DESIGN OF CONCRETE BRIDGES FOR …digitool.library.mcgill.ca/thesisfile106496.pdfi ABSTRACT Sustainable and durable infrastructure facilities, including bridges, require optimum use

DESIGN OF CONCRETE

BRIDGES FOR

SUSTAINABILITY AND DURABILITY

Juan Manuel Macía

Department of Civil Engineering and Applied Machanics

McGill University

Montréal, Canada

July 2011

A Thesis submitted to the Graduate and Postdoctoral Studies Office in partial fulfilment of the requirements of the degree of Master of Engineering (Thesis Option)

Copyright © by Juan Manuel Macía (2011)

To My Wife Maria Luisa

i

ABSTRACT

Sustainable and durable infrastructure facilities, including bridges, require optimum use of all

resources involving reduction of energy and water consumption during all project phases,

including planning, design, construction, maintenance, operations, repair and rehabilitation, and

finally decommissioning and disposal of the debris at the end of its service life. Design of a

sustainable and durable bridge structure requires consideration of a few feasible alternatives to

develop an optimum option to fulfill all of the relevant limit states, with the best life-cycle

performance and with the lowest life-cycle costs.

The current national standards do not account for the observed increases in operating loads and

the increasing deterioration of bridge structures over their service life. While these standards

emphasize quality control in choice of materials, design and construction, they do not provide

guidance and scientific tools to design and maintain a bridge structure for durability over its

service life, and include only prescriptive tools for preventing some deterioration modes.

This research program integrates sustainability and durability in the design of a conventional

bridge structure in a cold climate country, subjected to the various mechanical natural and man-

made loads and an aggressive environment, and considers the performance of the various

materials and structural components over the design service life. The latest available models of

the relevant deterioration modes have been incorporated in the life-cycle performance and design

considerations. The basic procedure adopts a multiple protection strategy for all deterioration

modes, resulting from the relevant aggressive actions, and integrates durability considerations

with structural calculations for the final design and defines maintenance strategies and any

needed supplementary protection techniques. The design-for-durability procedure is illustrated in

a worked out bridge design example.

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RÉSUMÉ

Les infrastructures durables, incluant les ponts, ont besoin de l’utilisation optimale de ressources

naturelles, en considérant une réduction de la consommation d’énergie, des matériaux et d’eau

pendant toutes les phases du projet, tels que la conception, la construction, l’entretien, l’opération,

la réfection, le renouvellement et finalement le démantèlement à la fin de la vie de service. La

conception de ponts sous le principe du développement durable demande la mise en

considération de quelques possibles solutions qui répondent aux différents états limites à

respecter, avec la meilleure performance et les coûts les plus bas pendant la vie de service de

l’ouvrage.

Les normes de conception de ponts au niveau national ne considèrent pas l’augmentation des

charges d’opération ni l’accroissement de la détérioration des ouvrages d’art pendant leur vie de

service. Bien que ces normes mettent l’accent sur le contrôle de la qualité pendant la sélection

des matériaux de construction, la conception et la construction, elles ne fournissent pas de

directives ni des outils scientifiques pour faire la conception et l’entretien des structures pour

atteindre une durabilité spécifique selon la vie de service requise. Ces normes incluent seulement

des outils prescriptifs pour prévenir quelques modes de détérioration.

Ce programme de recherche fait l’intégration des principes du développement durable avec la

conception classique des structures de ponts dans un pays de climat froid, soumis à différents

charges mécaniques et environnementales d’origine naturelle et artificielle. Il considère aussi la

performance des différents matériaux de construction et composants structuraux pendant la vie

de service du pont. Les plus récents modèles disponibles concernant les modes de détérioration

de matériaux de construction, ont été incorporés dans les considérations de la vie de service et la

conception de la structure. La procédure de base adopte une stratégie de protection multiple

contre tous les modes de détérioration qui résultent des actions environnementales agressives.

Elle intègre les considérations de durabilité avec les calculs de conception structurale d’une

manière itérative jusqu’à l’identification de la conception définitive. Cette procédure inclut

l’utilisation des mesures de protection supplémentaire, ainsi que la définition des stratégies

d’entretien. La procédure de conception pour la durabilité est illustrée à travers un exemple

détaillé de conception d’un pont.

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ACKNOWLEDGEMENTS

The author would like to express his deepest gratitude to Professor M.S. Mirza, McGill University,

for his guidance and advice throughout the preparation, development, improvement and

publication of this thesis. Professor Mirza has been a source of valuable encouragement, support,

constructive criticism and inspiration since the early stages of the research program. The author

would like to acknowledge the enthusiasm, encouragement and interest of Professor Mirza to

make public the progress and results obtained from this thesis.

The author would like to sincerely thank Professor Andrew, J. Boyd, McGill University, for his

advice and discussions related to the durability of materials aspects of this research work. The

author would like to thank the Civil Engineering and Applied Mechanics Department Staff for their

help and assistance during the research program.

The author has a profound gratitude to Mr. Guy Maurel, Mr. Santiago Saenz, and the rest of the

Bridge Engineering Design Team at AECOM, for their support and encouragement throughout

the development of this research work. The author would like to express his sincere appreciation

and gratitude to Pierre Nadon, François Charbonneau and the rest of the Bridge Engineering

Division at GENIVAR, for their constant support and the opportunity given to the author to

continue his career in the company as a Civil Engineer.

The author would also like to thank Professors Jorge I. Segura Franco and Carlos Iván Gutiérrez,

Universidad Nacional de Colombia, for their advice and encouragement since the beginning of his

career as a Civil Engineer.

The author would like to express his deepest love and sincere appreciation to his family. He

would like to express his heartfelt thanks to his parents Carlos Macia and Martha Niño, who

taught him that training, hard work and persistence are the key elements to accomplish all the

goals and projects that are fixed in life. Above all, the author is deeply grateful to his wife Maria

Luisa Reyes. Her unconditional love, patience, encouragement and support made possible the

completion of this project. The author affectionately dedicates this thesis to his wife and

companion.

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TABLE OF CONTENTS

ABSTRACT i

RÉSUMÉ ii

ACKNOWLEDGEMENTS iii

TABLE OF CONTENTS iv

LIST OF FIGURES ix

LIST OF TABLES xii

1. INTRODUCTION 1

2. GENERAL OVERVIEW OF THE BRIDGE PROJECT 2

2.1. Description of the structure 2

2.2. Structural system 5

2.3. Foundation system 6

3. PRINCIPLES OF DURABLE AND SUSTAINABLE CONCRETE BRIDGE DESIGN 7

3.1. Sustainable Development in Bridge Engineering 7

3.2. Durability Concerns about Reinforced Concrete Bridges 8

3.3. Durability of Construction Materials 9

3.4. Review of the Design and Construction Practice of Bridge Structures 10

3.5. Lessons from Previous Experiences 12

3.6. Durability Design Approach 14

3.7. Service Life Definition of Reinforced Concrete Bridges 17

3.8. Life Cycle Cost Analysis for Evaluating the Sustainability of Reinforced Concrete Bridges 18

4. SERVICE LIFE OF THE STRUCTURE 20

4.1. Required service life 20

4.2. Durability design formulation 20

4.3. Determination of lifetime safety factor 21

4.4. Design service life 24

5. BRIDGE DESIGN FOR DURABILITY 26

v

5.1. Durability design aspects 26

5.1.1. Identification of macro-climatic conditions 26

5.1.2. Identification of micro-climatic conditions 29

5.1.3. Environmentally induced mechanisms of deterioration 32

5.1.3.1. Frost attack 32

5.1.3.2. Abrasion of concrete by ice 33

5.1.3.3. Surface deterioration 34

5.1.3.4. Chloride-induced corrosion 35

5.1.3.5. Carbonation-induced corrosion 41

5.1.4. Minimum required conditions of the main construction materials 42

5.1.4.1. Minimum concrete cover 50

5.1.4.2. Type of steel 51

5.2. Concrete mixture design for durability 52

5.2.1. Concrete mixture requirements 52

5.2.2. Base materials specifications 53

5.2.3. Concrete mixture proportions 54

5.3. Concrete Handling, Placing and Curing 56

5.3.1. Concrete handling 56

5.3.2. Placing and finishing of concrete 56

5.3.3. Curing 56

6. STRUCTURAL DESIGN FOR DURABILITY 58

6.1. Materials properties 58

6.2. Superstructure design 59

6.2.1. Prestressed concrete girder design 60

6.2.1.1. Bridge deck parameters 62

6.2.1.2. Composite section 62

6.2.1.3. Diaphragms 63

6.2.1.4. Wearing surface 63

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6.2.1.5. Barriers 63

6.2.1.6. Load analysis 63

6.2.1.7. Prestressing steel 67

6.2.1.8. Ultimate limit states 72

6.2.1.9. Serviceability limit states 74

6.2.1.10. Girder design for the required service life 76

6.2.1.11. Girder performance with time 79

6.2.1.12. Supplementary protection measures 84

6.2.2. Reinforced concrete design for durability 85

6.2.2.1. General design parameters 85

6.2.2.2. Flexural design for durability 87

6.2.2.3. Check for shearing resistance 88

6.2.2.4. Deflection analysis 89

6.2.3. Deck slab design 92

6.2.4. Transverse bending moments in the bridge deck 92


6.2.4.2. Durability Parameters 94

6.2.4.3. Initial Conditions of the Bridge Deck Slab 94

6.2.4.4. Assumption for steel reinforcement and performance with time 94

6.2.5. Transverse bending moments in the cantilever overhang 96




6.2.5.4. Assumption for steel reinforcement and performance with time 98

6.2.5.5. Final design details 99

6.2.6. Transverse vertical shear 99



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6.2.6.4. Shearing resistance performance with time 101

6.2.7. Analysis of deflection with time 102

6.2.8. Design of the semi-continuity of the bridge deck slab 103


6.2.8.2. Durability parameters 105

6.2.8.3. Initial conditions of semi-continuity of the bridge deck 105

6.2.8.4. Semi-continuity performance of the bridge deck with time 107

6.2.9. Supplementary protective measures 108

6.3. Substructure design 108

6.3.1. Pier column design 109


6.3.1.2. Durability parameters 121

6.3.1.3. Initial conditions of the pier columns 121

6.3.1.4. Assumptions for reinforcement and performance with time 121

6.3.1.5. Supplementary protective measures 124

7. MAINTENANCE STRATEGIES 125

7.1. General principles for maintenance strategies 125

7.2. Design for maintainability 125

7.3. Preventive maintenance 125

7.3.1. Bridge inspection 126

7.3.1.1. Routine inspection 126

7.3.1.2. Detailed inspection 127

7.4. Corrective maintenance 129

8. SUMMARY AND RECOMMENDATIONS 130

8.1. Bridge design for durability 130

8.2. Holistic design approach 131

8.3. Basic conclusions and recommendations 132

viii

REFERENCES 134

APPENDIX 1 – DURABILITY CALCULATIONS 136

APPENDIX 2 – GIRDER DESIGN CALCULATIONS 140

APPENDIX 3 – MASS LOSS OF STEEL CAUSED BY CORROSION 141

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LIST OF FIGURES

2.1. Bridge cross-section A-A.

2.2. Plan view of the bridge.

2.3. Elevation view of the bridge.

2.4. Typical detail of the semi-continuous girders at the pier support.

2.5. Overview of the foundation system of the bridge at the piers.

3.1. a) Semi-probabilistic modelling of resistance and load effects. b) Increase of failure probability in time.

3.2. Current and proposed engineering practices.

3.3. Flow chart of the durability design procedure.

3.4. Qualitative cost flow diagram for the present worth of a bridge.

4.1. Relationship between mean service life and target service life.

4.2. Lifetime safety factor in terms of degradation process.

5.1. Montreal climate graph (Altitude: 57m).

5.2. Ice accretion in the Montreal metropolitan area.

5.3. Frost Index for southern regions of Quebec.

5.4. Microclimates on the bridge deck.

5.5. Microclimates of the bridge at the intermediate pier.

5.6. Diagram of the two limit states of corrosion in reinforced concrete elements.

5.7. Mass loss of reinforcement vs. time due to chloride-induced corrosion.

5.8. Response for deterioration mechanisms for: a) Bridge deck slab, b) Barriers.

5.9. Response for deterioration mechanisms for: a) Piercap, b) Abutments.

5.10. Response for deterioration mechanisms for: a) Pier columns, b) Caissons.

5.11. Response for deterioration mechanisms for: a) Edge girders, b) Internal girders.

5.12. Chloride-induced corrosion progress in the different members of the bridge over a stipulated service life of 150 tears.

5.13. Progress of carbonation front in the bridge.

6.1. NEBT girder characteristics.

6.2. Geometrical parameters of the composite section.

x

6.3. a) CL-625 design truck clearance envelope. b) CL-625 and CL-W design truck loads. c) CL-625 and CL-W design lane loads.

6.4. Definition of the vehicle edge distance, Dve.

6.5. Non-prestressed reinforcement of the bridge girder.

6.6. Prestressed reinforcement of the bridge girder.

6.7. Localized deterioration on the lower flange of a bridge girder.

6.8. Representation of the degradation of the composite slab-girder section.

6.9. Variation of internal compressive stresses at the top of the composite section vs. time.

6.10. Variation of internal tensile stresses at the bottom of the composite section vs. time.

6.11. Loss of flexural resistance of the composite section vs. time.

6.12. Loss of shearing resistance of the composite section vs. time.

6.13. Increment of the static deflections of the composite section caused by the live load.

6.14. Increment of the stress variation on the prestressing strands of the girder.

6.15. Reinforced concrete section parameters for the structural and durability design.

6.16. Cracked transformed concrete section for flexural analysis.

6.17. Representation of the compatibility of stresses and deformations.

6.18. Bridge deck slab resistance for positive bending moments vs. time.

6.19. Bridge deck slab resistance for negative bending moments vs. time.

6.20. Detail of the deck slab reinforcement.

6.21. Notation for cantilever moments.

6.22. Bridge deck slab flexural resistance at the cantilevers overhang.

6.23. Reinforcement details for the bridge deck slab overhangs.

6.24. Axle loads moving across the deck lanes.

6.25. a) Influence line for shear forces in the slab at the intermediate support. b) Shear force diagram for the loading location that generates the largest shear force at the intermediate support.

6.26. Bridge deck slab shearing resistance vs. time.

6.27. Deflections on the cantilever overhang of the bridge deck slab.

6.28. Bending moments diagram for the bridge deck under permanent loads.

xi

6.29. a) Influence line for bending moments at the firs intermediate support of the bridge deck and location of the truck loads that generates the maximum negative bending moment at the support. b) Bending moments diagram for the bridge deck under the truck loading condition.

6.30. Detail of the bridge deck reinforcement for the semi-continuity condition at the intermediate supports.

6.31. Bridge deck bearing capacity for negative bending moments with time.

6.32. Reinforcement of the pier column.

6.33. Interaction curves for pier column for abrasion by ice.

6.34. Interaction curves for pier column for frost attack.

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LIST OF TABLES

4.1. Lifetime safety factors determined by a normally distributed degradation function.

5.1. Depth of frozen soil in terms of frost index.

5.2. Microclimates, exposure conditions and mechanisms of deterioration of the different structural members of the bridge.

5.3. Environmental coefficient values for frost attack.

5.4. Environmental coefficient values for surface deterioration.

5.5. Rate of corrosion at anodic areas in carbonated and

5.6. Rate of corrosion at anodic areas in carbonated and chloride-contaminated concrete (Tuutti, 1982).

5.7. Corrosion rate found by Andrade et al (1994).

5.8. Required development lengths for corroding steel reinforcement in various concrete mixtures.

5.9. Environmental coefficient for carbonation-induced corrosion.

5.10. Air content coefficient for carbonation-induced corrosion.

5.11. Parameters a and b.

5.12. Durability parameters for the various bridge components according to the different modes of deterioration.

5.13. Rates of deterioration, initiation time for corrosion and carbonation coefficients for the various bridge components.

5.14. Concrete cover thicknesses for the analyzed members of the bridge.

5.15. Detailed proportions for the concrete mixture designs.

5.16. Summary of the concrete mixture proportions for the bridge members.

6.1. Material properties for the elements of the bridge.

6.2. Conditions for use of the simplified method of analysis for dead loads.

6.3. Conditions for use of the simplified method of analysis for live loads.

6.4. Bending moments on the girder produced by the different load cases.

6.5. Induced internal actions caused by prestressing.

6.6. Compressive stresses of concrete in the girder at transfer.

6.7. Compressive stresses of concrete in the girder at SLS.

xiii

6.8. Stress levels inside composite section.

6.9. Shearing forces cause by the different load cases.

6.10. Shear reinforcement for the girder.

6.11. Final deflection caused by permanent loads.

6.12. Geometrical properties of the NEBT1600 girder of different steps of deterioration.

6.13. Durability design parameters of the bridge deck slab.

6.14. Initial conditions of the bridge deck slab.

6.15. Maximum cantilever moments due to unfactored CL-625 truck wheel loads including the DLA (kNm/m).

6.16. Initial conditions of the bridge deck at the intermediate supports.

6.17. Durability design parameters for the pier columns.

6.18. Initial conditions of the reinforced concrete section of the pier columns.

INTRODUCTION

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1. INTRODUCTION

The main objective of this research is to establish the necessary steps to carry out a concrete

bridge design for durability based on the integration of durability and sustainability principles to

the standard structural design practices. The bridge design example will involve the detailed

design procedure for some key elements of the bridge structure such as bridge girders, bridge

deck slab and intermediate foundation units.

The purpose of this durability design is to establish how the construction materials and the

various bridge components will behave over time under certain and specific environmental

conditions, and thereby determining the overall performance of the structure over its service life.

This design procedure takes into account a rational understanding of how the construction

materials deteriorate by different mechanisms. It integrates different fields of engineering that

concern the durability of a reinforced concrete structure, including construction materials

engineering, structural design, construction practices, durability and sustainability concepts, life-

cycle costing, bridge management and maintenance strategies. It is necessary to recognise that

durability can only be achieved through a holistic design approach.

The durability design of the structure incorporates multiple protection strategies against the

different actions that may cause deterioration on the structure. The key steps that are considered

within the design-for-durability procedure include: determination of the service life of the structure,

analysis of the environmental effects, identification of the mechanisms of deterioration, selection

of adequate calculation models and determination of durability parameters, design and selection

of good-quality construction materials, integration between durability parameters and structural

calculations to define the final design, the identification of supplementary protection measures,

and the definition of the maintenance strategies.

This concrete bridge design for durability is proposed to guide the engineer to implement the

essential durability and sustainability principles into the current bridge engineering practice. The

design procedure is based on the fact that higher initial investments during the design and

construction of a bridge project will require lower maintenance, renovation and reparation costs

during its future operations. In addition, higher levels of maintenance will represent more durable

and hence, more sustainable structures.

GENERAL OVERVIEW OF THE PROJECT

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2. GENERAL OVERVIEW OF THE BRIDGE PROJECT

2.1. Description of the structure

The proposed bridge structure crosses a river and connects an intermediate town with a major

highway system. The town is located near the metropolitan region of Montreal, Quebec, Canada.

The total width of the bridge is 15.07m including four traffic lanes, two lanes for each direction,

two shoulders of 1m each, and one concrete barrier placed on both sides. The bridge has three

spans of 26m, with intermediate semi-continuous supports configured by the pier and bridge deck

arrangements. The bearing axes are parallel to each other and perpendicular to the axis of the

bridge.

The general overview of the bridge project is shown in Figures 1.1, 1.2 and 1.3.

Figure 2.1: Bridge cross-section A-A .


3

Figure 2.2: Plan view of the bridge.


4

Figure 2.3: Elevation view of the bridge.


5

This bridge is an important link between the town and the highway system. It represents a vital

corridor not only for the regular transportation of residents and visitors, but also for goods,

services and merchandise. This bridge is a lifeline structure and a part of the emergency system

which must remain operational during and after a mayor event, such as an earthquake.

Once the bridge is constructed and open to service, it is estimated an immediate traffic between

1000 and 4000 daily vehicles per lane using the bridge; this which is equivalent to a traffic

between 250 and 1000 daily trucks per lane. However, it is expected that the previous numbers of

daily vehicles and trucks will be exceeded within the bridge service life. For this reason, the

bridge will be designed for a daily traffic of more than 4000 vehicles, or 1000 trucks. Accordingly,

it is possible to classify this bridge to be part of a highway Class A (Clause 1.4.2.2)1.

The river that passes under the bridge is wide but shallow, and only small boats use it. However,

all bridge elements prone to vessel collision will be designed for such forces. The bridge must

remain functional and open to traffic after the incidence of abnormally unspecified loads such as

earthquakes, vessel collisions, floods, fires, and wind loads (among others). Accordingly, the

bridge can be classified as a lifeline bridge (clause 4.4.1)1.

2.2. Structural system

The structure consists of three sets of seven simply-supported girders, arranged in three spans.

The intermediate supports at the piers are designed to provide the necessary negative

reinforcement at the top fibres as well as a continuity of the bottom “fibres”.

Figure 2.4: Typical detail of the semi-continuous girders at the pier support.


6

Cast-in-place diaphragms are provided at the foundation units (girder ends) and at intermediate

location along the girder spans to ensure satisfactory transverse distribution of applied loads.

Additionally, some high-stress strength dowels are embedded at the middle of the pier along the

pier axis to provide lateral restraint to improve the continuity of the girders and to ensure the

condition of semi-continuity over the piers (see Figure 1.4).

The foundation units are basically defined by two intermediate rigid-frame piers 2, and two open-

end 3, cantilever abutments 4. These foundation units are reinforced concrete elements.

2.3. Foundation system

The foundation units are supported by a series of circular caissons that transfer the loads to the

bedrock. The number and the diameter of these shafts, as well as their arrangement are

designed for each one of the foundation units. This kind of foundation is chosen because of the

nature and characteristics of the bridge, being able to sustain effectively large compressive and

lateral forces. Additionally, they represent a feasible solution for this structure that overpasses a

river, where the caissons can be bored down through the soil strata to a depth where the bedrock

can be reached. The condition of end-bearing caissons involves a distribution of pressures at the

bearing stratum that provides a good resistance against overturning moments.

Figure 2.5: Overview of the foundation system of the bridge at the piers.

PRINCIPLES OF DURABLE AND SUSTAINABLE CONCRETE BRIDGE DESIGN

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3. PRINCIPLES OF DURABLE AND SUSTAINABLE

CONCRETE BRIDGE DESIGN

3.1. Sustainable Development in Bridge Engineering

The concept of sustainability can be related to reasonable use of natural resources to provide

important benefits to the society without compromising their supply for future generations.

Sustainable development in civil engineering is focused on the efficient use of natural resources

and energy at the planning, design, construction and operation stages of a project. This efficiency

is dependant on the positive impact of the structure on the community 7. This goal can be

achieved by emphasizing conservation measures, use of renewable resources, waste reduction,

recycling of used materials and elements, and the preparation of more complete environmental

and economic assessments, using valuable tools, such as life-cycle cost analysis.

A complete analysis of the life-cycle costs of the project is essential to ensure safety and

serviceability of the bridge during its design life. This analysis involves not only the initial costs of

design and construction, but also the future costs of maintenance, repair, rehabilitation, and

decommission of the structure at the end of its service life. The design of a durable structure,

complying with the principles of sustainability requires optimization of the design alternatives to

identify the option with the lowest life-cycle costs.8

As long as the bridge is well-designed and performs correctly during its desired service life, the

positive impact on society will be attained. The design of a bridge must be engineered to ensure

that the structure doesn’t attain any of the possible limit states (ultimate, serviceability, functional,

economic, environmental, and other limit states) during the specified service life of the structure.

Past experience shows that, when the performance of the bridge declines early, additional

amounts of resources are required for repair and rehabilitation of the system to upgrade its

performance to an acceptable level. In this case, the concept of sustainability is normally not

respected because of the excessive use of natural resources. According to this, it is possible to

establish that the sustainable performance of a bridge is proportional to its durability.

The inclusion of durability aspects in the structural design process of a bridge requires the

consideration of different models of the possible mechanisms of deterioration that may be

initiated by the environmental conditions (microclimate and macroclimate) present at the bridge.

The concepts of reducing, reusing and recycling (the basic three R’s in sustainable development)

8 must be present in all phases of the project. Reducing involves a reduction of the consumption

of natural resources and a reduction of waste generation. This implies that the construction of

new structures must be undertaken only when it is strictly necessary. Reusing implies that any


8

new element on the structure may be useful and applicable for the functions in other structures

with other characteristics. For instance, it is possible to mention the reuse of some of the

“composing elements” of the bridge, such as precast girders, precast panels for reinforced-earth

retaining walls, expansion joints, and bearings mechanisms (among others). Recycling means

that some of the elements of the bridge structure can be transformed or processed to create new

elements or components with different characteristics.

Only recent colossal transport infrastructure projects like the Confederation Bridge in Canada, the

Second Gateway Bridge in Australia, and the Great Belt Link in Denmark have been developed

following a coherent synchronization between the structural design procedure and the principles

of durability and sustainability with the purpose of attaining an impressive service life of more that

100 years. These projects have had major impacts in their countries. The synchronization of

criteria in these projects necessitate proper selection and design of the construction materials, the

use of high quality construction practices, the identification and detailing of critical parts, the

definition and implementation of an effective maintenance program and the monitoring of the

performance of the structural components.

It is essential to integrate sustainability and durability principles in the practice of structural design.

This involves being able to visualize how construction materials will perform over time under

specific and aggressive environmental conditions, and consequently defining the overall

performance of the structure over its design service life. The new approach to bridge design

requires significant advanced planning, rethinking and retroactive improvement of possible

solutions before the project is defined for execution.

3.2. Durability Concerns about Reinforced Concrete Bridges

Durability and sustainability issues have not been taken into account seriously in bridge design

practice over the last decades in North America and in many other regions of the world. This has

led to premature deterioration of bridges, which has become a serious problem for all nations

around the globe, creating immense social, economic and environmental impacts. This early

deterioration of infrastructure is directly associated to some aspects, like deferred maintenance,

employment and conception of inadequate construction materials, ignorance of the environmental

conditions that may affect the bridge, bad construction practices, and poor workmanship during

the different phases of the bridge project.

In the past, bridge engineering practice was focused mostly on mechanical and structural

behaviour. Structural design and material selecion were developed in separate ways until the last

few years, leading to the appearance of flaws in the real-life performance of bridge structures.

From the structural design point of view, all considerations about the required quality of the

construction materials have been mostly prescriptive, based on a series of recommendations


9

found in different bridge design codes and manuals. On the other hand, considerable research

has been undertaken on construction materials in such a way that the results and conditions of

the test are not representative of the field conditions of the “actual” bridge structures; this has

resulted in inaccurate recommendations for some bridge designs.

Presently, adequate knowledge is available for the different modes of deterioration that may

occur on bridge structures. However, the provisions adopted by different national codes about the

necessary actions to ensure durability of construction materials against the degradation imposed

by the aggressive environmental conditions vary significantly. Nonetheless, most of the provisions

in the codes concur in two major aspects, the identification of the local exposure conditions

(microclimate), and the necessity of low concrete permeability, a property that can be reached

following the concept of the four C’s for each type of environmental exposure: constituents of

concrete mixture, good-quality concrete cover thickness, compaction, and curing.8

3.3. Durability of Construction Materials

Satisfactory performance of reinforced concrete structures can be attained by a thorough

evaluation during the design procedure of all possible scenarios involving all possible loading

cases and environmental conditions that may affect the structure, excellent quality control during

construction, use of protective measures for the most critical structural elements against the

exposure to severe environmental conditions, planning and implementation of effective

maintenance strategies, and as an essential component, the design and employment of high-

quality construction materials. This means the use of construction materials and techniques that

require relatively low maintenance while they provide the required strength properties and

durability characteristics over the design service life of the structure. In the case of reinforced

concrete bridges, the level of performance of the structure depends directly on the quality control

exercised in design, construction and maintenance of the structure during operation.

When a reinforced concrete element is exposed to severe environmental conditions, such as

constant moisture, direct contact with chloride solutions, high temperature ranges, and wetting

and drying cycles, it is essential to provide and protect reinforcing steel to withstand these

conditions without generating active corrosion at early ages of the structure, or at least, being

able to develop corrosion rates slow enough to ensure that this degradation mechanism will not

affect the required performance during the design service life. Nowadays, some of the most

feasible solutions generally used in the construction industry is the use of galvanized steel and

epoxy-coated steel, provided that they are properly detailed and specified in the detailed

drawings and technical recommendations of the project. The use of stainless steel, although it is

more expensive, has also proved to be effective in mitigating or considerably delaying corrosion

of steel embedded in concrete. Additionally, there is a growing interest and research in the use of


10

fibre-reinforced composites to replace reinforcing steel; however, special care must be taken

during the design process to ensure the fulfillment of the different required limit states during its

service life.

The quality and level of performance of concrete depend strongly on the microstructure of the

hydrated cement paste (hcp), which involves the hydration products, the pore structure, and the

free and bound water on the pore walls. After the initial set, the concrete quality is governed

mostly by hcp, which is the one that controls low capillarity (low permeability), bonding with the

aggregates, high pH (around 13), production of calcium-silicate-hydrate gel, and the resistance to

the variations in humidity and temperature conditions. These properties are developed during the

various stages of concrete preparation and placing, and are controlled strongly by curing and

other on-site conditions. Under appropriate conditions, the water-tightness and strength of

concrete will increase due to the continuous hydration of the un-hydrated cement particles in the

concrete mixture. When concrete is fresh, the aggregates help to define some characteristic,

such as workability and segregation during placing. In addition, when concrete has hardened, the

aggregates help to increase the stiffness and stability of concrete. The concrete mixture needs to

be carefully designed for all of the above factors or characteristics, and mixed, placed,

consolidated and cured using high quality control through on-site supervision 8, and follow-up with

an excellent maintenance program through its service life.

3.4. Review of the Design and Construction Practice of Bridge Structures

It is important to take review of the past and current construction practices to understand, control

and mitigate the principal causes of premature deterioration in present reinforced concrete

infrastructure facilities.

Several mechanisms of deterioration are closely related to the ability of aggressive elements and

substances to penetrate into reinforced concrete elements. Therefore, deterioration is greatly

associated with permeability, which depends on factors, such as pore system interconnectivity

and cracking. Above certain crack widths (0.15 to 0.18mm) 9, aggressive substances can easily

ingress into the concrete structure, interacting with different elements and activating different

modes of deterioration. There are many different reasons for cracking in concrete; one

predominant factor which influences cracking at early age is the use of high-early strength

cements and concrete mixtures necessary to facilitate the high speed of modern construction.9

During the first decades of the twentieth century, concrete was produced using coarsely ground

cement, which caused the hydration process to be very slow, generating low heat of hydration,

and enabling concrete to develop strength at a relatively small rate. Consequently, thermal

cracking of concrete elements was not a problem. After the 1930’s, the increased fineness of


11

cement introduced new problems due to an increment of the rate of hydration and heat

generation, making it harder to adequately cure the concrete, and promoting early age cracking.

After the 1950’s, significant changes were introduced in concrete construction practices, including

the development of the ready-mixed concrete industry, the placement of concrete by pumping,

and the vibration procedures, using mechanical vibrators. During this period, achieving high

strength at early ages and providing more fluidity and versatility to concrete, engineers decided to

augment water content in the concrete mixture, increasing cement fineness and C3S content,

which is the component of cement responsible for early strength, normal setting, temperature rise

during hydration, and major contribution to creep and shrinkage. As a result of these changes,

concrete tended to be more permeable and consequently less durable.

During the 1970’s, a wide range of water-reducing admixtures were introduced. Having identified

the high w/c ratio as a clear factor responsible for the low durability of concrete, it was believed

that reducing the water content in the concrete mixture could improve the durability of concrete.

However, it must be emphasized that this factor is not the only one that determines the

performance of hardened concrete. The type of cements that were developed and used to

achieve the fast schedules of the construction industry produced concretes with high thermal and

drying shrinkage, low creep, and high elastic modulus at early ages. These concrete mixtures

were prone to cracking and prone to lack of durability.

Since the 1980’s, an increased use of water-reducing admixtures and the introduction of mineral

admixtures like puzzolanic materials, produced concrete mixtures able to develop high workability

at very low water-cementitious materials (w/cm) ratio, generating high compressive strengths and

high elastic modulus at early ages, and very low permeability in laboratory specimens. These

kinds of concretes were termed “high-performance concretes”. However, these very high

strengths and moduli were accompanied by a significantly reduced creep potential and high

brittleness, resulting in significant cracking at different ages of concrete, especially for those

elements subjected to important cyclic loading conditions, such as bridge decks.

From the structural design point of view, most of the current bridge design standards are based

on semi-probabilistic considerations of mechanical loads and resistances, considering various

limit states that the bridge structure fulfill during its service life (Figure 3.1a). This consideration

does not take into account for the future increase of loads and the deterioration of the structure

expressed in terms of loss of resistance of the construction materials. Consequently, the

probability of failure increases considerably over the service life of the structure, resulting in

premature deterioration with huge impacts on the surrounding environment (Figure 3.1b). 8

The Canadian standards provide detailed guidance to address the design of safe and serviceable

concrete structures in terms of structural mechanics, putting a special emphasis on certain


12

phenomena that are not as well understood, such as shear and torsion among others. However,

the different aspects related to durability and sustainability are not explained or introduced using

the same scientific and rational approach as used for structural analysis and design. The codes

deal with these subjects by providing some prescriptive requirements for durability of construction

materials, including the use of de-icing chemicals, minimum concrete covers, type of reinforcing

steel, chloride ion content, sulphate attack, and freezing and thawing cycles, among others. The

Canadian standards emphasize the need to employ high-quality concrete and steel. Additionally,

they indicate a minimum service life of 75 years that the bridge structures must provide. However,

the codes do not show clearly how to attain this service life by assuring enough durability of a

properly designed, constructed and maintained structure.

Figure 3.1: a) Semi-probabilistic modelling of resistance and load effects. 10

b) Increase of failure probability in time.5

3.5. Lessons from Previous Experiences

The experiences related to the evolution of bridge engineering practice throughout the 20th

century and the early years of the 21st century can be highlighted as follows:

As soon as the gain of strength became a priority in the construction industry, new deterioration

problems were introduced in bridge structures. Higher elastic moduli and resistances at early

ages, lower creep and higher thermal and drying shrinkages made modern and high performance

concrete more prone to cracking. This significant cracking of concrete at early ages was closely

connected to premature deterioration of reinforced concrete elements exposed to severe

aggressive environments.

Significant advances in the research about durability of construction materials have resulted from

some excellent laboratory work. Nevertheless, some conditions of the tests that simulate the

degradation mechanisms do not reflect the “actual” conditions on “real” bridge structures.


13

Therefore, some of the prescriptive measures adopted by the codes were ineffective when used

in the design and construction procedures.

These facts highlight the real deficiencies in the procedures and considerations regarding

durability and sustainability aspects in design and construction practices. Additionally, there was

a strong need to better study and analyze how the construction materials deteriorate by different

mechanisms. The most important change needed was to integrate the different fields of

engineering that concern the durability of a reinforced concrete structure, including construction

materials, structural engineering and construction quality. It is necessary to recognize that

durability can be achieved only through a holistic design approach.

A reduction of the speed of construction is needed to accomplish the goal of building durable and

sustainable structures. Additionally, some changes are required in the approaches adopted by

owners, builders, and designers in their professional practices. Any bridge project must seriously

take into account durability, economic and environmental considerations from the early stages of

the project, during the construction of the bridge, and maintained diligently throughout the entire

service life. Figure 3.2 shows the current engineering practices and how emphasis on some

related issues must change to attain sustainable and durable structures.

Figure 3.2: Current and proposed engineering practices.11

Some construction materials, such as steel, concrete and wood, are currently being produced at

a great cost to the environment; consequently, the effective use and the conservation of materials

must become a new priority in the construction industry. Accordingly, standard specifications

need to be changed from prescriptive to performance-based recommendations for materials. 9


14

3.6. Durability Design Approach

A new design procedure for bridge structures must achieve safety, serviceability, durability, and

general socio-economic and environmental benefits for the surrounding communities throughout

its specified service life. The basic steps involved in durability design are summarized in Figure

3.3.

Figure 3.3: Flow chart of the durability design procedure.12

During analysis and design stages, it is essential to consider the different load cases that may

affect the structure during its service life, including not only the conventional mechanical actions,

but also all possible environmental effects that may be developed globally and locally on the

structure. A semi-probabilistic analysis has been use in the various national standards and codes

to derive specific deterministic safety factors in terms of materials and load cases to ensure that


15

the specified limit sates are not exceeded. Moreover, there is a need to define a lifetime safety

factor for the bridge considering all the parameters involved on the definition of the design service

life and the target service life. The bridge is considered to attain its service life when the

maximum permissible level of deterioration and damage on the structure is reached, or when the

bridge is unable to provide service with an adequate level of safety and serviceability.

The durability design of the structure can be conceived by following required protection strategies

against the different actions that may cause deterioration of the structure. One of them can be

avoiding all possible degradation modes, another one may include the provision of a single

barrier of protection, and the other one involves the use of several shields of protection.

An essential part of the durability design process must be based on a life cycle cost analysis

approach, which must involve all costs of the project at its different life stages, including planning,

design, construction, maintenance, repair, rehabilitation and decommissioning. Additionally, the

possible socio-economic and environmental costs that correspond to the each one of these

stages during service life must be analyzed and weighted with the purpose of optimizing and

identifying the best design option that minimizes the overall life cycle cost of the bridge project.

Despite the fact that the bridge design for durability can be carried out by integrating the results of

ordinary mechanical design and durability design performed separately, the design approach can

be established in a more efficient way by combining the traditional structural design with the

durability parameters, to avoid overdesign, and unnecessary overuse of precious construction

materials. The key steps that need to be considered within the design-for-durability procedure of

a reinforced concrete bridge are briefly reviewed. 12

- Determination of the target service life and design service life

The target service life for the structure must be determined following the requirements

provided in the current regulations, standards and codes, according to the type of the

structure and the special needs of the client. The definition of the service life for the bridge

structure which is the subject of this work is discussed in Chapter 4.

- Analysis of the environmental effects

It is essential to analyze the environmental effects that define global and local conditions

acting on the different members of the structure, which create the macro- and microclimates

responsible for generating the different modes of degradation of the construction materials

and the bridge elements. It is essential to identify climatic conditions, such as temperature

and moisture fluctuations, rain, moisture condensation, freezing, solar radiation, and pollution.

It is also necessary to identify some geological conditions such as the presence of ground


16

water, sea water and contaminated soil. Additionally, all man-made activities that generate

specific microclimates on the structure, like the use of de-icing salts and abrasion by traffic,

need to be considered, along with any man-made hazards, such as terrorist activities aimed

at destroying the structure.

- Identification of the mechanisms of deterioration

A sound knowledge of probable environmental effects will enable identifying the different

mechanisms of deterioration that may take place in different parts of the structure. The

principal mechanisms of deterioration which may produce long-term effects on the bearing

capacity of the structure include chloride-induced corrosion, carbonation-induced corrosion,

mechanical abrasion, salt weathering, surface deterioration, and frost attack, among others.

These mechanisms of deterioration affect either concrete, or reinforcing steel, or both. They

destroy the material progressively from the outside towards the interior, causing a reduction

of the cross-sectional area of concrete and steel, loss of bond between the concrete and the

steel, splitting and spalling of concrete cover due to progressive corrosion of reinforcement,

and a generalized loss of bearing capacity of reinforced concrete elements.

- Selection of adequate calculation models and determination of durability parameters

The bridge engineer must make a careful selection of the various degradation models that

describe the mechanism of deterioration that affect the structure. It might be necessary to

make a preliminary evaluation of the rates of degradation for the different mechanisms to

accurately assess the service life of the structure. It should be noted that these models can

be readily substituted by newer models, developed using semi-probabilistic approaches.

Using the selected calculation models, it is possible to determine some durability parameters,

such as concrete cover thickness, durability of materials, detailing and supplementary

protective measures.

- Integration between durability parameters and mechanical calculations to define the

final design.

By integrating durability and mechanical calculations, it is possible to visualize the rate of the

loss of bearing capacity of the various components and the structure due to the degradation

caused by the various defined factors. Accordingly, an iterative improvement of the structural

characteristics of the bridge can be made, defining at the final stage, the minimum required

geometry for the structural members, the best reinforcement distribution, the quality of the

construction materials of the bridge, and the provision of supplementary shields of protection

for the critical members.


17

- Definition of the maintenance strategies

When the structural design for durability is completed, the engineer must conceive a

complete maintenance strategy that has to be followed during the entire service life of the

structure, executing some preventive, corrective and improvement activities in a periodic

manner. A complete description of the maintenance strategies for this bridge is presented in

Chapter 7.

3.7. Service Life Definition of Reinforced Concrete Bridges

The service life of a bridge and its components must be defined to respond to future needs or to

calculate life-cycle costs. In the past, service life was estimated by approximate projections

enhanced by judgement. Statistical analysis techniques of empirical data for durability of

materials are being used to develop mathematical models of the degradation modes. These

models are used to estimate deterioration projections that could be used to approximate service

life.13

Presently, to implement a durable and sustainable bridge design, engineers must explore outside

the current codes to evaluate possible environmental effects and ensure adequate material and

structural performance over the design service life. This requires an extrapolation of current

knowledge about climate, material properties as well as a projection of the models that describe

the different mechanisms of deterioration. The design, construction and maintenance teams are

required to go beyond the prescriptive durability requirements in the codes, to explore new ways

of attaining a required service life and to achieve higher levels of performance.7

Figure 3.4: Deterioration vs. time in terms of levels of maintenance expressed in terms of percentage of investment of the initial costs. 11


18

The economics of service life for civil infrastructure are based on the fact that higher initial costs

invested during the design and construction of a bridge project will generally involve lower

maintenance, renovation and reparation costs during its future operation. In addition to this,

higher levels of maintenance will represent longer periods of time for the structure to attain the

maximum acceptable levels of deterioration (Figure 3.4). For many cases, especially

infrastructure located in very congested urban areas, any replacement operation could represent

large costs, even many times higher than the initial costs, considering all possible socioeconomic

and environmental impacts, and great amounts of additional energy and natural resources,

thereby contradicting the principles of sustainable development.

3.8. Life-Cycle Cost Analysis for Evaluating the Sustainability of Reinforced Concrete

Bridges

The life cycle of a transportation infrastructure asset like a bridge regroups all the sequence of

events and actions involved during the life of the structure. These actions include the initial design

and construction, operation, monitoring and inspection, maintenance procedures, repairs, retrofit,

replacement, decommissioning and disposal. The life-cycle cost analysis of the project is a

process where the total economic worth of the bridge is evaluated by analyzing all initial costs,

and discounted future expenses related to the previously mentioned events and actions. The

present worth of the bridge project can be estimated using the following equation:

SpwfUCFRCICMCpwfFCPWnt

t

0

(3-1)

where:

FC = First (initial) cost

t = Time period of analysis (must be the same for all alternatives analyzed)

MC = Maintenance costs

IC = Inspection costs

FRC = Future rehabilitation costs

UC = Users costs (time delays, fuel consumption, user’s discomfort, vehicle operating costs,

accidents)

S = Salvage values costs (difference between salvage and decommissioning costs)

i = Discount rate

n = Number of years when cost (benefit ) will occur

pwf = Present worth factor = ni11


19

This can be clearly seen in the following cost flow diagram:

Figure 3.5: Qualitative cost flow diagram for the present worth of a bridge.

A detailed estimation of service life is essential to accurately assess the life cycle costs and

impacts of a bridge structure, and to evaluate the advantages that are brought by new

technologies and advances in bridge design. This estimate must include schedules of repairs,

rehabilitation, or additional costs between initial construction and decommissioning of the

structure. This evaluation of service life and maintenance are essential when determining future

impacts of construction repairs produced by increases in traffic flow and financial discounting

during the operation of the infrastructure asset. Additionally it is a very useful tool to compare

different alternatives for the design of a bridge structure, including the type of structural system,

the type of bridge, the materials to be used, and the general dimensioning of geometric design for

the project.

SERVICE LIFE OF THE STRUCTURE

20

4. SERVICE LIFE OF THE STRUCTURE

4.1. Required Service Life

According to the Canadian Highway Bridge Design Code, the required target service life of new

structures must be 75 years (Clause 1.4.2.3) 1. This minimum overall service life of the bridge of

75 years implies that the most critical elements of the structure, such as foundations (abutments

and piers) and girders must perform satisfactorily over this period of time and must fulfill all of the

required limit states for the structure.

Other components less critical to the integrity of the structure may have a shorter service life,

being necessary to replace them as part of the maintenance strategy. Some of these components

may be the asphalt concrete pavement, waterproofing membranes, protective coatings for

concrete and steel elements, expansion joints, bearings, barriers, drains, slope protection

systems and others.

The durability design of this project aims the fulfillment of the service life using a scientific

procedure that combines durability models with the current bridge structural design practice. This

procedure allows the visualization of the performance of the structure over time.

4.2. Durability Design Formulation

The service life of a structure is affected by the macro- and micro-environmental conditions that

determine the type and severity of the degradation mechanisms of the materials that compose

the different structural members of the bridge. Additionally, the quality of the construction

materials and the degree of exposure of the different members to the aggressive conditions may

vary significantly. Therefore, the performance and service life of this kind of structure should be

preferably treated stochastically. This stochastic procedure takes into account the real nature of

structural performance to produce a reliable structural design. 5

The various formulae for load, resistance and service life are quite complex; since many

degradation factors affect the performance of the structure, the application of this methodology

become even more complex. Therefore, it is preferable to apply the lifetime safety factor method.

Despite the fact that this method is based on the theory of safety and reliability, the formulation of

the procedure results in a deterministic form. The design service life is determined by multiplying

the target service life by a lifetime safety factor (Sarja and Vesikari, 1996)5.

gtd tt (4-1)


21

where:

td = design service life.

tg = target service life.

t = lifetime safety factor.

4.3. Determination of lifetime safety factor

The lifetime safety factor is the relationship between the mean service life obtained from a

distribution of probable service life values of a structure, and the target service life that is required

for the design purposes of the project.

g

Lt t

t (4-2)

where:


(tL) = mean service life.

tg = target service life.

Figure 4.1: Relationship between mean service life and target service life.5

Figure 4.1 represents a distribution of service life values and the relationship between the target

service life, failure probability and mean service life. The mean service life can be identified by the

point of intersection of the degradation curve with the limit state of durability. To ensure that the

mean service life is greater than the target service life, the lifetime safety factor t must be grater

than one. The lifetime safety factor depends on the maximum allowable failure probability. Low


22

values of maximum allowable failure probabilities require larger values of lifetime safety factor,

which also depends on the form of the service life distribution. 5

The durability of a bridge can be expressed in terms of the degradation process that is present in

the structural system due to the degrading effect of the environmental loads that act on the

different structural members (Figure 4.2).

Figure 4.2: Lifetime safety factor in terms of degradation process.5

It is evident from Figure 4.2 that the point where the degradation curve intersects the maximum

degradation at the design service life, a time that must be longer than the target service life by a

factor represented by the lifetime safety factor t. The range Dmax – D(t) in Figure 4.2 represents

the safety margin. 5

The lifetime safety factor can be determined by a stochastic method assuming a normally

distributed degradation around the mean value, and a standard deviation of this degradation

being proportional to the mean degradation. Following these assumptions, it is possible to obtain

the following expression for the lifetime safety factor (Sarja and Vesikari, 1996)5:

nDt

11 (4-3)

where:


= reliability index.

D = coefficient of variation of degradation.

n = degradation rate exponent.


23

The Canadian Highway Bridge Design Code establishes that for the design of new bridges,

components must not fail suddenly and abrupt collapse of the structure must be avoided.

Accordingly, the lifetime target of the reliability index can be defined as 3.75 for most bridge

structures for the ultimate limit state (ULS). For the serviceability limit state (SLS), the target value

of the safety index can be selected as zero for an appropriate period (Clause C1.4.2.1) 6.

The required reliability index and the probability of failure Pf for ordinary design determined by

Eurocode 1 are: 5

For the ultimate limit state (ULS):

= 3.8 (serious consequences of a durability failure). → Pf = 7.2 x 10-5.

= 3.1 (no serious consequences of a durability failure). → Pf = 9.7 x 10-4.

For the serviceability limit state (SLS):

= 2.5 (serious consequences of a durability failure). → Pf = 6.2 x 10-3.

= 1.5 (no serious consequences of a durability failure). → Pf = 6.7 x 10-2.

Table 4.1: Lifetime safety factors determined by a normally distributed degradation function.


24

The degradation rate exponent n will affect the determination of the lifetime safety (Equation 4-3)

in the following way: 5

n = 1 → represents a linear degradation.

n = 0.5 → represents a retarding degradation.

n = 2 → represents an accelerating degradation.

The calculation of different values of the lifetime safety factors for different values of the

parameters previously mentioned are shown in table 4.1.

4.4. Design service life

Aggressive environments can dramatically deteriorate the integrity and performance of different

members of a bridge structure. This is the case of the chloride-induced corrosion, which once it

starts the active phase, the deterioration occurs quite rapidly.

Chloride- induced corrosion is perhaps one of the most critical deterioration mechanisms that can

be generated in most bridge structures of North-America where de-icing salts are usually

employed for traction during the winter seasons. Accordingly, the use of a degradation rate

exponent of the order of 2 could closely represent the behaviour of a reinforced concrete element

subjected to such aggressive environments.

Because a series of widely varied micro-climates can be developed at different parts of members

of the bridge, depending on the degree of exposure against certain environmental conditions, the

variation of the deterioration mechanisms can be considered to be elevated. For this reason, high

values of D can be adopted for the determination of the lifetime safety factor. Additionally, it

should be emphasized that poor durability performance of the structure may lead into serious

resistance and serviceability consequences, which must be considered rationally in the durability

design process.

Considering that any premature failure of the structure would represent serious consequences on

the surrounding environment, a reliability index = 3.8 is adopted for ULS. Accordingly, the

lifetime safety factor at ULS for this structure can be determined as:

201.218.08.3 21 t

Consequently, the design service life for the ultimate limit state results:


25

752ULSdt years 150 years

Considering that any premature serviceability failure of the structure would represent serious

consequences on the use of the structure, a reliability index = 2.5 is adopted for SLS.

Accordingly, the lifetime safety factor at ULS for this structure can be determined as:

73.118.05.2 21 t

Therefore, the design service life for the serviceability limits state results:

7573.1 SLSdt years 13075.129 years

DURABILITY PARAMETERS

26

5. DURABILITY PARAMETERS

5.1. Durability Design Aspects

The environmental conditions around the bridge induce certain physical, chemical and biological

processes responsible for the different mechanism of deterioration that can occur and affect the

different composing elements, depending on their degree of exposure to these specific

environmental conditions.

5.1.1. Identification of Macro-climatic Conditions

The bridge project is situated in the southwest of the province of Quebec (Canada), near the city

of Montreal, with an approximate longitude of 73º 35’ west of the Greenwich meridian, and a

latitude of 45º 30’ north of the Equator.

Figure 5.1: Montreal climate graph (Altitude: 57m).15

The environment of this zone is greatly influenced by the confluence of several climatic regions

generating a climate classified as humid continental or hemiboreal, according to the Köppen

climate classification14. The average temperature is 7ºC, with a highest monthly average high


27

temperature of 26ºC in July and a lowest monthly average low temperature of -13ºC in January.

The range of average monthly temperatures is 31ºC. However, on the hottest and coldest days,

temperatures can go up to 37ºC and -40ºC respectively (Figure 5.1).15

This region receives an average of 1047mm of precipitation per year, and 87mm per month.

During the driest conditions in February, it is possible to have an average precipitation of 76mm

during 15 days. However, during the wet season in summer, it is possible to receive an average

precipitation of 98mm during 13 days. There is an average annual precipitation of 218cm of

snowfall, which occurs from November through March. Thunderstorms are common; they begin in

the late spring and usually finish in the early fall season. Tropical storm remnants can bring heavy

rains during the hurricane season at the Atlantic and the Pacific oceans. The average annual

relative humidity is 77.4% and the average monthly relative humidity varies from 71% in May to

83% in September. 14, 15

Figure 5.2: Ice accretion in the Montreal metropolitan area.1

During winter season, important snow and ice accretions take place in the rural and urban areas

of all regions in Canada. The degree of accumulation varies according to the geographic location.

In the case of the metropolitan region of Montreal, there exists an extreme possibility of

accumulation of ice, generating important mechanical and environmental loads that must be


28

considered (Figure 5.2). Additionally, due to the extreme cold temperatures during this period of

the year, saturated soil can be subjected to freezing up to a certain depth below the ground level.

The frozen depth of soil depends on different factors, where the most important is the cold

temperature. This factor can be estimated by the frost index, which is calculated by adding all the

mean daily temperatures of the air below 0ºC during the year. The depth of frozen soil is a

function of the frost index (Figure 5.3). According to the Ministry of Transportation of Quebec

(MTQ), the depth of frozen soil can be more than 1m. This factor must be considered when

designing the foundations of the bridge, placing the foundation elements at adequate depth to

protect the footings, shafts, piles and pile-caps against the effects of freezing and thawing cycles.

Consequently, the minimum required protection of the foundation units of a bridge has been

established by the MTQ considering the frost index and the classification of the road way (Table

5.1).

Figure 5.3: Frost Index for southern regions of Quebec.16

Frost index (ºC * days)

Depth of the frozen soil (m)

National Highway Regional and

Collector Road Local Road

< 1200 2.00 1.80 1.60 1200 – 1700 2.25 2.00 1.80

< 1700 2.50 2.25 2.00

Table 5.1: Depth of frozen soil in terms of frost index.16


29

The average wind speed around the Montreal’s metropolitan area is relatively constant

throughout the year, showing more intensity during the winter season. According to the Canadian

Highway Bridge Design Code, an hourly mean wind pressure of 461Pa can be estimated for the

analysis of wind loads for bridge structures located in the metropolitan area of Montreal for a

return period of 150 years. 1

The bridge project is meant to cross a river, 75m in width and 6m in depth at its deepest point.

This is generally calm and unaffected by tides; however, there is some fluctuation between high

water and low water levels during the driest season during the summer and the wet seasons from

fall to spring. During winter, the river does not get frozen; however, it carries ice blocks of large

sizes that may affect the foundations of the structure.

5.1.2. Identification of Micro-climatic Conditions

The macroclimate surrounding the bridge can create certain micro-climatic or local conditions at

different locations of the structure, depending on the degree of exposure of the structural

elements to the specific environmental conditions.

The different microclimate zones that can be identified in a typical cross-section of the bridge

deck are shown in the Figure 5.4.

Figure 5.4: Microclimates on the bridge deck.

It is possible to identify a wider range of microclimates at the foundation units, such as the

intermediate pier, where its different components can experience varying conditions and can

develop different mechanisms of deterioration. The possible microclimates that can be identified

at the central pier are displayed in the Figure 5.5.

A complete description of the microclimates, exposure conditions and mechanisms of

deterioration that take place on the different structural members of the bridge, is presented in the

Table 5.2.


30

Figure 5.5: Microclimates of the bridge at the intermediate pier.

Microclimate Bridge Element Exposure Conditions Deterioration Mechanisms

A Upper side of the deck slab

Rain and wind exposure; accumulation of de-icing salts; abrasive deterioration mechanisms due to traffic; freezing and thawing cycles; exposure to CO2; wetting and drying cycles; exposure to sunshine and daily temperature changes.

Frost attack; surface deterioration; carbonation and chloride-induced corrosion.

B Reinforced concrete barrier

Rain and wind exposure; accumulation of de-icing salts; splashing of chloride solutions and water from continuous traffic; freezing and thawing cycles; exposure to CO2; wetting and drying cycles; exposure to sunshine and daily temperature changes.



31

C

Reinforced concrete barrier; exterior girders; external faces of the piercap.

Exposure to rain; wind abrasion; accumulation of de-icing salts; wetting and drying cycles; freezing and thawing cycles; exposure to CO2; rundown of water carrying aggressive agents; exposure to sunshine and daily temperature changes.


D Lower face of the deck slab

Possible cracking and leakage of water carrying aggressive agents causing efflorescence; wetting and drying cycles; freezing and thawing cycles and frost attack.

Frost attack; surface deterioration and chloride-induced corrosion.

E Interior girders Exposure to wind; wetting and drying cycles; freezing and thawing cycles; frost attack.

Frost attack; surface deterioration and reinforcing steel corrosion.

F Upper parts of the pier columns and abutments.

Exposure to rain and wind; wetting and drying cycles; freezing and thawing cycles; frost attack.


G

Internal parts of the piercap; zone between the columns.

Exposure to wind; wetting and drying cycles; freezing and thawing cycles; frost attack.


H Middle part of the pier columns (spray zone).

Exposure to rain and wind; exposure to spray from the river; exposure to cyclical splashing; wetting and drying cycles; frost attack.


I

Lower part of the pier columns at the transition zone between high tide and low tide.

Frequent wetting - drying and freezing - thawing cycles; exposure to rain and wind; ice abrasion and ice impact.

Frost attack; surface deterioration and reinforcing steel corrosion; abrasion of concrete by ice.

J Lower part of the pier columns in the submerged zone.

Water and ice abrasion; ice impact; absence of air and free oxygen.

Surface deterioration; abrasion of concrete by ice.

K

Foundation of the pier columns through the sedimentary deposit of the river bed.

Embedded zones into the sedimentary stratum; absence of air. There is no significant potential for chemical attack due to the interaction with aggressive soil deposits.

No major deterioration is expected.

L Embedded zone of the pier columns into the bedrock.

Embedded zones into the bedrock; absence of air. There is no significant potential for chemical attack due to the interaction with aggressive soil deposits.

No major deterioration is expected.

Table 5.2: Microclimates, exposure conditions and mechanisms of deterioration of the different structural

members of the bridge.


32

5.1.3. Environmentally induced mechanisms of deterioration

It is possible to identify the main mechanisms of deterioration from all possible modes of

deterioration in a component, that may affect the integrity of the construction materials and hence,

the bearing capacity of the structural members of the bridge. The mechanisms are described in

the following sections.

5.1.3.1. Frost Attack

Frost attack causes a gradual disintegration of concrete surfaces as a result of repeated freezing

and thawing cycles. In environments with moist and freezing conditions, a decrease in concrete

strength takes place, generating an eventual disintegration and complete loss of material near the

surface after a prolonged period of exposure.

The main reason for the disintegration of concrete is the effect of water freezing inside the

capillary pore system. When water freezes, it expands by about 9% of its volume, inducing some

internal pressures inside the pore system that eventually are dissipated through the generation of

cracking. The rate of this mechanism of deterioration depends not only on the quality of the

concrete (strength, low permeability, air content, etc.) but also on the aggressiveness of the

environmental conditions. Additionally, the presence of de-icing salts worsens even more the

consequences of this mechanism by adding the effects of concrete flaking or scaling caused by

the crystallization of salts, solar heating and water migration towards the surface (Litvan’s

theory).10

The gradual weakening or disintegration of the concrete cover surface, resulting from repeated

freezing and thawing processes can be modeled by the equation: 5

4.17.0 cmagecurenv facccr (5-1)

where:

r = the rate of disintegration [mm/year].

cenv = the environmental coefficient.

ccur = the curing coefficient.

cage = the ageing coefficient.

a = the air content [%].

fcm = the mean cubic compressive strength of concrete [MPa]; fcm= (f’c/0.80).

The curing and ageing coefficients can be expressed by the following equations:


33

dccur

10log17.085.0

1

(5-2)

flslsfage ppp

c001.0008.0045.01

1

(5-3)

where:

d = the curing time [days].

psf = the proportion of silica fume with respect to the total weight of binding agent [%].

psl = the proportion of blast furnace slag with respect to the total weight of binding agent [%].

pfl = the proportion of fly ash with respect to the total weight of binding agent [%].

Different environmental coefficients (Cenv) have been evaluated and presented by Sarja and

Vesikari 5, depending on the macro-climate conditions of different latitudes around world. These

coefficients are shown in Table 5.3.

Class Conditions Cenv

1

Very Hard: Frost, snow, ice, numerous freezing and thawing cycles, salt water or de-icing salts, temperature and moisture variations. Latitudes 60º ± 5º.

80 - 160

2

Hard: Frost, snow, ice, numerous freezing and thawing, constant contact with water (no chlorides), temperature and moisture variation. Latitudes 60º ± 10º.

40 - 80

3 Moderate: Normal outdoor conditions, freezing and thawing effects. Latitudes: 60º ± 10º.

20 – 40

4 Favourable: No freezing and thawing effect.

< 20

Table 5.3: Environmental coefficient values for frost attack.

5

5.1.3.2. Abrasion of concrete by ice

The reinforced concrete columns of the intermediate pier can suffer chemical changes from the

permanent exposure to the flow of the river. Additionally, some physical changes can also occur

due to wetting and drying cycles at the transition zone between the high water level and the low

water level, and due to freezing and thawing cycles. However, the final cause for the detachment

of concrete is the impact and friction of ice blocks or ice sheets against the concrete surface.

There exist many models that are intended to represent the degradation of concrete caused by

ice abrasion. However, in structural design of reinforced concrete structures the following


34

approximations can be used for obtaining the rate of abrasion expressed in the loss of concrete in

time. 5

When the aggregate stones are not loosening due to frost attack, the rate of abrasion can be

estimated as:

vf

pr

ck

'3 (5-4)

When the aggregate stones are loosening due to frost attack, the rate of abrasion can be

estimated as:

vfp

rck

'

3 (5-5)

where:

= the movement of the ice blocks or ice sheets [km/year].

p’ = the total proportional volume of cement stone in concrete including aggregates up to =

4mm. (Approximately, p’ can vary between 0.4 and 0.6).

fck = the characteristic (cubic) compressive strength of concrete [MPa] 8 cmf

5.1.3.3. Surface Deterioration

This mode of deterioration represents different types of weathering mechanisms due to the

exposure of the structure to the different microclimates. The model for surface deterioration

(Equation 5-6) includes temperature and moisture fluctuations, leaching and efflorescence effects

on the internal structure of the concrete cover, and physical deterioration due to the accumulation

of salts.

Temperature cycles can cause gradual cracking in weak zones of concrete elements like corners

and edge zones. Wetting and drying cycles with climatic moisture changes can induce slight

cracking and small changes in the permeability of the concrete cover. The generation of cracking

and the increase of the concrete cover permeability can produce an increment in the ingress of

water. The flow of water through certain zones of concrete can generate leaching of concrete

minerals, affecting the durability properties of the affected regions. Salt weathering and

efflorescence is associated with the crystallization of salts and minerals in the pores of concrete

while water vaporises. An associated mechanism is the expansion and shrinkage of these

crystals as a result of hydration and dehydration leading to cracking and disintegration of the

concrete. This mechanism can be represented by the following equation: 5


35

3.3 ckcurenv fccr (5-6)

where:

r = the rate of disintegration [mm/year].


ccur = the curing coefficient.

Some environmental coefficients (Cenv) for the surface deterioration model have been evaluated

and presented by Sarja and Vesikari 5 depending on the macro-climate conditions found at

different latitudes around the world. These coefficients are shown in Table 5.4.

Class Conditions Cenv

1

Very Hard: “Gulf conditions” Marine structures or structures within the capillary rise of saline ground water, temperature and moisture variations. Latitudes 20º ± 10º.

100000 - 500000

2

Hard: Marine structures or structures within the capillary rise of saline ground water, temperature and moisture variations. Latitudes 40º ± 10º.

10000 - 100000

3 Normal: Normal outdoor conditions, small climatic changes. Latitudes 40º ± 10º.

1000 - 10000

4 Favourable: Air continuously dry, no sunshine.

< 1000

Table 5.4: Environmental coefficient values for surface deterioration.

5

5.1.3.4. Chloride-induced Corrosion

Chlorides present in de-icing salts can penetrate into the reinforced concrete elements

developing a gradient near the concrete surface. The time required for the chloride threshold

(critical level of chlorides) to be reached and destruction of the passive layer is normally defined

as the initiation time for corrosion.

The modelling of corrosion of reinforcement can be established by defining two main limit states,

the initiation period and the propagation of active corrosion (Figure 5.6). The first period

corresponds to the progression of the chloride ions into the concrete, moving towards the

reinforcing steel and increasing in concentration until the attainment of the chloride content

threshold that marks the end of the initiation period and the start of active corrosion.

The initiation period must be designed to be as long as possible to impede, or at worst diminish

the occurrence of active corrosion.


36

Figure 5.6: Diagram of the two limit states of corrosion in reinforced concrete elements.

In the case of prestressed concrete elements, the service life must corresponds to the initiation

period because once the prestressed strands get depassivated, localized pitting corrosion occurs

which can lead to brittle failures in prestressing steel subjected to high stress levels. In this

situation, and other cases where corrosion cannot be allowed, service life can be defined as:

0ttd (5-7)

where:

td = design service life

t0 = time to initiation of corrosion

In other cases, some corrosion may be allowed, which could generate some cracking in the

concrete cover due to the rust products that result from the corrosion reactions. In these cases,

the service life includes some of the propagation period of corrosion, which produces a

progressive decrease of the cross-section area of the reinforcing steel, a loss of bond between

steel and concrete, and a reduction of the effective cross-sectional area of the reinforced

concrete elements due to cracking and spalling of the concrete cover.5

In this case, the service life can be defined according to the following formula:

10 tttd (5-8)

where:

t1 = time for propagation of corrosion.

The propagation of active corrosion ends when a maximum allowable loss of cross-sectional area,

loss of bond, or crack width is attained.


37

The initiation period of corrosion can be determined based on the Fick’s second law of diffusion,

which describes the gradient of chloride content in the concrete cover as: 5

tD

xerfCC

c

sx2

1 (5-9)

where:

xerf error function x

t dte0

22

.

cx = the chloride content at depth x.

cs = the chloride concentration at the concrete surface.

x = the depth from the surface of the structure.

Dc = the diffusion coefficient of chlorides.

t = time

The initiation period of corrosion can be derived from the following equation: 5

2

210

112

1

s

thc

CC

c

Dt (5-10)

where:

cth = the critical chloride content.

c = concrete cover thickness.

t0 = time to initiation of corrosion.

Once the initiation period is terminated, active corrosion starts to deteriorate the reinforcing steel

bars. The length of the propagation period (t1) depends at first on the design criteria, whether

active corrosion is allowed or not within the design service life. If corrosion occurs during the

service life, the end of the propagation period will be determined when a critical reduction of the

reinforcing steel cross-section area and a maximum allowable reduction of the bearing capacity of

the reinforced concrete element are reached, attaining consequently the defined limit states of the

structure.

The length of this period depends on the rate of corrosion, which is controlled by different aspects,

such as ambient temperature, relative humidity, chloride content, w/cm ratio, type of cement,

degree of exposure to severe and aggressive micro-climates, and the presence of supplementary

protective measures, among other factors. It has been discovered that the rate of corrosion slows


38

gradually with time; however, in absence of detailed information about this phenomenon,

corrosion rate can be conservatively considered to be constant for structural design for durability.

Some researchers have determined the values of corrosion rate on-site and in laboratories

around the world. According to the experimental data reported by Tuutti (1982)5, some

approximate averages of rates for carbonation- and chloride-induced corrosion are presented in

Table 5.5.

Relative Humidity(%)

Carbonated Concrete(m/year)

Chloride-contaminated Concrete (m/year)

99 2 34 95 50 122 90 12 98 85 3 78 80 1 61 75 0.1 47 70 0 36 65 0 27 60 0 19 55 0 14 50 0 9

Table 5.5: Rate of corrosion at anodic areas in carbonated and

chloride-contaminated concrete (Tuutti, 1982). 5

The values in Table 5.5, as well as the factors that affect the rate of corrosion, are specified for a

temperature of +20ºC, and some corrections can be applied to obtain the actual rates of corrosion

at different temperatures. This is expressed in the following equation: 5

0rcr T (5-11)

where: r = the rate of corrosion.

cT = temperature coefficient.

r0 = the rate of corrosion at +20ºC.

Some values of temperature coefficients determined by Tuutti (1982), and average daily

temperatures for some cities in the northern hemisphere (Europe) are shown in the Table 5.6.

Some other values of corrosion rates were found in association with the relative humidity and the

nature of the mechanism of corrosion by Andrade et al. in 1994. These values are presented in

Table 5.7.


39

City

Rate of corrosion (m/year)

Exposed to rain Sheltered from

rain Sodankylä

Temperature coefficient,

CT

0.21 11 2.5 Helsinki 0.32 16 4

Amsterdam 0.47 24 6 Madrid 0.73 37 9

Doniphan County, Kansas, USA

Condition of concrete

Uncracked 1.35 -

Mission Creek, Shawnee County,

Kansas, USA

Uncracked 3.64 -

Cracked 9.37 -

Table 5.6: Temperature coefficients, condition of concrete and evaluated rates of corrosion for some cities in Europe. 5 , 27, 28

Relative Humidity(%)

Rate of Corrosionm/year)

Only aggressive action is carbonation 90 - 98 5 - 10

< 85 ≤ 2 Chloride-contaminated environments

100 ≤ 10 80 - 95 50 - 100

< 70 ≤ 2

Table 5.7: Corrosion rate found by Andrade et al (1994). 5

A major consequence of steel corrosion is the loss of bond between concrete and the steel

reinforcing bars. This loss of bond is closely related to the mass loss of the reinforcing bars.

Amleh (2000) 17 developed bond strength equations as a function of mass loss of reinforcing bars,

and other factors. These equations are shown in the Table 5.8.

Concrete Mixture Required Development Length

(Ld design)

Point Tupper fly ash concrete

(W/CM = 0.32) MLfd

c

dfL

cb

bydesignd

34.0'597.116.04.0

25.0

(5-12)

Thunder Bay fly ash concrete

(W/CM = 0.32) MLfd

c

dfL

cb

bydesignd

31.0'205.116.04.0

25.0

(5-13)


40

Sundance fly ash concrete

(W/CM = 0.32) MLfd

c

dfL

cb

bydesignd

31.0'337.116.04.0

25.0

(5-14)

Normal Portland cement concrete

(W/C = 0.32) MLf

d

c

dfL

cb

bydesignd

31.0'284.116.04.0

25.0

(5-15)

Normal Portland cement concrete

(W/C = 0.42) MLf

d

c

dfL

cb

bydesignd

31.0'425.116.04.0

25.0

(5-16)

Table 5.8: Required development lengths for corroding steel reinforcement in various concrete mixtures. 17

where: Ld = the required development length for the reinforcing bar under corrosion [mm].

c = the concrete cover thickness [mm].

ML = the mass loss of reinforcing bars [%].

db = the diameter of the reinforcing bar [mm].

fc = the concrete compressive strength [MPa].

fy =the reinforcing steel yield strength [MPa].

Figure 5.7: Mass loss of reinforcement vs. time due to chloride-induced corrosion. 17


41

To use these equations, it is necessary to determine the increment of mass loss of the reinforcing

bars in time as a result of the corrosion process. Mirza et al (2005) presented some charts

relating the progress of mass loss of reinforcing steel with time depending on the concrete cover

thickness and other factors. A typical chart is shown in Figure 5.7. 17

The rest of the charts are presented in the Appendix 3, for different chloride concentrations at the

concrete surface (Cs), ranging from 1% to 6%.

5.1.3.5. Carbonation-induced Corrosion

Carbon dioxide present in the air penetrates concrete, causing shrinkage of the hydrated cement

paste that leads to cracking, a reduction of the wearing resistance of the surface, and a decrease

of alkalinity of the pore solution of the concrete microstructure that creates a carbonation front

that progressively penetrates into the reinforced concrete element. When this carbonation front

reaches the reinforcement, the passive protective layer of the steel dissolves, allowing corrosion

to occur. The initiation time of corrosion is defined as the time required for the complete

carbonation of the concrete cover. The model that represents the progress of the carbonation

front from the external surface of the concrete cover towards the interior of the reinforced

concrete elements is represented by the equations 5-17 and 5-18. 5

tKd c (5-17)

bcmairenvc faccK

(5-18)

where:

d = the depth of carbonation at a time t [mm].

Kc = the carbonation coefficient.

t = age [years].


cair = the air content coefficient.

a, b = parameters depending on the binding agent.

Two distinctive environmental coefficients (Cenv) are evaluated and presented by Häkkinen

(1993)5 according to some specific micro-climatic conditions. These coefficients are presented in

Table 5.9.

Environment Cenv

Structures sheltered from rain 1 Structures exposed to rain 0.5

Table 5.9: Environmental coefficient for carbonation-induced corrosion.

5


42

Häkkinen et al. have presented some values for the air content coefficient (Cair). The typical

values are presented in Table 5.10.

Air porosity Cair No entrained air 1

Entrained air 0.7

Table 5.10: Air content coefficient for carbonation-induced corrosion.5

The same authors published some values of parameters a and b depending on the type of

cementing material present in the concrete mixture. These values are shown in Table 5.11.

Binder a b Portland cement 1800 -1.7

Portland cement + fly ash 28% 360 -1.2 Portland cement + silica fume 9% 400 -1.2

Portland cement + blast furnace slag 70% 360 -1.2

Table 5.11: Parameters a and b. 5

Accordingly, the initiation period of carbonated-induced corrosion can be calculated using the

equation:

2

0

cK

dt (5-19)

5.1.4. Minimum Required Conditions of the Main Construction Materials

The quality and thickness of the concrete cover as well as the type of reinforcing steel are two

key parameters controlling the durability of reinforced concrete members. These aspects are

some of the multiple measures of protection against aggressive agents present in the different

microclimates. The different modes of deterioration need to be evaluated for each member of the

bridge, depending on the specific environmental conditions of exposure.

After having followed an iterative process of analysis for the different modes of the deterioration, it

was possible to identify the durability requirements for the construction materials with the purpose

of attaining an appropriate performance during the service life of the bridge. The input parameters

are established according to the microclimatic conditions present in the different zones and

elements of the structure that were described earlier. These parameters are summarized in the

table 5.12. According to this information, it was possible to determine the different rates of

deterioration of the concrete members of the bridge described earlier, which were produced by

the various mechanisms of deterioration developed according to the microclimatic conditions.

These rates are described in the Table 5.13.


43

Table 5.12: Durability parameters for the various bridge components according to the different modes of deterioration.


44

As expected, the elements exposed to the most severe environmental conditions would

experience the greatest rates of deterioration. These elements are the bridge deck slab, barriers

and edge girders.

Corrosion initiates relatively soon in the girders, because the have smaller concrete cover

thicknesses among all of the elements of the bridge (Table 5.14). Active corrosion occurs at the

lowest rate for the interior girders since they are sheltered from direct exposure to the chlorides

and moisture. However, the edge girders present a significantly high rate of active corrosion. For

these elements, it is essential to establish appropriate additional protective measures to impede

the direct exposure of the member to chlorides and moisture, or at least to reduce it. These

measures could be membranes, coatings or even cornices that act as shields against the

exposure to rain and rundown of water carrying de-icing salts.

Table 5.13: Rates of deterioration, initiation time for corrosion and carbonation coefficients for the various bridge components.

Other elements, such as the bridge deck slab and the barriers have higher initiation times for

corrosion between 20 and 30 years, due to thicker concrete covers. Nevertheless, the rates of

active corrosion are the highest in the bridge, caused by the direct exposure of the required

components for corrosion which are moisture, oxygen and above all, high concentration of

chloride ions. The abutments, piercap and columns are the elements that present highest

initiation times of corrosion. Among these elements, columns which are more protected against

rain and high concentration of chloride ions are the ones that present the lowest rate of active

corrosion. Caissons and the submerged part of the columns present a significantly delayed

initiation time of corrosion of 74 years. Additionally, they experience the lowest active corrosion

rate in the bridge due to the absence of free oxygen under water.


45

The responses for the different mechanisms of deterioration for each of the bridge members are

presented in Figures 5.8 to 5.11 for a service life of 150 years.

a)

b)

Figure 5.8: Response for deterioration mechanisms for: a) Bridge deck slab, b) Barriers.


46

a)

b)

Figure 5.9: Response for deterioration mechanisms for: a) Piercap, b) Abutments.


47

a)

b)

Figure 5.10: Response for deterioration mechanisms for: a) Pier columns, b) Caissons.


48

a)

b)

Figure 5.11: Response for deterioration mechanisms for: a) Edge girders, b) Internal girders.


49

It is evident in Figure 5.10 that the abrasion of concrete by ice on columns and caissons results to

be a very aggressive mechanism that may be very difficult to control only with an adequate

design of concrete for these elements. In this case, it is essential to configure an additional

protective measure by means of ice barriers upstream, protective islands around the piers, and

the use of steel shields attached to the concrete sections, among other solutions.

The initiation and development of corrosion of reinforcing steel for the different elements of the

bridge are shown in Figure 5.12.

Figure 5.12: Chloride-induced corrosion progress in the different members of the bridge over a stipulated service life of 150 tears.

The progress of the carbonation front for the various bridge members is shown in Figure 5.13. In

this figure it is possible to identify three groups of elements that experience the same rate of

carbonation progress into concrete.

The elements, which are exposed to rainy conditions, present lower progress of carbonation,

because of the higher moisture saturation in these elements, which impede or make more difficult

the diffusion of CO2 into the concrete. This is the case of the bridge deck slab, barriers and edge

girders, which are normally quite saturated with moisture.

Other elements like the piercap, abutments and caissons, which are more protected against rain,

and therefore have a lower moisture content, demonstrate a higher progress of the carbonation


50

front. Also, the elements that are sheltered from rain, such as the internal girders and columns,

also experience the most rapid progress of the carbonation front.

Figure 5.13: Progress of carbonation front in the bridge.

5.1.4.1. Minimum Concrete Cover

From Figures 5.8 to 5.13, and the results of the rates of deterioration of concrete, it is possible to

determine the minimum concrete cover thickness necessary to ensure satisfactory performance

of the bridge members during the established service life.

For instance, the necessary concrete cover thickness for the top surface of the bridge deck slab

can be established by identifying the deterioration of the concrete cover that take place due to the

most severe mechanism of deterioration, which in this case is the surface deterioration. After a

design service life of 150 years, a depth of 62mm of concrete will be affected. However,

considering a simultaneous degradation process caused by carbonation, an additional 11 mm will

be necessary to protect the reinforcing steel. Therefore, the minimum concrete cover thickness

should be 62mm + 11mm = 73mm, which is larger than the minimum specified concrete cover

thickness in the different standards. CSA-S6-06 specifies for this element a concrete cover

thickness of 70 ± 20mm1, CSA A23.1 specifies 60mm18, and the Ministry of Transportation of

Quebec (MTQ) establishes a minimum concrete cover thickness of 60mm. Following the

principles of design for durability, the adopted concrete cover thickness for the top slab of the

bridge deck slab is 75mm. This is a considerable thick concrete cover. For this reason, especial

care will be taken for the design of this element, especially in terms of cracking control. The

concrete mixture for the slab will have microfilament polypropylene fibres to reinforce the


51

concrete cover. Additionally, a galvanized thin wire mesh can be provided at the middle of the

concrete cover to improve crack control in case of thick concrete covers.

Similarly, the concrete covers for other bridge members can be determined, following the same

procedure. Table 5.14 summarizes the minimum required and adopted concrete cover

thicknesses for the different bridge elements.

Table 5.14: Concrete cover thicknesses for the analyzed members of the bridge.

For design purposes, in the cases where the minimum required concrete cover thickness is

smaller than the minimum specified by the standards, the previous one will be adopted for the

element to respect the code provisions. For instance, in the case of barriers, the minimum

required concrete cover thickness is 56mm; however, the MTQ code establishes a minimum of

75mm for these elements, therefore the concrete cover adapted is 75mm.

5.1.4.2. Type of Steel

Development and progress of corrosion in reinforcing steel has been analyzed for this example,

using regular uncoated construction steel. However, following a multiple-stage protection strategy

for durability design, the recommended type of steel to be used in the analyzed member of the

bridge is galvanized steel or epoxy-coated steel. This kind of steel will increase the corrosion

initiation time or even better, under optimal conditions, it could impede this mechanism of

deterioration.

Hot-dip galvanized rebars can be used for large bridge members like abutments, transition slabs,

bridge deck slab, diaphragms, and piers. This kind of steel allows the use of cathodic protection

as another supplementary protection measure in the future when it is necessary. For other

members, such as barriers and girders, epoxy-coated steel can be used, following adequate

procedures of assembling and coating the reinforcing bars.

Especial care must be taken when using epoxy-coated steel in bridge construction. During

handling, bending, attaching and placing the reinforcement, the epoxy coating around the bar can

get damaged, leaving some unprotected spots and pinholes, where localized corrosion could start

and the spread along the bar. A good solution has been implemented for some important bridge


52

projects around the world using a technique called “the fluidized technique for epoxy coating”.

This technique consists of preparing welded reinforcement cages, which are blast cleaned,

heated to 160ºC, dipped into unheated fluidized powder of epoxy, and then cured and stored. The

epoxy powder fuses onto the reinforcing bars providing an even coating practically without any

pinholes, and covers all deformations as well as welding points in a reliable manner.19

5.2. Concrete Mixture Design for Durability

Because the concrete in different bridge elements is going to be exposed to diverse and specific

micro-climatic conditions, it must be able to cope with these different conditions, and therefore, its

quality can vary from element to element. Consequently, special concrete mixture designs need

to be prepared for the various elements of the structure, following certain durability requirements

and taking into consideration the precise environmental exposure of each bridge member.

5.2.1. Concrete Mixture Requirements

The microclimatic conditions described in Table 5.2 can be associated with the types of exposure

indicated in the Table 1 of the CSA Standard A23.118. The types of exposure conditions in this

table that concern the various bridge elements are:

C-XL: Reinforced concrete exposed to chlorides or other aggressive agents, affected or not

exposed to freezing and thawing cycles, and expecting higher durability performance exposures

classes C-1, A-1, and S-1.

C-1: Reinforced concrete exposed to chlorides and affected or not exposed to freezing and

thawing cycles. Examples: Bridge deck slabs, slabs and ramps for parking structures, coastal

structures located at the tidal and splashing zones, concrete structures exposed to splashing and

spraying of sea water or salted water from reservoirs.

C-3: Concrete permanently submerged, exposed to chlorides but not exposed to freezing and

thawing cycles. Example: submerged parts of coastal structures.

F-1: Concrete exposed to freezing and thawing cycles in saturated conditions but not exposed to

chlorides.

Considering the exposure condition of the bridge members, supplementary cementing materials

can be used in the concrete mixture to improve workability, enhance durability, decrease paste

porosity, decrease the heat of hydration and its rate of generation. The supplementary cementing

materials that are used tor the concrete mixtures are silica fume, blast furnace slag and fly ash.

Because these materials are added to the concrete mixture, prolonged and adequate curing

processes need to be implemented.


53

For protection against frost attack and freezing-and-thawing cycles, it is essential to use entrained

air for improved performance of the concrete in these bridge elements. A good microstructure of

concrete in terms of void size distribution and connectivity (permeability) is enhanced with the use

of low water to cementitious materials ratios. At lower W/CM ratios there is lower capillary

porosity in the concrete microstructure. Adequate slumps of concrete are required for some

bridge elements that may contain congested reinforcement, especially in the bridge deck slab and

the precast and prestressed girders. This fluidity of concrete must be attained without increasing

the W/CM ratio, which is a key factor to the durability of concrete. The use of plasticizers may be

necessary and recommended in some cases.

Corrosion protection of the reinforced concrete elements can be planned at the concrete mixture

design stage by using corrosion inhibitor admixtures. These admixtures contains anodic inhibitors

such as nitrites that block the corrosion reaction of the chloride ions by chemically reinforcing and

stabilizing the passive protective film on the steel, which is created by the high pH environment in

concrete. In effect, the chloride ions are prevented from penetrating the passive film and making

contact with the steel when this kind of admixture is added to the concrete.

To minimize cracking in the bridge members, especially those that have large surfaces,

shrinkage-reducing admixtures can be used. Additionally, as it has been mentioned before,

because some elements have thick concrete covers, the use of microfibers will help to control

crack generation. The use of fibres in concrete helps to control plastic shrinkage and settlement

cracking. Additionally, these fibres can improve impact, shatter and abrasion resistance of

concrete; they enhance durability and toughness of concrete, and can also help to reduce its

bleeding. Microfilament polypropylene fibres are recommended instead of steel fibres to avoid

corrosion problems that may initiate cracking.

5.2.2. Base Materials Specifications

The base materials for the concrete mixture designs, available at the various concrete mixing

plants in Montreal, are described as follows:

Cement: Type 20, relative density 3.14.

Silica Fume: Relative density 2.25.

Blast Furnace Slag: Relative density 2.90.

Fly Ash: Relative density 2.60.

Coarse Aggregate: Well-graded crushed rock with an oven-dry relative density of 2.68, absorption

of 0.5%, and oven-dry density of 1600kg/m3. The laboratory sample has a moisture content of 2%.


54

Fine Aggregate: Natural sand with some crushed particles with an oven-dry relative density of

2.64 and an absorption of 0.7%. The laboratory sample has a moisture content of 6%.

5.2.3. Concrete Mixture Proportions

The concrete mixture proportions are summarized in detail in the Table 5.15 for each bridge member.

Table 5.15: Detailed proportions for the concrete mixture designs.


55

The summary of the concrete mixtures design is shown in Table 5.16.

Table 5.16: Summary of the concrete mixture proportions for the bridge members.


56

5.3. Concrete Handling, Placing and Curing

5.3.1. Concrete Handling

The handling and transporting of concrete must be planned carefully to avoid the occurrence of

certain situations that could seriously affect the quality of the finished concrete. These situations

are delays, early stiffening and drying out, and segregation.

Additionally, it is essential to provide the use of the proper equipment for producing, transporting,

placing and finishing concrete in the most effective and efficient way. Considering some of the

construction characteristics for some elements, the use of the concrete pump is essential. As a

result, the concrete mixture has to be designed such that it will have adequate fluidity without

increasing the W/CM ratio; this is the case for the deck slab, abutments, columns and caissons.

For other elements, such as barriers and the piercap, concrete can be easily cast using the crane

and bucket system.

5.3.2. Placing and Finishing of Concrete

The placement and casting of concrete at the construction site or at the precast facility must be

performed carefully, following appropriate standards and procedures. The adequate placement

and handling of concrete will determine the quality of concrete. All concrete casting operations

need to be performed to avoid segregation that results in problems, such as rock pockets and

honeycombs, which results in increased ingress of aggressive agents into the concrete elements.

Consolidation is an essential process for a well-finished concrete element. It is related to vibration,

which promotes the uniformity of concrete and the removal of entrapped air. Nevertheless,

prolonged vibration must be avoided to prevent segregation of concrete and removal of entrained

air; it also influences the structure of internal voids and increasing W/CM ratios on the periphery

of the concrete elements inside the formwork and in the upper surfaces (concrete covers).

5.3.3. Curing

For the concrete mixtures described before, curing must start immediately after finishing. A

minimum curing period of seven days under moist conditions must be implemented for all cast-in-

place concrete elements. To lower the permeability of concrete surfaces exposed to the most

aggressive conditions, including the concrete slab and barriers, the curing period should be

increased to 14 days under moist conditions.

All recently placed and finished concrete surfaces must be cured and protected from drying, from

extreme changes in temperature, and from damage by subsequent construction operations and

traffic.


57

Certain limitations of the proportions of pozzolanic materials added to concrete should be

respected to ensure the effectiveness of the addition of these materials. Additionally, for

concretes with high dosages of supplementary cementing materials, the slow rate of pozzolanic

reactions requires prolonged periods of moist curing to achieve the benefits of the pozzolans;

otherwise, the addition of these materials become non-cementitious fillers. Pozzolans react with

Calcium Hydroxide and water to produce more Calcium Silicate Hydrate, which adds to the

strength and enhances the durability characteristics of the concrete. The extent of the pozzolanic

reactions is determined by the consumption of Calcium Hydroxide and moisture in the concrete.

Special care must be exercised during design and proportioning of the concrete mixture limiting

the dosages of the supplementary cementing materials. Elevated contents of silica fume can

make concrete highly cohesive with low aggregate segregation or bleeding. Low levels of bleed

water at the concrete surface contribute to the appearance of plastic cracking due to rapid water

evaporation at the surface, especially on hot and windy conditions. Special precautions are

needed when concrete is placed and finished during periods with adverse weather. During cold

weather, some arrangements, such as heating, covering, enclosing and insulating of concrete

must be provided to avoid freezing of concrete or a significant decrease in temperature. By

contrast, during hot-weather conditions, special precautions against rapid evaporation and drying

need to be taken.

The curing methods that maintain the presence of mixing water in concrete during the early

hardening period are recommended, especially for bridge deck slab, approach slabs, foundation

units and certain parts of piers and abutments. These methods include ponding, fogging, spraying

and employing saturated coverings.20

Other methods that involve the use of impervious membranes and that reduce the loss of mixing

water from the surface of concrete are recommended for most of the bridge members, including

the soffit of the bridge deck slab, diaphragms, girders, abutments, piers, and barriers. Steam

curing may be used for the girders in adequate facilities, considering that these elements are

normally precast, cured, and then transported to the construction site and installed.

Steam curing can enhance the gain of concrete strength at early ages; however, this is not the

main priority in this study. What is more important is to perform a proper curing process that will

ensure an adequate concrete performance, enhancing the mechanical properties, the durability

characteristics, the reduction of volume changes due to shrinkage, and an optimum pore size

distribution and structure that minimizes capillary porosity, and hence the permeability of concrete.

STRUCTURAL DESIGN FOR DURABILITY

58

6. STRUCTURAL DESIGN FOR DURABILITY

The structural design of the different bridge members is performed according to the guidelines

and specifications provided by the Canadian Highway Bridge Design Code (CHBDC) 1.

6.1. Materials Properties

Based on the durability requirements for the construction materials described earlier (Chapter 5)

as well as the necessary bridge performance requirements during the service life of the structure,

the summary of the material properties for the different elements of the structure is presented in

Table 6.1.

Material Properties

Properties Slab Barriers Piercap Columns Abutment Caissons Girders

Reinforced Concrete Elements

Unit weight, c (kgf/m³) 2284 2203 2272 2285 2337 2300 2273

Compressive strength at 28 days, f 'c (MPa)

45 45 35 35 35 35 50

Compressive strength at transfer, f 'ci (MPa)

- - - - - - 40

Modulus of elasticity, Ec (MPa) 28729 27216 25946 26158 27057 26417 29712

Reinforcing Steel

Yield strength, fy (MPa) 400 400 400 400 400 400 400

Modulus of elasticity, Es (MPa) 200000 200000 200000 200000 200000 200000 200000

Prestressing steel, high tensile 7 wire, low relaxation (stabilized) strands

Nominal strand diameter, dps (mm)

- - - - - - 12.7

Area of prestressing strand, Aps (mm²)

- - - - - - 98.7

Ultimate strength, fpu (MPa) - - - - - - 1860

Strand braking strength (kN) - - - - - - 183.58

Modulus of elasticity, Ep (MPa) - - - - - - 200000

Factor kp of steel strands 0.30

Table 6.1: Material properties for the elements of the bridge.

The edge girders will be designed, because of their exposure to a more aggressive microclimate,

therefore, only the material properties for the exterior girders are presented in Table 6.1 along

with the properties of materials for other structural elements.


59

6.2. Superstructure Design

The superstructure of the bridge can be analyzed using a simplified method for dead and live

loads. However, certain conditions must be respected according to Clauses 5.6.1.1 and 5.7.1.1 of

the Canadian Highway Bridge Design Code1, these are summarized in Tables 6.2 and 6.3.

Article 5.6.1.1 of CHBDC

Conditions for use Bridge properties Condition satisfied

(a) Constant width Constant bridge width = 15.07m

Yes

(b) Supports conditions are closely equivalent to line supports at ends and intermediate supports

Aligned simple supports at end and intermediate supports

Yes

(c) For slab-on-girder bridges, the skew parameter must be less than 1/18

The bridge is straight and rectangular. The skew angle is 0º, consequently, the skew parameter is 0

Yes

(d) Restrictions described in Clause A5.1.3.2 for bridges that are curved in plan and built with shored construction

The bridge is straight in plan. Therefore, the condition does not apply.

Yes

(e) Uniform slab cross section Uniform slab thickness = 225mm

Yes

(f) For a bridge with longitudinal girders and an overhanging deck slab, the overhang must not be greater than 1.8m and must not exceed 60% of the spacing of the two outermost adjacent webs.

Bridge deck overhang = 1.25m 60% spacing = 0.6 x 2.095m = 1.257m

1.25m < 1.8 1.25m < 1.257m

Yes

Table 6.2: Conditions for use of the simplified method of analysis for dead loads.

Article 5.7.1.1 of CHBDC

Conditions for use Bridge properties Condition satisfied

(a) Constant width Constant bridge width = 15.07m Yes (b) Supports conditions are closely equivalent to line supports at ends and intermediate supports

Aligned simple supports at end and intermediate supports

Yes

(c) For slab-on-girder bridges, the skew parameter must be less than 1/18

The bridge is straight and rectangular. The skew angle is 0º, consequently, the skew parameter is 0

Yes

(d) Restrictions described in Clause A5.1.3.2 for bridges that are curved in plan and built with shored construction

The bridge is straight in plan. Therefore, the condition does not apply.

Yes

(e) Uniform slab cross section Uniform slab thickness = 225mm

Yes

(f) At least three longitudinal girders of equal flexural rigidity and equal spacing, or with variations from the mean in rigidity and spacing of not more than 10% in each case.

There are seven identical girders distributed with the same spacing of 2095mm

Yes

(g) Requirements for box-girder bridges

For this structure, which is a slab-on-girder bridge, these requirements are not applicable.

N/A


60

(h) For a bridge with longitudinal girders and an overhanging deck slab, the overhang must not be greater than 1.8m and does not exceed 60% of the spacing of the two outermost adjacent webs.

Bridge deck overhang = 1.25m 60% spacing = 0.6 x 2.095m = 1.257m

1.25m < 1.8 1.25m < 1.257m

Yes

(i) Requirements for continuous spans given by the Clause A5.2.1.2

The girders are designed and built as simply supported beams. A condition of semi-continuity will be introduced in order to avoid the use of expansion joint at the intermediate supports.

N/A

(j) Requirements for box-girder bridges


N/A

(k) Requirements for box-girder bridges


N/A

Table 6.3: Conditions for use of the simplified method of analysis for live loads.

The Clause C5.7.1.1 of the Commentary on the Canadian Highway Bridge design Code 21

describes that for a deck-on-girder bridge with equally spaced girders, the most desired condition

of the cantilever overhang should be nearly 50% of the girder spacing. According to this, each

longitudinal girder can be associated to equal tributary areas and thus, a uniformly distributed

load over the entire deck area would then result in girders supporting equal external actions in

terms of mechanical loads. However, a maximum overhang of 60% of the girder spacing is

permitted, where the outer girders would be required to resist higher bending moments and shear

forces than the interior ones; nevertheless the assumption for uniformly distributed loads is still

acceptable.

In the present case, the cantilever overhang of the bridge deck is 1.25m / 2.095m = 0.60.

According to this, the geometric characteristics of the bridge deck of the structure respect all of

the previously mentioned requirements, thereby permitting the use of the simplified structural

analysis of the superstructure of the bridge.

6.2.1. Prestressed Concrete Girder Design

The design of the edge girders is presented in detail here, because these elements are more

critical than the intermediate girders, in terms of structural and environmental performance.

Following the standards determined by the Ministry of Transportation of Quebec (MTQ) and most

of the transportation agencies of the East Coast of North America, the sections that are

considered in this design exercise are the New England Bulb-Tee (NEBT) precast girders (Figure

6.1). A complete structural analysis is performed to determine the required structural parameters


61

for the girders in to withstand the external loading conditions. After several iterations, the girder

section, NEBT1600, was the one that revealed adequate bearing capacity immediately after

construction, and sufficient capacity in reserve to endure the external actions throughout the

entire service life. During this iterative process, the material properties and the general

configuration of the bridge were modified and improved to adjust the durability and mechanical

characteristics of the girders.

Figure 6.1: NEBT girder characteristics. 22, 16


62

6.2.1.1. Bridge Deck Parameters

The bridge deck parameters for the girder design are listed as follows:

Girder span (m) = 26 Center-to-center girder spacing (m) = 2.095

Width of the design lane (m) = 3.05 Shoulder width (m) = 1

Number of shoulders = 2 Number of lanes = 4

Girder NEBT1600 at year = 0 Number of girders = 7

Height (mm) = 1600 Width of the upper flange (mm) = 1200

Area, Ag (mm² x 10³) = 589 Centroid to bottom, Yb (mm) = 760.9

Moment of inertia, Ig (mm^4 x 10^9) = 204.8 Section modulus from the top, Stg (mm³ x 10^6) = 244.07

Section modulus from the bottom, Sbg (mm³ x 10^6) = 269.16 Weigth (kN/m) = 13.13

Web thickness, bv (m) = 0.18 6.2.1.2. Composite Section

According to Clause 5.8.2.1 of the CHBDC 1, the parameters of the composite section are shown

in Figure 6.2.

Figure 6.2: Geometrical parameters of the composite section.


63

6.2.1.3. Diaphragms

The number of diaphragms and their distribution are established according to the Clause 8.18.5.

Accordingly there is a total of five diaphragm distributed along each span. These diaphragms are

1.10m in height and 0.30m in width.

6.2.1.4. Wearing Surface

The wearing surface is meant to be composed by a waterproofing membrane welded on top of

the deck slab and an asphalt concrete layer of 65mm in thickness. The unit weight of the asphalt

concrete is considered as 23.5kN/m3. This pavement is placed over a width of 14.2m along the

three spans of the bridge.

6.2.1.5. Barriers

There are cast-in-place reinforced concrete barriers at each edge of the bridge deck. According to

the MTQ Standards 22, these barriers correspond to a barrier type 301. These are selected and

designed to have a level of performance PL-3 in accordance to the CHBDC 1. The cross-section

area of each of these barriers is 0.35m2.

6.2.1.6. Load Analysis

The load analysis is performed per metre length of the bridge deck. The results from this analysis

are presented in the following sub-sections.

6.2.1.6.1. Dead Loads

DLslab (kN/m) = 11.90 DLgusset (kN/m) = 1.34 DLgirder (kN/m) = 13.13

DLdiaphragms (kN/m) = 2.98 6.2.1.6.2. Superimposed Loads

DLwearing surface (kN/m) = 3.10 DLbarriers (kN/m) = 2.16

Total (kN/m) = 5.26 6.2.1.6.3. Live Loads

The design truck CL-625 described in the CHBDC is implemented for the design of the

superstructure (Figure 6.3). The design truck is placed centrally in a space of 3m within each

lane. Alternatively a CL-W lane load will be analyzed in order to determine the possible most

critical condition.


64

a)

b)

c)

Figure 6.3: a) CL-625 design truck clearance envelope. 1

b) CL-625 and CL-W design truck loads. 1

c) CL-625 and CL-W design lane loads.1

6.2.1.6.4. Acting ending Moments

The bending moments due to dead and live loads along a typical simply supported span of the

bridge are presented in Table 6.4.


65

Point Distance/L

Distance (m)

Bending moments (kNm) Mg Ms Md Msd Mt

0 0 0.00 0.00 0.00 0.00 0.00 0.1 2.6 399.46 362.01 90.58 186.64 881.50 0.2 5.2 710.16 643.58 161.03 331.80 1638.20 0.3 7.8 932.08 844.70 211.35 435.48 2165.70 0.4 10.4 1065.24 965.37 241.54 497.70 2267.20 0.5 13 1109.62 1005.59 251.60 518.43 2368.70 0.6 15.6 1065.24 965.37 241.54 497.70 2208.40 0.7 18.2 932.08 844.70 211.35 435.48 1854.90 0.8 20.8 710.16 643.58 161.03 331.80 1487.00 0.9 23.4 399.46 362.01 90.58 186.64 743.50 1 26 0.00 0.00 0.00 0.00 0.00

Table 6.4: Bending moments on the girder produced by the different load cases.

where:

Mg = Bending moments caused by the self weight of the girder.

Ms = Bending moments caused by the self weight of the deck slab.

Md = Bending moments caused by the self weight of the diaphragms.

Msd = Bending moments caused by the superimposed dead loads.

Mt = Bending moments caused by the deign truck.

6.2.1.6.5. Analysis of the Load Effects

The conditions of the Clause 5.7.1.1 of the CHBDC are satisfied and the simplified live load

analysis can be applied. According to the Clauses 3.8.4.5.1 and 3.8.4.5.3, the dynamic load

allowance (DLA) must be equal to 0.25. Accordingly, the truck load incremented by the DLA

produces the following maximum bending moment:

MTruckDLA (kNm) = 3014.75 The lane load without DLA will cause the following maximum bending moment:

MLane (kNm) = 2684.45 According to the Clause 3.8.9, the maximum bending moment caused by the truck load on the

shoulders is:

PShoulder (kN/m²) = 4.00 MShoulder (kNm) = 676.00


66

According to Clause 3.5.1, the load factors for the Ultimate Limit State (ULS) and the

Serviceability Limit State (SLS) are 1.7 and 0.9 respectively.

A. Design Bending Moment for ULS

According to Clauses 3.8.4.4 and 3.8.4.9, the average positive design bending moment is:

Mgavg (ULS) (kNm) = 2181.37 The ULS design bending moment Mg is obtained multiplying Mgavg by an amplification factor (Fm)

which accounts for transverse variation in the longitudinal moment intensity. The modification

factor is determined as follows:

Lane width modification factor, = -0.42

Width dimension for the loading distribution, F Internal part (m) = 10.35 External psrt (m) = 9.81 ULS, SLS

Percentage correction factor, Cf (%) = 9.04 Amplification factor Fm Internal part = 1.26

External part = 1.33

Design bending moment, Mg (ULS) Internal girders (kNm) = 3210.56

External girders (kNm) = 3389.34 B. Design Bending Moment for SLS

According to Clause 3.8.4.2, the average positive design bending moment is:

Mgavg (SLS) (kNm) = 1395.40

Design bending moment, Mg (ULS) Internal girders (kNm) = 2053.76

External girders (kNm) = 2168.13 C. Design Bending Moment for FLS

According to Clause 5.7.1.2.2.2, the bending moment for the fatigue limit state (FLS) is calculated

for one design truck on the span, and affected by the DLA. The average positive bending moment

results:

Mgavg (FLS) (kNm) = 430.68 The FLS design bending moment Mg is obtained by multiplying Mgavg by an amplification factor

(Fm) which accounts for transverse variation in the longitudinal moment intensity. It is necessary

to define the vehicle edge distance according to Figure 6.4.


67

Figure 6.4: Definition of the vehicle edge distance, Dve. 1

The modification factor is determined as follows:

Lane width modification factor, = -0.42 Width dimension for the loading distribution, F Internal part (m) = 5.09

External psrt (m) = 3.91 Percentage correction factor, Cf (%) = 0

Vehicle edge distance, Dve (m) = 2.06 FLSCorrection factor for vehicle edge distance, Ce Internal (%) = 0

External (%) = 34.89 Amplification factor Fm Internal part = 2.88

External part = 2.78

Design bending moment, Mg (FLS) Internal girders (kNm) = 1240.64

External girders (kNm) = 1196.20 6.2.1.7. Prestressing Steel

The prestressing strands used for the reinforcement of the girder are low-relaxation, grade 1860

(Table 6.1). The distribution and stress losses of these strands are:

Number of straight strands = 42 Number of inclined strands = 14

Total number of strands = 56

The prestress loss at transfer, fst (MPa) = 130

The final loss, fs (MPa) = 440 According to Clause 8.7.1, the stress in the prestressing strands at transfer is:

fst (MPa) = 1246.40 The prestressing force per strand at transfer is (kN) = 123.02

The total prestressing force at transfer (kN) = 6889.10


68

The prestressing stress at SLS, fse (MPa) = 936.40 The prestressing force per strand at SLS is (kN) = 92.42

The total prestressing stress at SLS (kN) = 5175.67 6.2.1.7.1. Prestressing Strands Profile

The prestressing strands profile is determined according to the following considerations:

The prestressing strands are inclined symmetrically at 0.4L (m) = 10.4 Concrete cover of the girders (m) = 0.045

Rebar diameter for girder stirrups (m) = 0.016 Concrete cover for the strands (m) = 0.061

Centre of gravity of the strands, Ycg: at midspan (m) = 0.153 at end (m) = 0.443

Strand eccentricity (e) from the girder centroid: at midspan (m) = 0.608 at end (m) = 0.318

The eccentricities and moments due to the initial and effective prestressing forces are presented

in Table 6.5.

Fraction of L

Distance (m)

e (m)

Initial prestressing

force, Pi (kN)

Bending moment due to Pi,

Mpsi (kNm)

Effective prestressing

force, Pf

(kN)

Bending moment

due to Pf, Mpsf (kNm)

0 0 0.318 6889.10 2189.60 5175.67 1645.01 0.1 2.60 0.390 6889.10 2689.06 5175.67 2020.25 0.2 5.20 0.463 6889.10 3188.52 5175.67 2395.48 0.3 7.80 0.535 6889.10 3687.98 5175.67 2770.72 0.4 10.40 0.608 6889.10 4187.44 5175.67 3145.96 0.5 13.00 0.608 6889.10 4187.44 5175.67 3145.96 0.6 15.60 0.608 6889.10 4187.44 5175.67 3145.96 0.7 18.20 0.535 6889.10 3687.98 5175.67 2770.72 0.8 20.80 0.463 6889.10 3188.52 5175.67 2395.48 0.9 23.40 0.390 6889.10 2689.06 5175.67 2020.25 1 26.00 0.318 6889.10 2189.60 5175.67 1645.01

Table 6.5: Induced internal actions caused by prestressing.

The angle between the centre of gravity of the strands at the end and at midspan of the girder

results to be 0.63º. The horizontal factor for the horizontal component of the prestressing is

0.9999. Accordingly, the horizontal component of the prestressing force is considered to be

constant.

6.2.1.7.2. Stress Limitations

According to Clauses 8.8.4.6 and 8.4.1.8, the maximum permissible concrete stresses in

compression and tension are:


69

At transfer: Compressive strength (MPa) = 24 Cracking strength, fcri (MPa) = 2.53

At SLS: Compressive strength (MPa) = 20

Cracking strength, fcr (MPa) = 2.83 6.2.1.7.3. Stress Conditions of the Girder

A. Initial Compressive Stresses

The concrete compressive stresses in the girder at transfer (initial prestress) at the top and

bottom fibres are:

Fraction of L

Distance (m)

ftop (MPa) fbottom (MPa)

Condition

0 0 2.73 19.83 OK 0.1 2.60 2.32 20.20 OK 0.2 5.20 1.54 20.90 OK 0.3 7.80 0.40 21.94 OK 0.4 10.40 -1.10 23.30 OK 0.5 13.00 -0.91 23.13 OK 0.6 15.60 -1.10 23.30 OK 0.7 18.20 0.40 21.94 OK 0.8 20.80 1.54 20.90 OK 0.9 23.40 2.32 20.20 OK 1 26.00 2.73 19.83 OK

Table 6.6: Compressive stresses of concrete in the girder at transfer.

B. Compressive Stresses at SLS

The concrete compression stresses in the girder at SLS (final prestress) at the top and bottom

fibres are:

Fraction of L

Distance (m)

ftop (MPa) fbottom (MPa)

Condition

0 0 2.05 14.90 OK 0.1 2.60 3.63 13.46 OK 0.2 5.20 4.52 12.66 OK 0.3 7.80 4.71 12.48 OK 0.4 10.40 4.22 12.93 OK 0.5 13.00 4.56 12.62 OK 0.6 15.60 4.22 12.93 OK 0.7 18.20 4.71 12.48 OK 0.8 20.80 4.52 12.66 OK 0.9 23.40 3.63 13.46 OK 1 26.00 2.05 14.90 OK

Table 6.7: Compressive stresses of concrete in the girder at SLS.


70

C. Stress Conditions of the Composite Section

Once the concrete of the deck slab has hardened, a composite section is formed. The stresses of

concrete in the composite section are shown in Table 6.8.

Section modulus of the composite section at top fibres, Stc (m³) = 0.629 Section modulus of the composite section at bottom fibres, Sbc (m³) = 0.374

The section modulus of the transformed composite section at the interface between the concrete

deck and the top fibre of the precast girder results:

Sic (m³)= 1.037

Section Distance

(m) Concrete Deck Concrete Girder

Condition ftc (MPa) fbottom (MPa) ftop (MPa) fbottom (MPa)

0 0 0.00 0.00 2.05 13.19 OK 0.1 2.60 1.60 0.97 8.25 7.01 OK 0.2 5.20 2.96 1.79 13.09 2.05 OK 0.3 7.80 3.90 2.37 16.04 -1.14 OK 0.4 10.40 4.14 2.51 16.11 -1.50 OK 0.5 13.00 4.33 2.62 16.98 -2.32 OK 0.6 15.60 4.05 2.46 15.81 -1.19 OK 0.7 18.20 3.43 2.08 14.46 0.52 OK 0.8 20.80 2.72 1.65 12.32 2.86 OK 0.9 23.40 1.39 0.84 7.55 7.75 OK 1 26.00 0.00 0.00 2.05 13.19 OK

Table 6.8: Stress levels inside composite section.

D. Prestress losses

It is necessary to evaluate different prestress losses that take place on the girder, like anchorage

slip and friction, elastic shortening of concrete, relaxation of tendons, creep of concrete,

shrinkage and other special conditions.

i. Elastic Shortening

Modulus of elasticity of the steel strands, Ep (MPa) = 200000 Modulus of elasticity of the concrete at transfer, Eci (MPa) = 27291.32

Concrete stress at the tendon centre of gravity at Mmax, fcir (MPa) = 18.12 Prestress loss due to elastic shortening, ES (MPa) = 132.82

ii. Relaxation of Tendons at Initial Prestressing

Age of concrete after casting, t (days) = 0.75 Stress in the prestressing strands at jacking, fsj (MPa) = 1450.8

Yield strength of prestressing steel, fpy (MPa) = 1674 Relaxation of tendons at initial prestressing, REL1 (MPa) = 12.82


71

iii. Prestress Loss at Transfer (Short Term)

Prestress loss at transfer (elastic shortening-short term), fs1 (MPa) = 145.64 iv. Losses at Transfer

a) Prestress Loss due to Creep

Mean annual relative humidity, RH (%) = 77.4 Factor of prestress loss due to creep of pre-tensioned concrete, Kcr = 2

Modulus of elasticity of prestressing steel, Ep (MPa) = 200000 Modulus of elasticity of the girder concrete, Ec (MPa) = 29712.26

Concrete stress at the tendons centre of gravity at transfer, fcir (MPa) = 18.12 Stress at the girder top fibre (loads-dead) at transfer, ftg (MPa) = 1.64

Stress at the girder bottom fibre (loads-dead) at transfer, fbg (MPa) = 4.55 Neutral axis at transfer, y (m) = 1.18

Concrete stress at the tendons centre of gravity (loads-dead) at transfer, fcds (MPa) = 3.96 Prestress loss due to creep, CR (MPa) = 246.82

b) Prestress Loss due to Shrinkage

Prestress loss of pretensioned members due to shrinkage, SH (MPa) = 35.73 c) Relaxation of Tendons at Transfer

Relaxation of tendons at transfer, REL2 (MPa) = 16.27 d) Prestress Loss After Transfer

Prestress loss after transfer (long term), fs2 (MPa) = 301.76

fs = fs1 + fs2 (MPa) = 447.40 Diference with respect to the initial assumed prestress loss (%) = 1.65

The prestress loss is about the same as the earlier assumed value at the beginning of Section

6.2.1.7.

e) Differential Shrinkage Between Cast-in-place Deck and Girder

Differential shrinkage strain, sh = 1.00E-04 Modulus of elasticity of the concrete deck slab, Esc (MPa) = 28729.32

Cross-sectional area of the concrete deck slab (m²) = 0.47 Force caused by the differential shrinkage (SLS), F (kN) = 338.56

Additional stress at the top of the girder (composite section) (MPa) = 1.42 Additional stress at the bottom of the girder (composite section) (MPa) = -0.19

Total stress of the girder at top fibres (composite section) (MPa) = 17.55 Stress limitation at top fibres = OK

Total stress of the girder at bottom fibres (composite section) (MPa) = -2.40 Stress limitation at bottom fibres = OK


72

6.2.1.8. Ultimate Limit States

Concrete resistance factor, c = 0.75

Prestressing steel strands resistance factor, p = 0.95 Resultant compressive and tensile force in the girder (kN) = 9766.56

Depth of the compressive block, a (m) = 0.146

The neutral axis is located within the concrete slab

Distance from the extreme compressive fibre to the C.G. of strands, dp (m) = 1.72

Depth ratio of rectangular compressive block to depth to the Cu, 1 = 0.845 Neutral axis location at ULS, Cu (m) = 0.173

Cu/dp = 0.101

The reinforced section is under-reinforced

The condition of the girder is satisfactory

The stress in the prestressing tendons at ULS, fps (MPa) = 1804 Factored flexural resistance of the section, Mr (kNm) = 15617.61

Maximum bending moment at ULS, Mu (kNm) = 7289.25 Mu < Mr OK

Cracking moment, Mcr (kNm) = 3407.54 Mr > 1.2Mcr OK

6.2.1.8.2. Anchorage Zone Reinforcement

Reinforcing bars resistance factor, s = 0.9 Area of stirrups for achorage zone, (mm²) = 23.15

Anchorage zone length, (m) = 0.4 Bar type = 10M

Selection of rebar size = OK Number of stirrups = 4 Stirrup spacing (m) = 0.13

6.2.1.8.3. Longitudinal Reinforcement at the Exterior Support

Yield strength of prestressing tendons, fpy (MPa) = 1674.00 Prestressing stress at SLS, fpe (MPa) = 1004.40

Compression factor, kp = 0.28 Distance from the extreme compression fibre to the neutral axis, c (m) = 0.70

Effective shear depth at ends, dv (m) = 1.35 Distance from the extreme compression fibre to the centroid of tendons, dp (m) = 1.47

c/dp = 0.48 Stress in prestressing tendons at factored resistance, fpr (MPa) = 1612.29

Development length of pretensioned strands, ld (m) = 1.73 Prestressing force at SLS, Pf (kN) = 5175.67

Angle between the C.G. of strands at end and at midspan (°) = 0.63


73

Horizontal component of Pf (kN) = 5175.36 Vertical component of Pf, Vp (kN) = 56.52

Bearing distance (m) = 0.5 Available resistance at the face of the bearing for all strands, Pfa (kN) = 1065.85

Design bending moment at distance dv from the bearing, Mf (kNm) = 1216.56

Fraction of L

Distance (m)

Shearing forces (kN) Vg Vs Vd Vsd Vt

0 0 170.71 154.71 38.71 68.36 339.04 0.1 2.6 136.57 123.77 30.97 54.69 339.04 0.2 5.2 102.43 92.82 23.22 41.02 289.04 0.3 7.8 68.28 61.88 15.48 27.35 39.04 0.4 10.4 34.14 30.94 7.74 13.67 39.04 0.5 13 0.00 0.00 0.00 0.00 39.04 0.6 15.6 -34.14 -30.94 -7.74 -13.67 -135.96 0.7 18.2 -68.28 -61.88 -15.48 -27.35 -135.96 0.8 20.8 -102.43 -92.82 -23.22 -41.02 -285.96 0.9 23.4 -136.57 -123.77 -30.97 -54.69 -285.96 1 26 -170.71 -154.71 -38.71 -68.36 -285.96

Table 6.9: Shearing forces cause by the different load cases.

Design shearing force at distance dv from the bearing, Vf (kN) = 1033.80

Longitudinal strain, x = 0.0000

Shear resistance factor of cracked concrete, = 0.221

Angle of the principal diagonal compressive stress to the horizontal axis, (º) = 41 Shear resistance of concrete, Vc (kN) = 285.16

Shear resistance provided by the sturrups, Vs (kN) = 694.95 The force at the face of the bearing, Flt (kN) = 1621.09

The resulting axial force (kN) = 555.23 Required area of longitudinal reinforcement (mm²) = 1388.08

Bar size for longitudinal reinforcement = 15M Number of bars required= 6.94

Adjusted number of logitudinal bars = 8 6.2.1.8.4. Shear Resistance

The shear reinforcement calculations for the girder are summarized in Table 6.10.

6.2.1.8.5. Interface Shear

A cohesion resistance for the interface shear is assumed as c = 0.5MPa, considering that fresh

concrete of the deck slab is placed against hardened concrete of the girders, with the surface

clean and free of laitance, and intentionally roughened to a full amplitude of about 5mm and a

spacing of about 15mm.

Parameter dependent on the density of concrete, 1 = 1 Type of concrete = Normal-density concrete


74

Friction coefficient, = 1 Unfactored permanent load normal to the interface area, N (kN) = 0.00

Area of shear-friction reinforcement, Avf (m²) = 0.028 Area of concrete resisting shear transfer, Acv (m²) = 31.2

The ratio Avf / Acv, = 0.0008974

Compressive stress across a shear-friction plane, (Mpa) = 0.36 Shear resistance at the plane, v (MPa) = 0.64

Average factored horizontal shear stress from end to midspan, vfv (MPa) = 0.31 The provision of vertical stirrups that go into the slab is satisfactory

Section of

L Distance

(m) Ycg (m) d (m) dv (m) Mf (kNm) Vf (kN)

0 0.0 0.443 1.56 1.35 0.00 1078.28 dvg 1.2 0.411 1.53 1.35 1216.56 1033.80 0.1 2.6 0.371 1.48 1.34 2673.04 977.90 0.2 5.2 0.298 1.41 1.27 4872.92 792.52 0.3 7.8 0.226 1.34 1.21 6422.16 267.13 0.4 10.4 0.153 1.27 1.14 6986.21 166.75 0.5 13.0 0.153 1.27 1.14 7289.25 66.36 0.6 15.6 0.153 1.27 1.14 6886.25 -331.52 0.7 18.2 0.226 1.34 1.21 5893.80 -431.90 0.8 20.8 0.298 1.41 1.27 4615.88 -787.29 0.9 23.4 0.371 1.48 1.34 2438.44 -887.67

L-dvg 24.8 0.411 1.53 1.35 1112.62 -943.58 1 26.0 0.443 1.56 1.35 0.00 -988.05

Vp (kN) Vc (kN) Vs (kN) Rebar size s (m) smax (m) sadopted (m) 56.52 285.16 739.42 15M 0.30 0.3 0.3 56.52 285.16 694.95 15M 0.32 0.3 0.3 56.52 282.24 641.96 15M 0.34 0.3 0.3 56.52 268.46 470.36 15M 0.44 0.6 0.4 56.52 254.68 0.00 15M - 0.6 0.6 56.52 240.90 0.00 15M - 0.6 0.6 0.00 240.90 0.00 15M - 0.6 0.6

56.52 240.90 36.92 15M 5.08 0.6 0.6 56.52 254.68 123.53 15M 1.61 0.6 0.6 56.52 268.46 465.13 15M 0.45 0.6 0.3 56.52 282.24 551.73 15M 0.40 0.6 0.4 56.52 285.16 604.72 15M 0.37 0.6 0.3 56.52 285.16 649.20 15M 0.34 0.3 0.3

Table 6.10: Shear reinforcement for the girder.

6.2.1.9. Serviceability Limit States

6.2.1.9.1. Deflections

Prestressing force at transfer, Pi (kN) = 6889.10 Prestressing force at SLS, Pf (kN) = 5175.67


75

Modulus of elasticity at transfer, Eci (MPa) = 27291.32 Second moment of area of the girder about its neutral axis, Ig (m^4) = 0.2048

Moment due to prestressing force at transfer, Mo (kNm)

at ends = 2189.60 at 0.4L = 4187.44

The deflections of the girder are calculated using the conjugate-beam deflection method. The

deflections of the girder at the different construction stages are determined as follows:

Vertical deflection due to the initial prestress, p (m) = 0.057 ↑

Downward deflection due to selfweight of the girder, g (m) = -0.014 ↓

Net camber at transfer, net (m) = 0.043 ↑

Deflection due to fresh concrete slab, s (m) = -0.013 ↓

Deflection due to diaphragms load, diaph (m) = -0.002 ↓

Deflection due to superimposed dead load, sdl (m) = -0.002 ↓ The deflections of the girder during the construction stages and during service are summarized in

Table 6.11.

Time Stage Deflection

factor Deflection,

(m) Direction

18 hours Transfer 1 0.043 ↑

2 months Deck

casting 1.7 0.058 ↑

3 months Hardened

deck 1.7 0.047 ↑

1 year Deck

surfacing 1.8 0.047 ↑

5 years Service 2 0.050 ↑

Table 6.11: Final deflection caused by permanent loads.

6.2.1.9.2. Superstructure Vibrations

According to the Clause 3.4.4, the verification of the structure vibrations for the Serviceability

Limit State is made. Accordingly, the acting loads on the deck are amplified by the load factors for

SLS, the DLA, and the amplification factors. The results are presented as follows:

Distribution factor, Fu = 0.439

Deflection, (m) = 0.006 Weight of the bridge per unit width, Wd (kN/m per metre width) = 16.52

Fundamental flexural frequency, FFF (Hz) = 6.47 Static deflection (m) = 0.012

The dynamic response of the bridge is satisfactory


76

6.2.1.9.3. Fatigue

In non-prestressed and fully prestressed members, the change in the stresses of the rebars and

strands due to repetitive live loads is usually not critical. But in partially prestressed members,

repetitive loads can cause fatigue damage to these parts of the structure. The allowable stress

range in straight rebars and straight strands is 125MPa (Clause 8.5.3.1 and 8.5.3.2). The

calculations for the variation in tensile stresses in the strands are presented as follows:

Variation in the tensile stress in the prestressing steel, fs (MPa) = 157.03 The stress range in straight steel bars and strands is (MPa) = 125

The stress range for steel reinforcement is exceeded Additional non-prestressing steel is required

Number of bars added at between the two lower layers of strands = 9 Size of rebar = 30M

Resultant compressive and tensile force in the girder (kN) = 2520 Depth of the compressive block, a (m) = 0.038

Effective depth, d (m) = 1.78 Incremented flexural resiatance at ULS, Mr (kNm) = 19617.86

New variation in the tensile stress in the strands, fs (MPa) = 114.08 The response of the girder for FLS is satisfactory

The final step concludes the design of the precast and prestressed concrete girder. It also

represents the performance of the girder immediately after construction. A general overview of

the girder’s reinforcement is presented in the Figures 6.5 and 6.6.

6.2.1.10. Girder Design for the Required Service Life

Considering the durability parameters described earlier in this chapter, it is possible to identify

that the controlling mechanism of deterioration for the composite section of the girders along the

service life is the surface deterioration (Section 5.1.3.3). This mechanism acts with a rate of

degradation of 0.41mm per year for the upper face of the deck slab, and 0.17mm per year for the

perimeter of the girder. Based on several in-situ inspections of actual bridges under active

deterioration, the degradation of the surface of concrete on the girders is focused around the

lower flange, which is the zone where most of the aggressive agents are concentrated (Figure

6.7). The carbonation front acts simultaneously with the surface deterioration, with a rate that

varies proportionally to the square root of time, affected by a carbonation coefficient which is 0.92

and 0.90 for the bridge deck slab and the exterior girders, respectively (Table 5.13). The

integration of these rates of deterioration cannot be made directly in the formulation of the

different parameters involved in the structural design of a prestressed concrete bridge girder.

However, the input information for these parameters can be affected by the rates of deterioration

at different time intervals, establishing an iterative process that simulates the progressive

deterioration of the bridge girder and its corresponding loss of bearing capacity and serviceability.


77

Figure 6.5: Non-prestressed reinforcement of the bridge girder.


78

Figure 6.6: Prestressed reinforcement of the bridge girder.


79

Figure 6.7: Localized deterioration on the lower flange of a bridge girder.

6.2.1.11. Girder Performance with Time

The surface deterioration and the progress of the carbonation front were calculated for time

intervals of five years, starting from the time t = 0 years when the bridge is put to service, until t =

150 years, which represents the end of the design service life. At every time interval, the

geometrical properties of the composite slab-girder section were evaluated, along with the effects

on the reinforcement. Once these basic parameters were established, a structural analysis and

design iteration was performed, including verifications of all entities calculated at time interval, t =

0 years presented previously. In Figure 6.8, the progress of deterioration of the composite slab-

girder section is represented as a grey-coloured zone that moves into the girder and the slab with

time according to the rate of deterioration acting on the bridge deck.

Figure 6.8: Representation of the degradation of the composite slab-girder section.

It is important to note that after 150 years, the progress of deterioration, including the carbonation

front, did not reach the level of the steel reinforcement, or the prestressing steel. This

corresponds to the appropriate design of the quality and thickness of the concrete covers for the


80

girders and the deck slab. However, some supplementary protection measures must be used for

these elements to diminish the deterioration in the element. These protective measures are

described in the Section 6.2.1.1.2.

From each one of the structural analysis and design iterations, the basic results with respect of

the limit states were identified and plotted, to prepare a graphic representation of the loss of

bearing capacity and serviceability of the bridge girders during the design service life. The

evaluated limit states for the bridge deck included the ultimate limit state (ULS), the serviceability

limit state (SLS) and the fatigue limit state (FLS).

Time (years)

Height (mm)

Cross-sectional

area (mm² x 10³)

Centroid to

bottom, Yb

(mm)

Second moment of

inertia (mm4 x 109)

Section modulus for the top fibre (mm³ x 106)

Section Modulus for the

bottom fibre (mm³ x 106)

Weight (kN/m)

Web thickness

bv (m)

Width of the upper

flange (mm)

0 1600 589.0 760.9 204.8 244.071 269.155 13.13 0.18 1200 5 1599 587.4 762.0 204.0 243.728 267.717 13.10 0.18 1200

10 1598 585.3 763.1 203.2 243.382 266.282 13.05 0.18 1200 15 1597 583.4 764.2 202.4 243.036 264.852 13.01 0.18 1200 20 1597 583.4 764.2 202.4 243.036 264.852 13.01 0.18 1200 25 1596 581.4 765.4 201.6 242.716 263.392 12.96 0.18 1200 30 1595 579.4 766.5 200.8 242.366 261.970 12.92 0.18 1200 35 1594 577.5 767.6 199.9 241.893 260.422 12.88 0.18 1200 40 1593 575.5 768.7 199.1 241.538 259.009 12.83 0.18 1200 45 1592 573.6 769.9 198.3 241.212 257.566 12.79 0.18 1200 50 1591 571.7 771.0 197.5 240.854 256.161 12.75 0.18 1200 55 1590 569.8 772.2 196.7 240.523 254.727 12.70 0.18 1200 60 1590 569.8 772.2 196.7 240.523 254.727 12.70 0.18 1200 65 1589 567.9 773.3 195.8 240.039 253.201 12.66 0.18 1200 70 1588 566.0 774.5 195.0 239.705 251.775 12.62 0.18 1200 75 1587 564.2 775.6 194.2 239.339 250.387 12.58 0.18 1200 80 1586 562.3 776.8 193.4 239.001 248.970 12.54 0.18 1200 85 1585 560.4 778.0 192.5 238.538 247.429 12.49 0.18 1200 90 1584 558.6 779.2 191.7 238.196 246.022 12.45 0.18 1200 95 1583 556.8 780.3 190.9 237.822 244.649 12.41 0.18 1200

100 1583 556.8 780.3 190.9 237.822 244.649 12.41 0.18 1200 105 1582 554.9 781.5 190.0 237.352 243.122 12.37 0.18 1200 110 1581 553.1 782.7 189.2 237.004 241.727 12.33 0.18 1200 115 1580 551.3 783.9 188.4 236.654 240.337 12.29 0.18 1200 120 1579 549.5 785.1 187.5 236.176 238.823 12.25 0.18 1200 125 1578 547.7 786.3 186.7 235.822 237.441 12.21 0.18 1200 130 1577 545.9 787.5 185.8 235.339 235.937 12.17 0.18 1200 135 1577 545.9 787.5 185.8 235.339 235.937 12.17 0.18 1200 140 1576 544.2 788.7 185.0 234.980 234.563 12.13 0.18 1200 145 1575 542.4 789.9 184.2 234.620 233.194 12.09 0.18 1200 150 1574 540.6 791.2 183.3 234.159 231.673 12.05 0.18 1200

Table 6.12: Geometrical properties of the NEBT1600 girder of different steps of deterioration.


81

The internal stress levels in the girder at the ULS for different steps of degradation of the bridge

deck are shown in Figures 6.9 and 6.10.

Figure 6.9: Variation of internal compressive stresses at the top of the composite section vs. time.

Figure 6.10: Variation of internal tensile stresses at the bottom of the composite section vs. time.


82

With time, compressive and tensile stresses at the girder extremes increase as a result of the

loss of cross-section area of the composite slab-girder section. The tensile stresses increased

more than the compressive stresses during the service life. However, none of these internal

stresses attained the maximum allowable levels.

The gradual loss of flexural and shearing resistances (ULS) is shown in the Figures 6.11 and 6.12.

Figure 6.11: Loss of flexural resistance of the composite section vs. time.

Figure 6.12: Loss of shearing resistance of the composite section vs. time.


83

The losses of flexural and shearing resistances vs. time are described by parabolic curves. The

loss of shearing resistance was more pronounced than the loss of flexural capacity. However,

none of these ultimate limit states were attained by the composite section.

The increment of the static deflections with time that are accounted for the analysis of vibrations

produced by the action of the truck loads (SLS) is presented in Figure 6.13. Again, the

serviceability limit state of deflection is fulfilled throughout the service life of the bridge.

Figure 6.13: Increment of the static deflections of the composite section caused by the live load.

To complete the analysis and design for durability of the prestressed girders, the performance of

the composite girder in terms of the FLS vs. time is presented in the Figure 6.14. The increase of

static vibration is again represented by a second-degree curve, similar to the increments of the

internal stresses in the composite section. As for the other limit states, the serviceability limit state

for vibration is fulfilled throughout the service life of the bridge. The stress variation started at

103MPa at the beginning of the service life, and reached a value of 109MPa at the end of the 150

years of the design service life, and was within the maximum allowable strand stress range of

125MPa.

As a general conclusion, the selection of the type of girder, as well as the careful design and

selection of the construction materials for the deck elements produce a satisfactory performance

of the composite slab-girder section during the design service life within the different limit states

considered in the design of the bridge superstructure for durability.


84

Figure 6.14: Increment of the stress variation on the prestressing strands of the girder.

6.2.1.12. Supplementary Protective Measures

Some additional measures can be implemented to enhance the performance of the girders and

to reduce the effects of the degradation mechanisms on the construction materials. One of

these measures is the waterproofing of the girder surfaces, which may be exposed to various

microclimates. This waterproofing can be established by means of sealing treatments, coatings

and/or membranes on the surfaces.

Further to the previous measures and the fact that the concrete of the girders has been

carefully designed to be as impermeable as possible, it is important to improve the protection of

the prestressing strands against corrosion. For this reason, regular steel for the passive

reinforcement of the girders is implemented, with the purpose of reducing the electrochemical

potential for corrosion of the prestressing strands.

The installation of long cornices at the edges of the bridge deck can provide additional

protection to the edge girder from the direct exposure to moisture, de‐icing salts, wetting and

drying and freezing and thawing cycles. These cornices can be designed on precast concrete, or

as a synthetic material resistant enough to withstand the effects of the microclimates in these


85

zones. Additionally, these cornices must be designed such that their installation, maintenance

and replacement can be executed easily.

6.2.2. Reinforced Concrete Design for Durability

Reinforced concrete elements, such as the bridge deck slabs, the concrete barriers, seismic

shear keys, and the foundation units, allow a more direct integration of the durability parameters

into the structural design equations. The definition of a model that integrates all of these

parameters is presented in the following sections.

6.2.2.1. General Design Parameters

A reinforced concrete section is assumed to be affected by a progressive deterioration of its

construction materials. The concrete will deteriorate depending on the controlling mechanism of

deterioration that affects the element. At the same time, the carbonation and chloride ingress

fronts move into the reinforced concrete section and initiate the active corrosion process at a

moment defined by the corrosion initiation time (t0) described in the previous sections.

Figure 6.15: Reinforced concrete section parameters for the structural and durability design.


86

The basic parameters that define the resistance of the reinforced concrete section are identified

in the Figure 6.15, where it is clear how these parameters are affected by the rates of

deterioration of the construction materials.

Figure 6.15 (a) presents the reinforced concrete section to perform an analysis per unit width of

the element (deck slab, barrier, beam, etc.), where the concrete covers on both sides are

subjected to different mechanisms of deterioration. This will be the case for a bridge deck slab

which is subjected to two microclimates with specific rates of deterioration taking place differently

at the top and bottom of the element. Figure 6.15 (b) represents the situation where the concrete

cover is completely deteriorated and then the reinforcing steel begins to corrode. The active

corrosion process starts once the initiation period for corrosion is over, and then it continues

according to the rates of corrosion and the other durability parameters that have been defined in

the Sections 5.1.3 and 5.1.4. Figure 6.15 (c) shows that corrosion first starts at the most exposed

point of the reinforcing bar and the corrosion will continue around the perimeter of the bar

according to the rate of corrosion 17. Once the bar corrosion has occurred over the entire

perimeter, corrosion will continue towards the inside of the rebar, reducing the cross section of

the bar in proportion to the rate of corrosion (Figure 6.15 (d)).

The parameters involved in the resistance of this reinforced concrete section are:

b = width of the reinforced concrete section.

h0 = initial height of the section.

d0 = initial effective height of the section.

c01 = initial concrete cover at the internal face of the section.

c02 = initial concrete cover at the external face of the section.

rc1 = depth of deterioration of concrete that takes place at the internal face of the section.

rc2 = depth of deterioration of concrete that takes place at the external face of the section.

ht = height of the section being affected by the rates of deterioration rc1 and rc2.

dt1 = effective height of the section being affected by the rate of deterioration rc1.

dt2 = effective height of the section being affected by the rate of deterioration rc2.

ct1 = concrete cover of the section affected by the rate of deterioration rc1.

ct2 = concrete cover of the section affected by the rate of deterioration rc2.


87

db1 = initial diameter of the reinforcing bars near the internal face of the section.

db2 = initial diameter of the reinforcing bars near the external face of the section.

dbt1 = diameter of the reinforcing bars affected by the rate of corrosion rs1.

dbt2 = diameter of the reinforcing bars affected by the rate of corrosion rs2.

rs1 = depth of corrosion of the reinforcing steel at the internal layer of reinforcement.

rs2 = depth of corrosion of the reinforcing steel at the external layer of reinforcement.

The suffixes 1 and 2 make reference to the side of the element being analyzed, such as the

internal or external faces of the reinforced concrete element, or the internal or external layers of

reinforcing steel. The general equations of the reinforced concrete section resistance do not

include these suffixes; however, they will be adjusted for every specific case of analysis and

design in terms of the durability and structural parameters corresponding to each one of the sides

of the element.

6.2.2.2. Flexural Design for Durability

The flexural resistance of a reinforced concrete section can be determined using the conventional

analysis for ultimate bending resistance based on strain compatibility and equilibrium using

material resistance factors and material properties. According to this the flexural resistance can

be expressed in terms of the following equation:

2'

121 t

cc

ystystn bd

f

ffM

(6-1)

where:

t = ratio of non-prestressed tension reinforcement = tst bdA .

Ast = area of reinforcing steel =

2

2

btd

.

fy = yield strength of non-prestressed reinforcing steel.

f’c = compressive strength of concrete.

s = resistance factor for non-prestressed reinforcing steel = 0.9. 1


88

c = resistance factor for concrete = 0.75. 1

1 = ratio of average stress in the rectangular compression block.

1 = 0.85 – 0.0015 f’c ≥ 0.67. 1, 23

According to the Clause 8.8.4.3, the flexural reinforcement must be designed such that the

resulting factored flexural resistance, Mn of the section is at least 1.20 times the cracking

moment, Mcr. However, this minimum reinforcement could be lowered if the factored flexural

resistance provided is at least one-third greater than the resistance required according to the

factored loads on the member. In accordance to Clause 8.8.4.4, the cracking moment is the one

that induces tensile cracking stresses, fcr in the concrete section. The cracking stress is

calculated according to the equation: 1

'4.0 ccr ff (6-2)

The cracking moment is the given by:

ct

gcrcr y

IfM ;

ct

gtcrcrt y

IfM (6-3)

where:

fcr = cracking strength of concrete , according to Clause 8.4.1.8.

Igt = second moment of area of the gross concrete section about its centroidal axis, neglecting the

reinforcing steel bars. The geometrical property of the concrete section changes with time

because of the reduction of the cross-section area of the slab as a result of the degradation

induced by the mechanisms of deterioration.

yct = the distance from the centroidal axis of the gross section to the extreme concrete fibre in

tension, neglecting the reinforcing steel bars. For the present case of the bridge deck slab,

222210 cct

ct

rrhhhy

(6-4)

6.2.2.3. Check for Shearing Resistance

Despite the fact that the CHBDC provides an empirical method for the design of the

reinforcement of the deck slab, a detailed analysis of the shearing resistance is performed. A


89

sectional design model is assumed to determine the shearing resistance of the reinforced

concrete section. A beam action with a critical section extending in a plane across the design

width and located at a distance, d, from the face of the reaction area, or from any change in slab

thickness is analyzed. Since no shear reinforcement is planned to be distributed in the slab, the

shearing resistance provided by the concrete itself must be adequate to withstand the external

factored shearing actions. According to Clause 8.9.3.4, the shearing resistance of concrete is

determined according to the following equation:

vcrcc bdfV 5.2 (6-5)

where:

dv = effective shear depth. According to Clause 8.9.1.5, the effective shear depth must be

considered as the larger of the values of 0.72h and 0.9d. Using the durability parameters defined

earlier, the effective shear depth is given by:

21072.072.0 cct rrhh (6-6)

101 9.09.0 ct rdd ; 202 9.09.0 ct rdd (6-7)

= a factor accounting for the shear resistance of cracked concrete. According to Clause 8.9.3.6,

for reinforced concrete sections not containing transverse reinforcement but having a specified

nominal maximum size of coarse aggregate not less than 20 mm, β shall be equal to:

vd1000

230 (6-8)

6.2.2.4. Deflection Analysis

The deflections of the deck slab caused by the external actions will increase with time due to the

ongoing degradation of the construction materials, according to the specific microclimates that

are developed in this part of the bridge. For this reason the derivation of the deflection formulation

will include the rates of deterioration of steel and concrete, and will result in an equation, which is

a function of time.

For a rectangular concrete section, the second moment of inertia can be calculated according to

the following equation:

12

3tt

gt

hbI (6-9)


90

The parameters for the second moment of area of the cracked transformed section, Icrt are

defined in Figure 6.16.

Figure 6.16: Cracked transformed concrete section for flexural analysis.

The transformed section is based on an additional equivalent area of concrete that is equivalent

to the reinforcing steel area in tension induced by flexure. The equivalent additional concrete

section is obtained by multiplying the reinforcing steel area by the modular ratio, n, which is

expressed by the following equation:

c

s

E

En (6-10)

where:

Es = the modulus of elasticity of the reinforcing steel = 200000MPa.

Ec = the modulus of elasticity of concrete. This parameter can be calculated in terms of the

compressive strength, f’c and the unit weight of concrete according to Clause 8.4.1.7 as:

5.1

'

230069003000

c

cc fE

(6-11)

Figure 6.17: Representation of the compatibility of stresses and deformations.


91

In accordance with the compatibility of deformations of the concrete section in the elastic range

(Figure 6.17), it is possible to formulate the required parameters for the analysis of deflections of

a reinforced concrete element under flexure.

In Figure 6.17,

cctct E ; sstst E

TC ; ststttct Abc 2

1

stttstc AcdEbcE 2

2

1 ; stttstc AcdEbcE 2

2

1

stttc

st Acd

E

Ebc 2

2

1 ; stttt Acdnbc 2

2

1 ; 02

1 2 tsttstt dnAcnAbc

b

dnAbnAnAc

tsttstst

t

2

142

(6-12)

Multiplying equation (6-12) by dt / dt at both sides of the equation, gives:

22

22 2

t

tststt

t

tstt

db

dbnAAnd

bd

dnAc

(6-13)

Since t

stt bd

A , we obtain:

ttttt nnndc 222 (6-14)

The second moment of area of the cracked transformed section is defined the by:

223

212 ttstt

tt

crt cdNAc

bcbc

I

; 2

3

3 ttstt

crt cdNAbc

I (6-15)

The deflections and rotations of the reinforced concrete section can be calculated using the

effective second moment of area, Iet, according to Clause 8.13.3.3.


92

gtA

crtcrtgtcrtet I

M

MIIII

3

(6-16)

Here, MA corresponds to the maximum bending moment in the concrete section at a load stage

for which the deflection is calculated. At this moment, it is possible to calculate the deflections by

means of the conventional structural analysis for deflection, using the effective second moment of

area, Iet, of the concrete section in the relevant deflection equations.

6.2.3. Deck Slab Design

The bridge deck slab has to be analyzed for positive and negative bending moments resulting

from loads applied on the different panels of the slabs. Moreover, the structural analysis has to

consider the bending moments induced in the longitudinal direction, resulting from the longitudinal

overhangs at the expansion joints of the abutments, and the condition of semi-continuity of the

bridge deck established at the intermediate support over the pier. The cantilevered parts of the

deck slab are analyzed for transverse negative bending moments that are generated as a result

of selfweight of the reinforced concrete elements and truck loads applied to the cantilever

portions, as well as the horizontal loads applied to the barriers.

6.2.4. Transverse Bending Moments in the Bridge Deck

The design of the internal segments of the bridge deck slab is carried out considering some basic

steps that are listed and explained as follows.


To determine the bending moments in the slab, it is necessary to identify the load cases that may

generate these actions. These load cases can be permanent loads, such as the self weight,

transitory loads, such as the truck loads, the strains and deformation effects, wind loads,

differential settlement; or exceptional loads, such as earthquake loads, stream forces, ice

accretion, and collision forces. However, the most relevant load cases for the analysis and design

of these elements are the dead loads and the truck loads.

According to Clause 5.7.1.7.1, concrete deck slabs that are supported on longitudinal girders may

be analyzed for transverse bending moments using the simplified elastic method, where the

maximum unfactored transverse negative and positive bending moments caused by the design

truck CL-625 in the portion of the slab between the outer girders may be determined by the

equation:


93

DLAPS

M eCL

110

6.08.0625 (6-17)

where:

M = the negative and positive transverse bending moment [kNm/m].

Se = the effective transverse span, considered to be the smaller of the centre-to-centre spacing of

the girder webs and the clear span between girder webs plus the deck thickness [m].The centre-

to-centre spacing of the girder webs = 2.095m, and the clear span between girder webs plus the

deck thickness = 1.915m + 0.225m = 2.175m. Therefore, Se = 2.095m.

P = the maximum wheel load of the CL-625 truck, which is 87.5kN.

DLA = dynamic load allowance, which is 0.4 because only one axle of the CL-625 truck is used

for the analysis.

Having defined the required parameters, the transverse bending moment caused by the truck

load is:

kNmkNm

M CL 41.264.0110

5.876.0095.28.0625

The dead load effects can be determined in terms of the self weight of the slab. The permanent

loads corresponding to the bridge deck slab are:

- Selfweight of the reinforced concrete slab: (0.225m)(24kN/m3)(1m) = 5.4kN/m.

- Waterproofing membrane and asphalt concrete: (0.065m)(23.5kN/m3)(1m) = 1.53kN/m.

The distributed load on a one-metre-wide slab strip results: w = 5.4kN/m+1.53kN/m = 6.93kN/m.

For the internal segments of a multi-span bending element, the resulting bending moments at the

supports and at the midspan can be calculated as:

- A the supports:

kNmmmkNwS

M eD 53.2

12

095.293.6

12

22

- At midspan:

kNmmmkNwS

M eD 27.1

24

095.293.6

24

22


94

According to Clause 3.5, the design bending moment for this element must be calculated

considering some load factors, which are 1.2 and 1.7 for the dead load and the live load

respectively, giving as a result the following design bending moment:

- At the supports: kNmkNmkNmM u 93.4741.267.153.22.1

At midspan: kNmkNmkNmM u 42.4641.267.127.12.1

6.2.4.2. Durability Parameters

The durability design parameters were determined in the Section 5.1.4. A summary of these

parameters for the bridge deck slab are listed in Table 6.13.

Element Rates of

deteriorarion (mm/year)

Cocrete Top 0.411

Bottom 0.131 Reinforcing

steel Top 0.031

Bottom 0.019 Carbonation coefficient Location Kc

Top 0.92 Bottom 1.17

Initiation time for corrosion

Location (years)

Top 18 Bottom 25

Table 6.13: Durability design parameters of the bridge deck slab.

6.2.4.3. Initial Conditions of the Bridge Deck Slab

The general conditions of bridge deck need to be assumed initially and then improved after

several iterations during the design for durability process in the same way that was used for the

prestressed concrete girders. The conditions of the bridge deck slab immediately after

construction are summarized in Table 6.14.

6.2.4.4. Assumption for Steel Reinforcement and Performance with Time

Starting with the minimum reinforcement provisions for flexure, a preliminary reinforcement is

assumed for the slab. Then, after having performed several iterations with the structural and

durability models for this element, it is possible to adjust and improve the rebar distribution for its

adequacy.


95

Table 6.14: Initial conditions of the bridge deck slab. The reinforcement of the slab is established as 20M @ 0.1m in the top layer, and 20M @ 0.15 in

the bottom layer, for the transverse (principal) reinforcement.

Figure 6.18: Bridge deck slab resistance for positive bending moments vs. time.

Figure 6.19: Bridge deck slab resistance for negative bending moments vs. time.


96

As for the longitudinal reinforcement, it is assumed to be two thirds of the transverse

reinforcement, in accordance to Clause 5.7.1.7.1, and Clause 8.2.2.2 of the MTQ Manual 22. This

means that the longitudinal reinforcement is 20M @ 0.15m at the top and bottom layers.

Considering the durability parameters, initial slab properties, and reinforcement distribution, the

performance of the bridge deck slab with time is shown in Figures 6.18 and 6.19.

It is evident from Figures 6.18 and 6.19 that with the assumed reinforcement distribution, the

adequate selection of materials, and the careful design of the concrete mixture, the slab presents

an adequate performance throughout the service life of the bridge.

6.2.4.5. Final Design and Details

The summary of the deck slab reinforcement for the internal sections is shown in Figure 6.20.

Figure 6.20: Detail of the deck slab reinforcement.

6.2.5. Transverse Bending Moments in the Cantilever Overhang

The design of the cantilever overhang of the bridge deck slab is carried out similarly considering

some basic steps as follows.


According to Clause 5.7.1.6.1.1, it is possible to calculate the design moment intensity due to the

CL-625 truck for a cantilever slab of constant or linearly varying thickness, using Table 6.15,

which is accompanied by Figure 6.21, and explains the basic parameters that determine the

bending moments generated in the deck slab overhang.

According to the bridge deck characteristics defined in Chapter 2, the transverse distance from

the longitudinal external edge of the bridge deck to the supported edge of cantilevered slabs

located at the outside face of the web of the external girder, Sc is equal to 1.16m. Considering

that the thickness of the slab remains constant across the bridge deck, and a cast-in-place barrier


97

is provided at the edge of the slab; the maximum cantilever moments due to the unfactored

CL-625 truck wheel loads results in a bending moment of 34kNm/m, including the dynamic load

allowance.

Table 6.15: Maximum cantilever moments due to unfactored CL-625 truck wheel loads including the DLA (kNm/m). 1

Figure 6.21: Notation for cantilever moments. 1

The dead load effects can be determined in terms of the self weight of the slab and the concrete

barrier placed at the edge of the deck. The permanent loads corresponding to the bridge deck

slab are:

- Selfweight of the reinforced concrete slab: (0.225m)(24kN/m3)(1m) = 5.4kN/m.

- Selfweight of the concrete barrier: (0.35m2)(1m)(24kN/m3) = 8.4kN.

- Waterproofing membrane and asphalt concrete: (0.065m)(23.5kN/m3)(1m) = 1.53kN/m.

The distributed load on a one-metre-wide slab strip results: w = 5.4kN/m+1.53kN/m = 6.93kN/m.

The resulting bending moments at the supported edge of the cantilever overhang of the slab can

be calculated as:


98

2

435.0

2

435.0 2 mSP

mSwM c

cD (6-18)

Therefore,

kNmm

mkNmmmkN

M D 74.92

435.016.14.8

2

435.016.193.6 2

According to Clause 3.5, the design bending moment for the slab overhangs must be calculated

considering some load factors, which are 1.2 and 1.7 for the dead load and the live load

respectively, giving the design bending moment as:

kNmkNmkNmM u 49.69347.174.92.1


For the cantilever part of the deck slab, the durability design parameters remain the same as the

ones described previously for the internal panels of the slab in Section 6.2.4.2.


Similarly, the initial physical conditions of the cantilever overhang of the bridge deck slab

correspond exactly to those described for the rest of the slab in Section 6.2.4.3.

6.2.5.4. Assumption for Steel Reinforcement and Performance with Time

The calculated factored design bending moment can be used to determine the minimum required

reinforcement. Then, this reinforcement provision can be refined by using the integration of the

structural and durability models, following an iterative process.

Adequate reinforcement for the slab at the cantilever overhangs works out to be 20M @ 0.10m in

the top layer, alternating the rebars with the previous bar distribution of 20M @ 0.10m that was

provided for the upper layer of the internal panels. This results in a rebar distribution of 20M @

0.05m for the cantilever part of the slab. The lower layer and the longitudinal reinforcement

remains the same as the one established for the rest of the slab. This reinforcement in the bridge

deck slab results in the performance with time as shown in Figure 6.22.

It is evident from Figure 6.22 that the proposed design for the slab performs adequately with time,

ensuring adequate bearing capacity throughout the design service life. At the end of the 150

years of design service life, the element still has some residual resistance of about 15kNm.


99

Figure 6.22: Bridge deck slab flexural resistance at the cantilevers overhang. Following the same multi-stage protection strategy that has been used for the rest of the bridge

elements, additional protective measures will help to reduce the impact of the deterioration

mechanisms on the most vulnerable parts of the bridge deck slab. These supplementary

protection measures are described in Section 6.2.9.

6.2.5.5. Final Design and Details

The summary of the deck slab reinforcement for the internal sections is shown in Figure 6.23.

Figure 6.23: Reinforcement details for the bridge deck slab overhangs. 6.2.6. Transverse Vertical Shear

The verification of shearing resistance is necessary to ensure that the bridge deck slab can

withstand the induced shearing forces without any shearing reinforcement during the required

service life.


100


The shear stresses in the slab are calculated considering the permanent loads and the truck

loads that act on the bridge deck. The dead or permanent load generates the following vertical

shear:

- At the internal sections of the slab: kNmmkNVD 27.13915.1/93.6 .

- At the cantilever overhangs: 16.44kN16.1/93.64.8 mmkNkNVD .

The transverse vertical shear on the bridge deck slab that is produced by the truck loads is

determined by the effects of the wheel loads of the heaviest axle moving across the bridge lanes.

This loading condition is shown in Figure 6.24.

Figure 6.24: Axle loads moving across the deck lanes. Using the influence line theory, it is possible to determine the position of loads that generates the

maximum internal shear forces in the slab. The analysis is aimed of determining the design

shearing force at the intermediate support.

The relevant influence line is presented in the figure 6.25 (a). The loading location and the

consequent shear diagram required by the influence line in Figure 6.25 (a), is shown in figure

6.25 (b). From these diagrams, it is possible to identify the maximum shear force,

VTruck = 89.04kN, induced in the slab due to the truck loads.

This load must be increased by the dynamic load allowance of 0.4 considering that a single axle

is considered in this analysis. The maximum factored shearing force induced in the slab is

determined considering the applicable load factors as:

kNkNkNVu 49.25096.964.017.144.162.1


101

Figure 6.25: a) Influence line for shear forces in the slab at the intermediate support. b) Shear force diagram for the loading location that generates the largest shear force at the intermediate support.


The durability design parameters remain the same as the ones described previously for the

flexural design of the slab in Section 6.2.4.2.


The initial physical conditions of the slab correspond exactly to those described in Section 6.2.4.3.

6.2.6.4. Shearing Resistance Performance with Time

The shearing resistance of concrete is a property that varies with time due to the progressive

deterioration of the slab caused by the relevant mechanism of deterioration, and hence causing a

progressive reduction of the slab effective height that accounts for the shearing resistance. The

performance of the bridge deck slab shearing resistance with time is presented in Figure 6.26.


102

Figure 6.26: Bridge deck slab shearing resistance vs. time.

It is clear from Figure 6.26 that the bridge deck slab performs satisfactorily in shear throughout

the service life. As for the rest of the bridge elements, some supplementary protection measures

are adopted to decrease the rates of deterioration and the loss of slab shear resistance.

6.2.7. Analysis of Deflections with Time

The analysis of deflections is based on the principles outlined in Section 6.2.2.4. According to

these design assumptions, the deflections produced by the dead and live loads will increase with

time due to the progressive deterioration of the section, with a reduction of the height of the

reinforced concrete slab and the corrosion of the reinforcing steel bars, which decreases the

rigidity of the of the concrete element, and consequently its stiffness against deflection. The slab

deflections are evaluated at the cantilever overhangs, which is the part of the slab that presents

the maximum values of deflections produced by the acting loads. The maximum allowable

deflections is considered to be (2Sc) / 240 which is the limit value established for slabs supporting

non-structural elements (reinforced concrete barriers) not likely to be damaged by large

deflections, according to the CSA A23.3 Standard 23.

The performance of the slab for the serviceability limit states in terms of deflections is presented

in Figure 6.27, which shows that after 135 years of service life the bridge deck slab will reach the

limit of its serviceability limit state in terms of deflections at the cantilever overhangs. However, an

increase of the slab section in this part of the deck is not required. Considering that some

supplementary protection measures, such as waterproofing coatings and membranes are being


103

adopted, the degree of deterioration over the slab will be reduced, extending the service life of

this bridge element.

Figure 6.27: Deflections on the cantilever overhang of the bridge deck slab.

6.2.8. Design of the Semi-continuity of the Bridge Deck Slab

The bridge superstructure is designed as a series of precast, prestressed concrete girders which

will resist their own dead load over a simple span. However, continuity will be established at

intermediate supports after the girders have been installed. These three-span continuous

systems will then resist the live loads (trucks) and any other superimposed loads as a continuous

girder with negative bending moments over the intermediate piers. This continuity is implemented

by providing a specially designed diaphragm over the intermediate piers along with high-strength

steel dowels. Additional reinforcement is arranged at the top fibres of the deck slab to provide the

necessary resistance for negative moments due to live loads. Moreover, reinforcement is

provided at the lower part of the girder ends to provide continuity at the bottom fibres at the

intermediary support.

For this bridge, the girders have been designed to withstand not only the dead loads, but also the

truck loads, following the philosophy of a durable design under the principles of multi-protection

approach design. This design assumption accounts for the event of the loss of continuity at the

intermediate supports under the event of deterioration, repair, modification or renovation of the

bridge deck. The main purpose for this semi-continuity setup is to eliminate the deck joint at the

pier supports and hence, eliminate the source of important mechanisms of deterioration that

would occur during the bridge service life if some expansion joints are installed at these locations.


104

These expansion joints over the piers have been a source of serious ingress of aggressive

elements and significant deterioration of girder ends, bearings, pier caps and piers; eliminating

these joints would cut off the ingress of aggressive elements and the resulting extensive

deterioration. The reinforcement for negative bending moments must be enough to provide at

least an appropriate crack control of the slab. However, the negative moment at the supports is

calculated based on the assumption of full structural continuity, following the recommendations of

Clause 8.19.4.3 1.


The maximum bending moment caused by the permanent loads and the truck can be determined

considering a three-span bridge continuously supported over the intermediate supports at the pier.

The permanent loads can be considered as uniformly-distributed loads that cause flexural

moments in the bridge deck: These uniformly distributed loads are:

- Slab: mkNmkNmm 38.812407.15225.0 3

- Gussets: mkNmkNmm 08.1072405.02.1 3

- Diaphragms:

mkNm

mkNmm81.8

263

562419.330.0 32

- Wearing surface: mkNmkNkNmm 69.215.23)20.14065.0 3

- Barriers: 332 80.162435.02 mkNmkNm

Therefore, the distributed permanent loads are wD = 138.76 kN/m. The maximum negative

bending moment produced by the permanent loads results -9377.49kNm, according to the

bending moment diagram presented at Figure 6.28.

Figure 6.28: Bending moments diagram for the bridge deck under permanent loads. The maximum bending moment caused by the truck live loads can be determined using influence

lines for the moments at the first intermediate support (Figure 6.29). This analysis allows the

identification of the precise location of the truck loads that generates the worst-case scenario for


105

the maximum negative bending moment at this support. Accordingly, the maximum negative

bending moment caused by the truck loads is -2233.85kNm.

Figure 6.29: a) Influence line for bending moments at the firs intermediate support of the bridge deck and location of the truck loads that generates the maximum negative bending moment at the support. b) Bending

moments diagram for the bridge deck under the truck loading condition. The design bending moment, considering the DLA and the load factors described in the CHBDC 1

results:

kNmkNmkNmM u 92.1599925.0185.22337.149.93772.1


The durability design parameters remain the same as the ones described previously for the

flexural design of the slab in the Section 6.2.4.2.

6.2.8.3. Initial Conditions of Semi-continuity of the Bridge Deck

The initial physical conditions of the bridge deck at the support correspond to the condition set by

the integration of the deck slab, the end-diaphragms and the girders of the bridge deck at the

intermediate supports. These initial conditions will be affected by the presence of the degradation

mechanisms over the bridge deck at the previously mentioned locations. The most aggressive

degradation mechanisms will take place at the top of the deck slab due to the action of the

surface deterioration, which is the controlling mechanism that determines the durability

parameters for the design of this part of the bridge deck. The degradation of the bridge deck at


106

the top of the slab will produce a loss of effective height and hence, a loss of bearing capacity for

negative bending moments at the support.

The conditions of the bridge deck at the intermediate supports immediately after construction are

described in Table 6.16.

Table 6.16: Initial conditions of the bridge deck at the intermediate supports. The determination of the bridge deck reinforcement for the negative bending moments under the

semi-continuity condition at the intermediate supports is determined after an iterative process

following the guidelines for a structural design for durability.

After several iterations, the adequate deck reinforcement for negative bending moments at the

intermediate supports results in a distribution of 25M rebars @ 0.15m (Figure 6.30). This

reinforcement at the top of the deck is extended up to the point where no negative reinforcement

is required. After having analyzed the bending moment diagrams for permanent and truck loads,

it can be established that the negative moment reinforcement must be extended 7m on each side

of the deck from the axis of the pier, then it will be overlapped with the top longitudinal

reinforcement of the slab for the intermediate portion of the deck.

Figure 6.30: Detail of the bridge deck reinforcement for the semi-continuity condition at the intermediate supports.


107

The positive moment reinforcement over the supports is proportioned for structural continuity to

resist the moments due to creep, shrinkage, temperature change, and live load in farther spans.

According to Clause 8.19.4.2, a minimum reinforcement for positive bending moments cannot be

lower than (1.5 x 1600) mm2 at the location of every girder. This reinforcement corresponds to

2400mm2 that can be distributed among 4 u-shaped 20M rebars that are adequately embedded

in the bottom flange of the girders beyond the strand transfer length and anchored into the

diaphragm over the continuity supports.

6.2.8.4. Semi-continuity Performance of the Bridge Deck with Time

The performance of the bridge deck reinforcement resistance for negative bending moments with

time is presented in Figure 6.31.

Figure 6.31: Bridge deck bearing capacity for negative bending moments with time.

Figure 6.31 shows that the semi-continuity of the bridge deck is satisfactorily available until about

140 years of service life. There is a slight gain of bearing capacity for negative bending moments

over the first 20 years of use, which is caused by the loss of the concrete cover in top of the

bridge deck slab that generates a loss of weight and hence, a reduction of dead load at this

location of the bridge. However, after the year 20, the initiation of active corrosion begins and

then, a reduction of the rebar cross-sections occurs, continuously reducing the flexural resistance

of the deck section for negative bending moments. The service life of the semi-continuity

condition of the bridge deck can be increased by applying additional protection measures that are

described in Section 6.2.9, and by ensuring an adequate maintenance strategy.


108

6.2.9. Supplementary Protective Measures

The performance of the bridge deck slab design for durability as detailed and reviewed in the

previous sections confirms satisfactory performance of the deck throughout the design service life

of the bridge. However, following the principle of a multi-front protection strategy, some

supplementary protection measurements are adopted, including the use of galvanized steel,

which may reduce significantly the rates of corrosion, extending the satisfactory behaviour of the

structural elements beyond the required service life.

The installation of waterproofing coatings and membranes attached on top of the concrete slab

are required for the purpose of reducing the direct exposure to wetting and drying cycles, and

freezing-and-thawing effects, preserving the top concrete surface for longer time. This will

diminish the ingress of moisture and solutions carrying de-icing salts, delaying the initiation time

for active corrosion to occur.

Additionally, it is planned to use microfilament polypropylene fibres in the concrete mixture of the

bridge deck slab; these fibres help to control plastic shrinkage and settlement cracking.

Additionally, they may improve impact, shatter and abrasion resistance. They enhance durability

and toughness of concrete, and may also help to reduce bleeding.

6.3. Substructure Design

The structural design of the substructure elements for durability is focused on the intermediate

piers. The piercaps will be subjected to different load cases that generate different flexural

conditions depending on the way that these loads are applied. The design loading conditions may

include different load cases, such as dead loads, live loads, earthquake effects, wind loads, wind

effects over the vehicles, braking forces from the traffic, and the effects caused by strains,

deformations and displacements suffered by the bearings placed on top of the piercap, caused by

temperature change and temperature differentials, concrete shrinkage, differential shrinkage and

creep. The flexural conditions on the piercaps may be generated in different planes of the

element, resulting in a condition of biaxial flexure. For these reasons, adequate reinforcement

must be provided for this part of the piers, considering as well the durability considerations

required for these elements over the service life of the bridge. The previously developed

procedure for flexural design of reinforced concrete for durability can be used for this purpose.

The lower part of the piers composed by the pair of columns can be subjected to flexural and

axial loads, caused by the combination of the previously described load cases. For the purpose of

the present bridge design, the substructure design will be focused on durability of the pier

columns.


109

6.3.1. Pier Column Design

The structural design for durability of pier columns is presented using a different approach due to

the complexity of the model of reinforced concrete columns subjected to flexural and axial loads

simultaneously. This approach is addressed by analyzing in an iterative manner, the different

reinforced concrete sections of the columns that are being deteriorated by the various

degradation mechanisms, identifying their bearing capacity against the applied loads by using the

interaction diagrams of the different column sections for flexure and axial loads.


A detailed description and analysis of the different load cases considered for the design of the

pier column of the bridge is presented in the following sections.

6.3.1.1.1. Dead Loads (D)

The dead loads that affect the foundation units comes from the self weight of the bridge elements

of the bridge that are involved in the tributary area that accounts for each one of the intermediate

piers. These loads are:

- Self weight of the slab: kNmkNmmm 83.2115242607.15225.0 3

- Self weight of the gussets: kNmkNmmm 06.259142485.122.105.0 3

- Selfweight of the barriers: kNmkNmm 8.4362242635.0 32

- Self weight of diaphragms:

Intermediate: kNmkNmmm 71.2333624915.125.013.1 3

End: kNmkNmmmm 41.748224730.0589.0675.019.3 322

- Selfweight of the girders: kNmmkN 81.24982785.1289.13

- Selfweight of the wearing surface: kNmkNmmm 95.5635.232620.14065.0 3

- Selfweight of the pedestals: kNmkNmmm 21.537243.29.0153.0 3

- Selfweight of the piercap: kNmkNmm 6.1407243.250.25 32


110

- Selfweight of the pier columns: kNmkNmm 78.90422461 32

The total dead load of the structure acting on each one of the intermediate piers is D =

9222.15kN.

6.3.1.1.2. Live Loads (L)

The live loads considered for the design of the pier columns correspond to the case where a

maximum number of trucks can be located near an intermediate pier within the tributary area of

the foundation unit. In this case, it is assumed that both of the traffic lanes are occupied with

trucks traveling on both directions of the bridge. According to the geometry of the bridge deck, it

is possible to arrange one complete design truck CL-625, plus the first and the last load axles

from the other trucks located in front and behind of the truck considered. This loading condition

can be found on each one of the four traffic lanes of the bridge. Therefore, the maximum live load

acting on the intermediate piers is:

kNkNkNkNkNkNkNkNL 330045015017512512550150

6.3.1.1.3. Wind Loads (W)

According to the recommendations in he CHBDC 1, the wind loads can be calculated according to

the following equations:

hgeh CCqCF ; vgev CCqCF (6-19)

where:

Fh = the horizontal wind pressure on the structure.

Fv = the vertical wind pressure on the structure.

q = the hourly mean reference pressure of 400Pa for a return period of 50 years.

Ce = the wind exposure coefficient of 1.0, according to the height of the superstructure.

Cg = the gust effect coefficient of 2.0, according to the span length of the bridge.

Ch = the horizontal wind load coefficient of 2.0.

Cv = the vertical wind load coefficient of 1.0.

Therefore, the resulting vertical and horizontal wind pressures are:


111

kPaPaPaFh 6.116000.20.20.1400

kPaPaPaFv 8.08000.10.20.1400

These wind pressures are applied horizontally and vertically over the exposed areas of the bridge

in a perpendicular direction to the longitudinal axis of the bridge. The vertical and horizontal wind

loads are considered to act simultaneously. The vertical wind load is considered to be applied

downwards or upwards, resulting in two different load cases. To consider a possible eccentric

effect, the vertical load resultant is applied at the quarter point of the transverse bridge deck width.

According to the geometrical characteristics of the bridge deck and the eccentricity for vertical

wind load is 3.77m. The intermediate piers are designed for wind-induced loads transmitted by

the bridge deck, and for wind loads acting directly on them. The foundation units are designed for

directly applied horizontal wind forces calculated with a horizontal wind load coefficient (Ch) of 0.7

for circular pier columns, according to Clause 3.10.3.3.

Considering the vertical and horizontal surfaces of the bridge involved in the tributary areas of the

intermediate piers, the resulting wind loads are:

- Horizontal wind force on the bridge deck:

kNmmkPaW deckh 8.1242636.1 (Load case: WHD)

- Vertical wind force on the bridge deck:

kNmmkPaW deckv 46.3132607.158.0 (Load case: WVD)

- Lateral wind force on the pier:

kNmkPaW pierh 85.702.147.00.20.14.0 2 (Load case: WHP)

- Frontal wind force on the pier:

kNmkPaW pierhf 17.2595.447.00.20.14.0 2 (Load case: WFP)

6.3.1.1.4. Wind Loads on Traffic (V)

According to the CHBDC 1, the wind effects on the moving vehicles are calculated from the

applied wind pressure, which is determined using a horizontal wind load coefficient (Ch) of 1.2.

This wind pressure is applied over the exposed surface of the design truck. The resulting wind

force over the vehicles is:


112

mkNmkPaW vehiclesh 88.232.10.20.14.0

This load should be applied over the entire length of the superstructure. According to this, the

resulting wind force on the intermediate piers resulting from the traffic over the bridge is:

kNVpier 37.82

6.3.1.1.5. Stream Pressure Loads (F)

Following the recommendations of the CHBDC1, the load acting longitudinally on the intermediate

piers of the bridge can be calculated according to the following equation:

2

2AvCP D (6-20)

where:

P = the total load due to flowing water acting on the pier columns in the direction of the

longitudinal axis of the piers [N].

CD = the longitudinal drag coefficient of 0.7 for circular pier columns.

= the density of water of 1000kgf/m3

A = the area of a pier column exposed to the flowing water, projected parallel of the longitudinal

axis of the pier. This area is 6m2.

v = water velocity of the design flow. It is taken to be 0.55m/s.24

The lateral load on the pier columns, generated from the water flow depends on the angle

between the water flow and the longitudinal axis of the piers. This force can be determined by the

following expression:1

2

2HLvCP L

p

(6-21)

where:


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Pp = the total load due to flowing water acting on the pier columns in the horizontal direction

perpendicular to the longitudinal axis of the piers [N].

CL = the lateral load coefficient of 0.7 for an angle between the water flow and the longitudinal

axis of the piers of 8.5º.

H = depth of the flowing water at the pier equal to 3.5m.

L = Length of the pier along the longitudinal axis. This length will be considered as 2m for each

column.

Having defined these parameters, the stream pressure loads acting on the pier columns are:

kNN

smmmkgfP 635.025.635

2

55.0610007.0 223

(Load case: FH)

kNN

smmmmkgfPp 741.013.741

2

55.025.310007.0 23

(Load case: FF)

6.3.1.1.6. Ice Loads

According to the location of the bridge crossing a river, the foundation units placed in the water

can be subjected to significant forces due to ice effects during the winter season. The ice effects

can be related with the pressures developed on the piers due to moving ice, ice impact forces, ice

jams, ice adhesion and ice accretion. The loading cases are described in the following

subsections.

A. Pressures Due to Moving Ice (F)

Following the recommendations of the CHBDC 1, the resulting horizontal force caused by moving

ice can be determined by the following expressions:

The lesser of bF or cF → for 0.6tw

F (6-22)

cF → for 0.6tw

where: 2ptCF nb ; ptwCF ac

where:

Fb = the horizontal ice load caused when ice floes fail by flexure [kN].


114

Fc = the horizontal ice load caused when ice floes fail by crushing [kN].

Cn = the coefficient of pier nose inclination. This coefficient is equal to 0.133.1

Ca = the coefficient allowing for the ratio of the pier width to ice thickness when the ice fails by

crushing. This coefficient is equal to 1.41.1, 24

p = the effective crushing strength of ice of 1500kPa, when ice movement occurs at temperatures

below the melting point.1

t = the ice thickness expected to make contact with the piers. This is equal to 0.4m.24

w = the frontal pier width at the level of ice action where ice is split or crushed, perpendicular to

the direction of the ice motion. For the analyzed pier columns, this value is equal to 2m.

Therefore, the ice load on the pier columns caused by moving ice is:

Fm

m

t

w 5

4.0

2is the lower value between bF and cF

kNmkPaFb 92.314.01500133.0 2

kNF 92.31

kNmmkPaFc 169724.0150041.1

B. Ice Impact Forces (FII)

To account for ice impact forces, the following cases are analyzed:

- Case 1: A longitudinal force F (load case: FIIH) with a transverse force equal to 0.15F (load

case: FIIF).1

- Case 2: A longitudinal force 0.5F (load case: FIIH) with a transverse force equal to 0.34F

(load case: FIIF).1

C. Ice Jams (FIJ)

Considering the span lengths of the bridge of 26m, some floating ice accumulations can be

developed generating some pressures of the order of 10kPa 1 on the exposed pier columns, and

applied longitudinally and laterally to the pier orientation above the level of still water for the

expected thickness of ice jam. For the actual conditions of the bridge, ice jams are expected to be

developed on a thickness around 0.75m 25 around the pier columns. According to this, the ice

force that can be generated on the foundation units results to be:


115

kNmmkPaFIceJams 1575.0210 (Load cases: FIJH and FIJF)

D. Ice Adhesion (FIA)

According to the CHBDC1, when there exist fluctuations of the water levels, the forces generated

on circular pier columns frozen to an ice formation can be calculated using the following equation:

75.02 13.005.11250 tRtFv (6-23)

where:

Fv = the vertical force on a bridge pier caused by to ice adhesion [kN].

t = the ice thickness expected to make contact with the piers equal to 0.4m.24

R = radius of the circular pier column. This is equal to 2m.

Therefore, the ice adhesion force generated on each pier columns is:

kNm

mmFv 39.313

4.0

213.005.14.01250

75.0

2

(Load case: FIA)

E. Ice Accretion (A)

The ice accretion loads can be developed on all exposed surfaces of the bridge deck. The

thickness of the ice accretion depends on the geographical location of the bridge. For this

purpose, it is possible to obtain the ice thickness accretion from Figure 5.2.1 According Figure 5.2,

the ice accretion thickness for the Montreal region corresponds to 31mm.

Considering the exposed surfaces of the bridge as the sum of the surfaces of the bridge deck

slab, the concrete barriers and the external faces of the edge girder, the resulting ice accretion

load applied vertically on the bridge deck is:

kNmkNmmF AccretionIce 62.651807.9031.057.2143 32

6.3.1.1.7. Earthquake Loads (EQ)

Based on the considerations for earthquake load analysis for multi-span bridges presented in the

CHBDC 1, the equivalent uniformly distributed static seismic loading can be determined from the

following equation:


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L

WCP sm

e (6-24)

where:

W = the effective weight of the bridge. According to the Section 6.3.1.1.1, the weight of the bridge

is 29978.86kN, say 30000kN.

L = the total length of the bridge, which is 78m.

Csm = the elastic seismic response coefficient. This coefficient can be determined using the

equation: 1

AI

T

SACsm 5.2

2.132

(6-25)

where:

A = the zonal acceleration ratio, which is considered to be 0.20 for the Montreal metropolitan area.

S = the site coefficient. This parameter depends on the type of soil where the foundation of the

bridge structure is built. For the current bridge design, the foundation stratum for the substructure

corresponds to profile of stiff clays, sands and gravels with a depth varying from 70m to 85m.

According to Clause 4.4.6.1, the foundation soil corresponds to a soil profile Type I, which is

related to a site coefficient S = 1.2.

I = the importance factor, which for a lifeline bridge (Section 2.1), has a value of 3.0.

T = the natural period of the structure [s]. This period can be calculated according to the

equation:1

gK

WT 2 (6-26)

where:

g = the acceleration due to gravity equal to 9.807m/s2.

K = the lateral stiffness of the bridge. This parameter can be calculated from the following

equation: 1


117

max,s

o

V

LpK (6-27)

where:

po = an arbitrary uniform lateral load applied on the structure.

Vs,max = the maximum static displacement of the bridge due to the arbitrary load po.

The lateral stiffness of the bridges results to be:

mkN

m

mmkNK 4875000

0016.0

78100

Therefore, the natural period of the bridge structure is:

smkNsm

kNT 157.0

4875000807.9

86.299782

2

Therefore, the elastic seismic response coefficient is:

686.0157.0

2.12.02.132

sCsm ; 5.10.32.05.25.2 AI

Thus, 686.0smC

Finally, the uniformly distributed static seismic load is:

mkN

m

kNPe 75.263

78

86.29978686.0 (Load cases: EQH and EQF)

According to the provisions of Clause 4.4.9.2, the elastic seismic effects on the principal axes of

the piers resulting from analyses in the two perpendicular and principal horizontal directions must

be combined in such a way that 100% of the static seismic resulting force is applied in one of the

principal horizontal direction, simultaneously with the application of 30% of the resulting seismic

force in the other perpendicular principal direction. The load cases consider the application of

100% of the resulting seismic force in each one of the principal directions of the piers, in the

different senses, and in each one of the considered directions. This is reflected in the load

combination list that is presented in Section 6.3.1.1.9.


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6.3.1.1.8. Vessel Collision Forces (H)

The bridge must remain open to all traffic after a vessel collision (Section 2.1). This consideration

classifies this structure as a Bridge Class I, according to the CHBDC 1.

According to the CHBDC, the vessel collision forces on the bridge can be divided in two groups:

vessel collision forces on the substructure, and vessel collision forces on the superstructure. The

ship collision force on the pier columns can be calculated using the equation: 1

4.85.0 VDWTPs (6-28)

where:

Ps = the equivalent static vessel collision force [MN].

DWT = the dead weight of the vessel [t].

V = design collision velocity [m/s].

For the current design case, the dead weight of the vessel is assumed to be equal to 2650lbf

(1.20t), which corresponds to the upper limit of the small-vessel category. The collision impact is

considered to be to the order of 45mi/h = 20.11m/s, according to the selected type of vessel.

Therefore, the static vessel collision force applied on the pier columns is:

kNMNsmtPs 59.262762.24.8/11.2020.1 5.0 (Load cases: HFH and HFF)

The analysis of the loads resulting from a ship collision with the superstructure can be made

based on the following collision scenarios: 1

- Collision with bow: 1 SBHBH PRP (6-29)

where:

PBH = the ship collision force [MN].

RBH = the ratio of exposed superstructure depth to the total bow depth. According to the

characteristics of the type of boat assumed for design, there are no potential exposed surfaces.

For the small-vessel category, the bow height of a typical boat is much smaller than the bridge

clearance for navigation.

For this reason, the collision bow force is not going to be considered.


119

- Collision with deck house: 1 SDHDH PRP (6-30)

where:

PDH = the vessel deck collision force.

RDH = the reduction factor calculated using the following equation:

2.010.0100000

2.12.010.0

1000002.0

DWTRDH

Therefore, the vessel deck collision force is:

kNkNPDH 52.52559.26272.0 (Load case: HDH)

- Collision with mast: 1 DHMT PP 10.0

kNkNPMT 55.5252.52510.0 (Load case: HMT)

For pier design, the collision forces are applied as equivalent static forces, considering 100% of

the force acting in the direction parallel to the alignment of the centreline of the navigable channel,

and simultaneously, 50% of the force acting normally to this direction.

The superstructure must design for an equivalent static impact force applied perpendicularly to

the bridge deck elements vulnerable for collision, according to Clause A3.3.8.2.

6.3.1.1.9. Load Combinations

The load factors and load combinations for the design of the bridge structure are the ones

described in Clause 3.5 of the CHBDC 1. These combinations are reproduced here for

completeness.

- Fatigue Limit State:

1. D + L

- Serviceability Limit States

1. D + 0.9L 2. 0.9D

- Ultimate Limit State:

1. 1.2D + 1.7L


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2. 1.2D + 1.6L 3. 1.2D + 1.4L + 0.5WHD + 0.5WVD + 0.5WHP + 0.5WFP + 0.5 4. 1.2D + 1.4L - 0.5WHD - 0.5WVD - 0.5WHP - 0.5WFP - 0.5 5. 1.2D + 1.4L + 0.5WHD - 0.5WVD + 0.5WHP - 0.5WFP + 0.5 6. 1.2D + 1.4L - 0.5WHD + 0.5WVD - 0.5WHP + 0.5WFP - 0.5 7. 1.2D + 1.4L + 1.65WHD + 1.65WVD + 1.65WHP + 1.65WFP + 1.65 8. 1.2D + 1.4L - 1.65WHD - 1.65WVD - 1.65WHP - 1.65WFP – 1.65 9. 1.2D + 1.4L + 1.65WHD - 1.65WVD + 1.65WHP - 1.65WFP + 1.65 10. 1.2D + 1.4L - 1.65WHD + 1.65WVD - 1.65WHP + 1.65WFP - 1.65 11. 1.25D + EQH + 0.3EQF 12. 1.25D + 0.3EQH + EQF 13. 1.25D - EQH - 0.3EQF 14. 1.25D - EQH + 0.3EQF 15. 1.25D + EQH - 0.3EQF 16. 1.25D - 0.3EQH - EQF 17. 1.25D - 0.3EQH + EQF 18. 1.25D + 0.3EQH - EQF 19. 1.2D + 1.3FH + 1.3FF 20. 1.2D - 1.3FH + 1.3FF 21. 1.2D + 1.3FH - 1.3FF 22. 1.2D - 1.3FH - 1.3FF 23. 1.2D + 0.9WHD + 0.9WVD + 0.9WHP + 0.9WFP + 0.9V + 1.3A 24. 1.2D - 0.9WHD - 0.9WVD - 0.9WHP - 0.9WFP - 0.9V - 1.3A 25. 1.2D + 0.9WHD - 0.9WVD + 0.9WHP - 0.9WFP + 0.9V - 1.3A 26. 1.2D - 0.9WHD + 0.9WVD - 0.9WHP + 0.9WFP - 0.9V + 1.3A 27. 1.2D + 1.3FI 28. 1.2D – 1.3FI 29. 1.2D + 1.3FIIH + 1.3FIIF 30. 1.2D - 1.3FIIH + 1.3FIIF 31. 1.2D + 1.3FIIH - 1.3FIIF 32. 1.2D - 1.3FIIH - 1.3FIIF 33. 1.2D + 0.65FIIH + 2.95FIIF 34. 1.2D - 0.65FIIH + 2.95FIIF 35. 1.2D + 0.65FIIH - 2.95FIIF 36. 1.2D - 0.65FIIH - 2.95FIIF 37. 1.2D + 1.3FIJH + 1.3FIJF 38. 1.2D - 1.3FIJH + 1.3FIJF 39. 1.2D + 1.3FIJH - 1.3FIJF 40. 1.2D - 1.3FIJH - 1.3FIJF 41. 1.2D + 1.3FIA 42. 1.2D - 1.3FIA 43. 1.2D + HFH + HFF 44. 1.2D + HFH – HFF 45. 1.2D - HFH + HFF


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46. 1.2D - HFH – HFF 47. 1.2D + HDH 48. 1.2D – HDH 49. 1.2D + HMT 50. 1.2D – HMT 51. 1.35D

After analyzing the bridge structure according to the above load combinations, it is possible to

identify the maximum factored axial loads and bending moments acting in the principal directions

of the pier columns. These factored actions are:

kNPu 22.12998 ; kNmM xxu 75.6965 ; kNmM yyu 30.4598

These actions induced by the different load cases acting on the structure need to be evaluated in

terms of the bearing capacity of the columns, which is reflected in the interaction diagrams for

flexural compression of the columns.


The durability design parameters have been determined in Section 5.1.4. The controlling

mechanisms of deterioration are frost attack and abrasion by ice. A summary of the durability

parameters for the pier columns is presented in the following table:

Rates of deterioration (mm/year)

Concrete Ice abrasion 2.3835

Frost attack 0.1024

Reinforcing steel 0.002

Carbonation coefficient, Kc 2.36

Initiation time for corrosion 28 years

Table 6.17: Durability design parameters for the pier columns.

6.3.1.3. Initial Conditions of the Pier Columns

The initial conditions of the pier columns are assumed at first and then, after several iterations of

analysis, making the integration of the structural and durability parameters, it is possible to

identify the proper characteristics of the reinforce concrete sections. A summary of the general

initial conditions of the pier columns is presented in Table 6.18.

6.3.1.4. Assumptions for Reinforcement and Performance with Time

Initially, a reinforcement ratio of 1% of the cross-sectional area of the columns is assumed, which

results in 48 30M rebars that can be arranged in such a way that 36 bars are distributed on an


122

outer reinforcement layer and the remaining 12 rebars are arranged in an inner layer, as shown in

Figure 6.32. The performance of this reinforced concrete section under the conditions of flexural

compression and deterioration mechanisms is analyzed initially for abrasion by ice as the

controlling mechanism of deterioration. This performance is shown in the Figure 6.33, which

shows that the abrasion by ice on the pier columns produces a significant deterioration of the

reinforced concrete section, up to the point of inducing failure of the column at the ULS after 145

years of service life. This failure condition is related to flexural compression of the column with

respect to an axis direction perpendicular to the longitudinal axis of the pier (Mux-x, Pu).

Diameter Concrete

cover Material properties

D0 (m) c (mm) fy (MPa) f'c (MPa) c s2 75 400 35 0.75 0.9

Table 6.18: Initial conditions of the reinforced concrete section of the pier columns.

Figure 6.32.: Reinforcement of the pier column.

The piers of a bridge play a critical role in the integrity of the bridge structure, but at the same

time they are very difficult and costly to replace. Therefore, it is very important as a minimum

requirement, that these elements attain the required design service life of the bridge. For this

reason, it is required to limit the aggressive action of ice abrasion on the pier column by

implementing additional protection measures represented by the installation and construction of

protective islands around the piers, made of protective rock layers placed and arranged to

diminish the flow around the pier columns.

Once these protective islands are installed, the controlling mechanism of deterioration acting on

the pier columns is frost attack. The performance of the pier column over the design service life,

X X

Y

Y


123

considering the effect of frost attack, is presented in the Figure 6.34, which shows that the

performance of the pier column over the design service life is satisfactory after having analyzed

and integrated the structural resistance and durability considerations. At the end of the service life

there is enough resistance of the column against the ultimate flexural compression conditions

developed on the two principal directions of the columns (Mux-x, Pu; Muy-y, Pu).

Figure 6.33: Interaction curves for pier column for abrasion by ice.

Figure 6.34: Interaction curves for pier column for frost attack.


124

6.3.1.5. Supplementary Protective Measures

Some additional measures can be incorporated to protect the pier columns and to enhance their

service life by using a multi-layer protection system. As mentioned previously, protective islands

were created around the piers to diminish the effects of abrasion by ice. At the same time, these

islands can help to reduce the risk of vessel impact against the piers, due to the difficulty of

navigating near these elements.

If the construction of these protective islands results to be inconvenient from the point of view of

cost or severe hydraulic disruption of the flow, another possibility could be to install some steel-

plate lining around the columns covering the zone that may be affected by ice abrasion and

vessel impacts. However, if this approach is adopted, it is recommended that the use of the steel

lining be extended along the entire length of the pier column. As a matter of fact, it would be

useful to leave the steel caisson driven up to the foundation stratum, and then fill it with reinforced

concrete. This can not only serve to protect the columns, but also significantly improves the

columns resistance besides enhancing its durability.

Rock protection of the river bed (Rip-rap) near the location of the bridge will also help to ensure

satisfactory river flow, thereby avoiding any hydraulically-related problems that may affect the

performance of the foundation units.

MAINTENANCE STRATEGIES

125

7. MAINTENANCE STRATEGIES

7.1. General Principles for Maintenance Strategies

The service life commences as soon as the bridge construction is completed. At this time, the

external actions start to affect the bridge structure, and the construction materials begin to suffer

progressive degradation. These external actions involve mechanical loads and environmental

actions.

The structural design for durability must consider these kinds of external actions that interact with

the structure and produce an ongoing deterioration in each one of the structural elements. The

results from a durable structural design approach are related to a defined service life for the

bridge. The maintenance procedures must be addressed to ensure the attainment of the design

service life for each bridge element, from the foundation to the superstructure.

The maintenance strategy must consider different kinds of maintenance procedures to be

performed on the bridge during its service life, including its maintainability, and the needed

preventive and corrective maintenance.

7.2. Design for Maintainability

This stage of maintenance strategy concerns all the aspects considered during the design

process to facilitate all maintenance operations to be performed in a practical and efficient

manner. These aspects involve, for example, the definition of a simple geometry of the structure

that provides easy access for inspection of all structural elements, access to hidden places of the

structure, such as the space between the diaphragms and the front wall of the abutments; and

the installation of monitoring mechanisms 29 that could provide valuable information on the future

behaviour and performance of the structure.

7.3. Preventive Maintenance

This stage of maintenance strategy concerns all of the actions that have to be performed to

preserve the original conditions of the bridge when it was opened to traffic. The preventive

maintenance procedures must consider cleaning activities for the different elements of the

structure, specially the most affected ones because of regular use. These cleaning actions must

involve:

- Cleaning of the draining system to ensure its proper operation, and to avoid the generation of

degradation sources (microclimates) caused by its malfunctioning.


126

- Cleaning of the expansion joints at the abutments to clear the accumulation of debris, dust

and water carrying different aggressive agents.

- Removing the accumulated debris, dust and water carrying aggressive substances

underneath the expansion joints, especially at the abutments seats.

- Cleaning the bottom flanges and the webs of the girders where several deleterious

substances may accumulate, especially at the extremities supported on the abutment seats.

- Cleaning and maintaining the proper conditions of the fixed and mobile bearings to ensure

their proper performance.

- Performing a thorough washing of the entire deck, especially in the splash zones at the

barriers, at the expansion joints, and any other critical zones conductive to accumulation of

salts and any other aggressive agents.

These procedures must be planned to be performed periodically. For this reason, it is necessary

to configure a guideline for seasonal cleaning procedures, especially in the spring, when the use

of de-icing salts is over, with the purpose of eliminating the maximum amount of chlorides

accumulated in the structure, and reducing the aggressiveness of the microclimates acting on the

structural elements, especially during the most severe conditions of temperature and relative

humidity that can be found during the summer season.

7.3.1. Bridge Inspection

As a part of the maintenance strategies, it is important to establish a plan for regular inspections

on the structure during preventive maintenance. The purpose of these inspections is to monitor

the state of the construction materials and the various bridge components over time.

These inspections must be performed by qualified personnel with the necessary training to

develop adequate criteria when making decisions after analysis of the information collected from

the inspections.

As for the cleaning procedures, these inspections must be performed periodically. However, the

degree of detail of each inspection may vary. For these reasons, different kinds of inspections

need to be considered, such as routine inspections, detailed inspections, and special inspections.

7.3.1.1. Routine Inspection

Routine inspections need to be performed more frequently, say about once a month. The

purpose of this kind of inspections is to determine the overall health of the bridge, and to identify

the obvious flaws developing in the various structural elements, which could lead to future


127

deficiencies in the quality of the materials and performance of the components of the bridge. This

inspection must to cover the structural elements of the superstructure, such as the concrete deck

slab, expansion joints, barriers, cornices, girders, diaphragms and the approach slabs.

The inspection of the different elements of the substructure must involve the abutments,

pedestals, bearings, front-walls, wing-walls, piers, piercaps, dowels, and shear-keys. All visible

and accessible parts must be inspected, including the visible parts of the footings and caissons,

the embankments, and the approaches to the bridge.

Under-water inspections are required to assess the condition of the submerged zones of the

foundations. These inspections cannot be as frequent as the routine inspection; however, they

are useful in establishing any deterioration in the foundation units and scour in the river bed which

may affect the stability of the foundations and the structure.

7.3.1.2. Detailed Inspection

This inspection is undertaken based on the findings from the routine inspection. Depending on

the importance and degree of precision of the required detailed inspection, two main groups are

established: general inspection and major inspection.

7.3.1.2.1. General inspection

The general inspection must be performed annually. Their purpose is to verify the detailed

condition of each element of the bridge, identifying all possible flaws. The extent of the flaws

must be determined by on-field measurements, using the necessary equipment and standard

instruments. All of the flaws found on the different structural elements must be recorded

pictorially, and on suitable forms and sketches prepared before the inspection. Immediately after

this process, a written report on the conditions and performance of the materials and the

structural elements of the bridge must be prepared.

7.3.1.2.2. Major Inspection

The major inspection may be required based on the information acquired from the general

inspection. These inspections demand a detailed examination of any of the previously identified

affected, or deteriorated structural elements. These kinds of inspections may be necessary after a

period of time of about 5 years or less, depending of the degree of deterioration found on

previous inspections.

7.3.1.3. Special Inspection

After a major event like an earthquake, a flood, a hurricane, a tornado, an important overload, or

a major accident, such as an explosion, a complete evaluation of the bridge performance must be


128

carried out. This must include testing of materials and a detailed analysis of the entire bridge

structure. These actions must be specially focused on the most critical elements that constitute

the basic structural system of the bridge, including girders, piercaps, piers, abutments and

foundations.

Figure 6.1: Framework for decision-making in bridge management.26


129

Immediately after the analysis is completed, a detailed report must be prepared, showing the

conclusions on all possible distress and the loss of load bearing capacity of the structure. The

report must also include all of the recommendations that must be implemented in terms of repairs

and strengthening, to restore the structural capacity of the bridge, and to provide additional

service life.

7.4. Corrective Maintenance

Any needed corrective maintenance must be defined as a result of the observations from the

different inspections performed on the structure. In this stage of maintenance strategies, it is

essential to define a framework based on the results of the evaluation of the quality of the

different construction materials and the performance of the various structural elements. A

tentative framework for decision making is shown in Figure 6.1.

Corrective maintenance involves repair, rehabilitation, and strengthening. The rehabilitation

process engages all necessary repairs that must be performed with the goal of restoring the

service levels, safety and serviceability of the bridge as closely as possible to the original

conditions.

Strengthening also involves the improvement of the load bearing capacity of the bridge at a

certain time, by increasing the strength and the stiffness of the relevant components of the

structural system, as closely as possible to the original capacity of the bridge.

The corrective maintenance procedure will normally be directed to different kinds of elements in

the structure. These elements are:

- Short-life elements, including pavement, waterproofing membranes and coatings, expansion

joints, elastomeric joint seals, membranes and coatings for the concrete elements exposed to

the most aggressive conditions, including barriers, cornices, drainage meshes, and pipes.

- Medium-life elements, including concrete deck slab, girders, diaphragms, bearings,

embankments, embankment protections, and protective islands.

- Long-life elements, including piercaps, piers, abutments, wing-walls, retaining walls,

abutments, and foundation units.

SUMMARY AND RECOMMENDATIONS

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8. SUMMARY AND RECOMMENDATIONS

8.1. Bridge Design for Durability

To design a bridge structure for durability, it is essential that all the possible loads that could

affect the structure considered in the analysis, and general performance studies performed as an

initial stage of the design process. Consequently, it will be necessary to select construction

materials with sufficient resistance against all anticipated loads and other actions.

One important step during the design process is the definition of the mechanisms of deterioration

developed as the action of each microclimate that can be generated in different parts of the

structure due to the interaction of the macroclimates and the characteristics of the different

elements of the structure.

The various microclimates acting on the various structural elements can generate different

mechanisms of deterioration that can affect the bridge structure. These include freezing and

thawing cycles, carbonation, chloride ingress, chemical attacks (sulphate attack and acid attack),

alkali-aggregate reactions, biological attacks and steel corrosion. Normally, the mechanisms that

affect concrete act as the initiation phase of the deterioration of a reinforced concrete element.

The mechanism of corrosion of steel reinforcement takes place more rapidly, and consequently it

can be considered to be the propagation phase of deterioration.

The rates of these mechanisms of deterioration depend on how the ingress of moisture, air and

aggressive substances occurs into the concrete by different mechanism of transport, such as

capillary suction, diffusion and permeability. Wetting and drying cycles, and cracking of concrete

produced before or after hardening, can accelerate the rate at which these substances can

penetrate the concrete. There exist some mathematical models to describe the rate of the

mechanism of deterioration with time. Some models are more accurate than others and they

have been developed through extensive experimental and field experiences, and theoretical

studies and deductions from the different physical and chemical laws. Finally, these models allow

the prediction of the amount of deterioration with time, and hence these models have to be

evaluated while determining the concrete cover thickness. Additionally, these models should be

integrated in the different formulations of the traditional structural design principles and methods.

According to the information available after the analysis using these models, one important

objective of the structural design for durability is to provide the structure with appropriates

concrete cover thicknesses for the different elements, by ensuring a good quality of the concrete

(low permeability, high resistance to chemical attacks), a sufficient concrete cover thickness

(while also evaluating the risk of cracking depending on the thickness of this cover), and


131

controlling the executions process of concrete placing, compaction, vibration and curing during

construction; these must be implemented using a very high quality control.

The geometrical forms and shapes of the structure may affect the rate of deterioration, because

these forms can make some elements more sensitive to accumulation of aggressive agents and

increased deterioration. In these cases, different kinds of imperfections and low-quality levels

during construction can be generated, resulting in increased ingress of aggressive substances at

early stages of the service life of the structure. Additionally, the cost of construction, execution

and maintenance procedures could increase significantly.

It is necessary to acquire all necessary information and parameters to work with the different

models of deterioration. The precision of these parameters will determine the accuracy of the

estimation of the service life of the bridge.

According to the evaluation of the required design service life and the evolution of different types

of degradation in the structure, the design process could be oriented towards two different

philosophies: to avoid the degradation processes on the structure by changing the environment,

selecting non-reactive materials for concrete and steel, and by inhibiting the reactions by using

protective measures, such as cathodic protection for example. However, it is very difficult to

ensure that all protection measures will perform adequately during the service life. For this

reason, the second type of philosophy is based on considering the existence of the degradation

of the materials and therefore, selecting optimal material compositions and protective measures

to allow the structure to resist the degradation processes that take place during its service life.

For this second approach, the integration between the degradation models and the traditional

design techniques is necessary.

In conclusion, a design procedure for durability must consider: the identification of the

microclimates and mechanisms of deterioration, the definition of the geometry of the structure,

the determination of the composition of concrete, the concrete cover thickness, the type of

reinforcement to use, the definition of protective measures for concrete surfaces and steel bars,

the necessity of the limitation of crack development according to the concrete cover thickness of

the bridge elements, and the determination of the maintenance strategies during the service life

of the structure.

8.2. Holistic design approach

Civil engineers must follow a holistic durability design process to be able to design and build

sustainable and durable bridges. However, the holistic principle involves the interaction of many

considerations that can contribute to the creation of not only durable structures, but also

sustainable practices and procedures of design and construction, seeking a balance between the


132

economic aspects and the environmental impact. Sustainable development is not an option

anymore; it is a must to find a balance between the economical development and the protection

of the environment.

It is time now to define strength through durability rather than durability through strength in the

present and future bridge design practice. The durability of materials cannot be assured only by

focusing on the achievement of a specific strength. The extreme damage and early deterioration

caused by the present practice is manifested in many infrastructure assets around the world, just

because the environmental effects and the deterioration processes that take place on the

structures due to the environment were not considered in their design, construction and

maintenance.

It is important to consider the durability aspects as essential steps during the design process, with

the purpose of incorporating concrete with low permeability, low heat of hydration, low w/c ratio,

enough entrained air, and non-reactive and strong aggregates. This concentration on the

production of durable materials can be guided towards the production of the required strength

concrete for a specific bridge project. The design of the concrete mixtures involves the choice of

proper materials and other ingredients used for the preparing the concrete mixture, including

aggregates, water, type of cement, air, pozzolanic materials and other admixtures. Since the

design process should be holistic, adequate protections and the type of the reinforcement steel

should be considered as well, to be able to build durable reinforced concrete structures.

The holistic design of structures for durability must involve an integrated design procedure,

involving the resistance of materials, their degradation with time, principles of sustainability, life

cycle costing, and environmental protection. If these concepts are implemented in the

engineering and construction practices around the world, there will be a reduction of the amount

of waste of materials, and hence a lower consumption of the natural resources of the planet;

there will be a more effective and efficient use of the construction materials, which can be

reflected in benefits for the present society, and also for the future generations.

8.3. Basic Conclusions and Recommendations

The following basic conclusions can be drawn from this research work:

- A complete structural design for durability is essential to ensure the attainment of the required

service life.

- A holistic design approach is necessary to define an adequate durable design.

- Different interdisciplinary studies need to be coordinated within the holistic approach.


133

- The feasibility of a bridge project must be based on a life-cycle performance and cost

analysis, converting all future operation and maintenance costs into a present value.

- A thorough understanding of the macroclimatic and microclimatic conditions that surround the

bridge is necessary to simulate the decline of the bearing capacity of the bridge, and to

formulate the necessary supplementary protective measures.

- The Integration between structural design and materials engineering is basic to define the

time-dependant performance of the bridge structure.

- The structural design must be accompanied by a maintenance strategy for the owner or the

operator to ensure satisfactory performance of the bridge throughout its service life and

beyond.

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REFERENCES

1. Canadian Highway Bridge Design Code. CAN/CSA-S6-06. National Standard of Canada. Canadian Standards Association (CSA). Standards Council of Canada. November 2006.

2. Wang, L.; Gong, C. Piers and Columns. Bridge Engineering Handbook. Edited by :Chen, Wai-Fah; Duan, Lian. CRC Press LLC, 2000.

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4. Lindsell, P. Substructures. The Manual of Bridge Engineering. Edited by: Ryall, M. J.; Parke, G.A.R.; Hearding, J.E. Thomas Thelford Publishing. The Institution of Civil Engineers. 2000.

5. Sarja, A; Vesikari, E. Durability Design of Concrete Structures. Report of RILEM Technical Committee 130-CSL. RILEM (The International Union of Testing and Research Laboratories for Materials and Structures). Technical Research Centre of Finland. VTT Building Technology. Espon, Finland. E & FN SPON. 1996.

6. Commetary on CAN/CSA-S6-06, Canadian Highway Bridge Design Code. S6.1-06. CSA Special Publication. Canadian Standards Association (CSA). November 2006.

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9. Mehta, P. K.; Burrows, R. W. Building Durable Structures in the 21st Century. The Indian Concrete Journal. July 2001.

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13. Hong, T.-H.; Chung, S.-H.; Han, S.-W.; Lee, S.-Y. Service Life Estimation of Concrete Bridge Decks. KSCE Journal of Civil Engineering. Construction Management Vol.10, No 4, July 2006, pp 233-241.

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18. CSA A23.1-04/A23.2-04. Béton : Constituants et execution des travaux/Méthodes d’essai et pratiques normalisées pour le béton. Association Canadienne de Normalisation. Septembre 2005.

19. Rostam, S. Design for Durability: The Great Belt Link. Concrete Technology: New Trends, Industrial Applications. COWIconsult, Copenhagen, Denmark. © RILEM. E&FN Spon. London, 1995.

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APPENDIX 1 – DURABILITY CALCULATIONS

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The effects of the various mechanisms of deterioration for each bridge element are presented as

follows:


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138


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APPENDIX 2 – GIRDER DESIGN CALCULATIONS

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APPENDIX 2 – GIRDER DESIGN CALCULATIONS

The different verifications of the various limit states considered for the girder design for durability

are presented for time intervals of five years.

APPENDIX 3 – MASS LOSS OF STEEL CAUSED BY CORROSION

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The following Figures summarize the reinforcement mass loss due to chloride-induced corrosion

over a period of 100-year service life, for concrete cover thicknesses ranging from 25 to 70mm in

thickness, chloride concentrations at the concrete surface ranging from 1% to 6 %, a critical

chloride threshold of 0.4%, and a chloride diffusion coefficient of 1.5x10-12m2/s.17


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