J.W., DiJulio, R.M. and Bavarian, B.B. (1996a), “Failure

Analysis Of Welded Moment Resisting Frames Damaged In The Northridge

Earthquake”, ATLSS Report No. 96-04. Lehigh University. Bethlehem, Pennsylvania.

Kaufmann, E.J., Xue, M,, Lu, L.W. and Fisher, J.W. (1996b), “Achieving Ductile

Behavior Of Moment Connections”, Modern Steel Construction. American Institute of

Steel Construction, Inc., January.

Mao, C., Ricles, J., Lu, L.W., and Fisher, J. (2001). “Effect of Local Details on Ductility

of Welded Moment Connections,” Journal of Structural Engineering, ASCE, 127(9):

1036-1044.

Paret, T.F. and Freeman, S.A. (1997), “Is Steel Frame Damage Being Diagnosed

Correctly”, Building to Last; Proceedings of Structures Congress XV, Portland, 13-16

April 1997. Vol. 1. ASCE, Washington. 261-266.

Popov, E.P., Blondet, M., Stepanov, L. and Stojadinovic, B. (1995), “Full-Scale Steel

Beam-Column Connection Tests”, Technical Report: Experimental Investigations of

Beam-Column Subassemblages, Report No. SAC 95-09, SAC Joint Venture, Sacramento.

22

SAC Joint Venture (1995), “Analytical And Field Investigations Of Buildings Affected

By The Northridge Earthquake Of January 17, 1994”, Report No. SAC 95-04, SAC Joint

Venture, Sacramento.

Shuey, B.D., Engelhardt, M.D. and Sabol, T.A. (1996), “Testing Of Repair Concepts For

Damaged Steel Moment Connections”, Technical Report: Experimental Investigations of

Beam-Column Subassemblages, Report No. SAC 95-09, SAC Joint Venture, Sacramento.

Steel Committee of the Kinki Branch of the Architectural Institute of Japan (1995),

“Reconnaissance Report on Damage to Steel Building Structures Observed from the 1995

Hyogenken-Nanbu Earthquake,” (In Japanese).

Stojadinovic, B., Goel, S., Lee, K.H., Margarian, A.G. and Choi, J.H. (2000), “Parametric

Tests on Unreinforced Steel Moment Connections,” Journal of Structural Engineering,

ASCE, 126(1): 40-49.

Tide, R. (1995), “ Fracture Of Beam-To-Column Connections Under Seismic Load”,

Habitat and the High-Rise, Tradition and Innovation; Proceedings of the Fifth World

Congress, Amsterdam, 14-19 May 1995.Council on Tall Buildings and Urban Habitat.

Bethlehem, Pennsylvania. 1273-1286.

Tide, R. (1998), “Stability Of Weld Metal Subjected To Cyclic Static And Seismic

Loading”, Engineering Structures, (20),4-6.

Tsai K.C. and Popov, E.P. (1988), “Steel Beam-Column Joints In Seismic Moment

Resisting Frames”, Report No. UCB/EERC - 88/19, Earthquake Engineering Research

Center, University of California at Berkeley.

Uang, C.M. and Bondad, D. (1995), “Static Cyclic Testing Of Pre-Northridge And

Haunch Repaired Steel Moment Connections”, Technical Report: Experimental

Investigations of Beam-Column Subassemblages, Report No. SAC 95-09, SAC Joint

Venture, Sacramento.

Uang, C.M. and Fang, C.C. (2001). “Cyclic Stability Criteria for Steel Moment

Connections with Reduced Beam Section,” Journal of Structural Engineering, ASCE,

127(9): 1021-1027.

Whittaker, A., Bertero, V.V. and Gilani, A. (1995), “Full-Scale Seismic Testing Of Steel

Beam-Column Assemblies”, Technical Report: Experimental Investigations of Beam-

Column Subassemblages, Report No. SAC 95-09, SAC Joint Venture, Sacramento.

Yang, T. and Popov, E.P. (1995), “Behavior Of Pre-Northridge Moment Resisting Steel

Connections”, Report No. UCB/EERC - 95/08, Earthquake Engineering Research Center,

University of California at Berkeley.

Youssef N.F.G, Bonowitz D. and Gross J.L. (1995) “ A Survey Of Steel Moment-

Resisting Frame Buildings Affected By The 1994 Northridge Earthquake.” NIST Report

No. NISTIR 5625, National Institute of Standards and Technology. United States

Department of Commerce Technology Administration. Washington.