Prevention Failure in Soft Ground

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PREVENTION OF FAILURES RELATED TO GEOTECHNICAL

WORKS ON SOFT GROUND

S.S. Gue1 & Y.C. Tan2

ABSTRACT

The success of geotechnical works on soft ground relies on important factors such as proper planning,

analysis, design, construction control and supervision. However, this is usually easier said than done andtherefore there are still repeated failures of geotechnical works such as embankment, foundation and

excavation. Most of the failures are quite similar in nature that they are caused by failing to comply withone or a combination of the above factors. This paper presents case histories of geotechnical failures

investigated by the Authors. The causes of failures, remedial works proposed and lessons learned arediscussed. Finally, some simple guidelines to prevent failures related to geotechnical works on softground are presented.

Keywords: Failure; Soft Ground; Embankment; Excavation; Foundation; Bridge

1. INTRODUCTION

The success of geotechnical works on soft ground relies on important factors of proper planning, analysis,design, construction control and supervision. However, most of the geotechnical failures investigated by

the Authors are usually quite similar in nature that they are caused by failing to comply with one or acombination of the factors stated above. This paper presents the statistic of the geotechnical failures

investigated by the Authors over the recent four years. Case histories of geotechnical failures of embankments, foundations and excavations are also presented together with the causes of failures,

remedial works proposed and lessons learned. Finally, some simple guidelines to prevent failures are alsodiscussed.

2. CATEGORY OF GEOTECHNICAL FAILURES

Failures of projects on soft ground in this paper can be broadly classified into two broad categories. Thefirst category includes those of total or partial collapse of embankments, excavations, foundations, etc.This category often needs reconstruction and/or strengthening measures. The second category of failuresis those due to lateral and vertical movements resulting to severe distortion to completed or adjacent

structures causing loss of serviceability. The affected structures usually need expensive repairs or strengthening works.

The Authors have reviewed 55 cases of failures investigated over the recent four years. The results of theinvestigations are shown in Table 1 which indicates nearly 50% of the failures are largely due toinadequacy in design. The inadequacy is generally the result of lack of understanding and appreciation of

the subsoil and geotechical issues. Hence inadequate assessments, analyses and checks on various modesof failures are the main causes. Failures due to construction either of workmanship, materials and/or lack

1 Managing Director, Gue & Partners Sdn Bhd, Kuala Lumpur, Malaysia2 Director, Gue & Partners Sdn Bhd, Kuala Lumpur, Malaysia



of supervision account for only 15%. The remaining 40% of the failures are attributed to combination of both design and construction.

Table 1 : Cases of Failures due to Design and Construction

Category Design only Construction only Both Design and Construction

Number of Cases 25 8 22Percentage (%) 45% 15% 40%

From the 55 cases of failures investigated, two third of them are due to differential settlements causingdistortion to completed and/or adjacent structures as presented in Table 2. The results also reveal that allthese failures are avoidable if extra care and input from engineers having the relevant experience in

geotechnical engineering were consulted.

Table 2 : Mode of Failures

Mode of Failures Complete or Partial Failure Damage due to Differential Settlement

Number of Cases 18 37

Percentage (%) 33% 67%

2. EMBANKMENT FAILURES

Two case histories of embankment failures investigated by the Authors are presented with causes of failures and lessons learned. The failures of two embankments (namely Embankment A and EmbankmentB) occurred during the construction and are situated at the same expressway but at different locations.

2.1 Failures of Embankment A

Embankment A was initially constructed using vacuum preloading method with prefabricated verticaldrains. Figure 1 shows the cross-section of the proposed embankment. After the 1st failure, the remedial

works of stone columns were proposed and constructed. The embankment with stone columns failedwhen the embankment reached 3.2m of the planned 5.5m fill height. Figure 2 shows the embankmentafter 2

ndfailure. .

The embankment is sitting on very softsilty Clay of 4.5m thick and follows by a

layer of soft sandy Clay to a depth of 12m.Beneath these very soft to soft cohesivesoils is a layer of loose clayey Sandfollows by layers of medium to stiff silty

Clay. Figure 3 shows the undrained shear strength (su) profile of the subsoil obtainedfrom field vane tests.

The effectiveness of the vacuum preloading method is dependent on manyfactors like the pump capacity, the airtight

seal between the edge of the geomembraneand the subsoil; integrity of the

geomembrane at the ground surface, effectiveness of the vertical drains and etc. This method requiresclose monitoring of the pore water pressures in the subsoil during filling to prevent failure.

Very Loose Clayey SAND

Medium to Sti ff Si l t y CLAY and Clayey SILT

Embankment Fil l(Without Vacuum Preloading)

Embankment Fill (Failed Area)(Vacuum Preloading with Vertical Drains)

Vertical Drains

Liner and Sand Layer

for Vacuum System

Scale (m)

0 5 10

Soft Sandy CLAY

Very Soft Sil ty CLAY

Figure 1 – Cross-section of Embankment A.



In view of this, instruments like piezometers, settlement gauges and vacuum meters have been installed atsite with the intention to monitor the performance of the embankment treated with vertical drains andvacuum preloading. The construction sequence of Embankment A and changes of pore water pressure of

the piezometers in the subsoil at depths 3m, 6m and 8m throughout the construction are shown in Figure 4.Embankment A failed not long after reaching the final fill height of 5.5m. As shown in Figure 4, from

Stage C filling onwards, the pore water pressure measured from piezometers PZ-A2 and PZ-A3 at depthsof 6m and 8m respectively increased beyond the design pore pressure until failure at Day162 after reaching the final fill height. Piezometer PZ-A1 at 3m deep did not show increase in pore water pressureuntil it was out of order after Day 130. In brief, the measurement from piezometers PZ-A2 and PZ-A3 at

Embankment A had indicated that the vacuum suction at these depths were not functioning properly andwas unable to prevent the increase of pore water pressures in the cohesive subsoil with respect to the

embankment loads.

The trend of increase in pore water pressures have been observed for more than one month but no

contingency action was taken by the Contractor and the Consultant, who was also responsible for thedesign, to investigate the causes and to stop the filling until the pore water pressure in the subsoil drops

Sheer Drop and Cracks

Figure 2 – Failure of Embankment A treated with Stone Columns.

Heave Up

16

14

12

10

8

6

4

2

0

e p

m

0 10 20 30Sensitivity, St

Su-Undisturbed from VS-A

Su-Remolded from VS-A

Su-Undisturbed from VS-B

Su-Remolded from VS-B

In-Situ Vane Shear Test

VS-A

VS-B

Su = 10 kPa

Su = 8 kPa

Su = 13 kPa

Su = 17 kPa

Su = 19 kPa

0 10 20 30 40 50 60

Undrained Shear Strength, Su (kPa)

Figure 3 – Undrained Shear Strength Profile.

Fist Crack Observed on Day 162

Excess Pore Water Pressuregenerated at PZ-A3,

U = + ve

Excess Pore Water Pressuregenerated at PZ-A2,

U = + ve

Stage BStage C

Stage D

Stage EStage F

∆

-2

0

2

4

6

8

10

12

e z o m

e t e r

e a

m

Piezometers at Location A

at 3.0m depth

at 6.0m depth

at 8.0m depth

0 50 100 150 200 250 300 350Days0

2

4

6

e

g

t

m

Designed Water Headis 3m at PZ-A1

Designed Water Headis 6m at PZ-A2.

Designed Water Headis 8m at PZ-A3 ∆

Figure 4 – Construction Sequence and

Monitored Pore Water Pressure Changes.



below the allowable design values. Details of back-analyses and methodology of monitoring usingobservational method are presented by Gue et al. (2001) and Tan & Liew (2000) respectively.

From the monitoring results, it is clear that 1st failure of Embankment A could have been avoided if

observational method (Peck, 1969) was employed properly.

After the 1st Failure, stone columns were proposed and constructed by the Contractor using vibro-replacement process as remedial measures to support the re-construction of the new embankment. The

stone columns are of 1m diameter with grid spacing of 2.5m centre-to-centre up to a depth of 20m.Crushed stones of size 15mm to 100mm were used as backfill medium for the stone columns. During re-

construction of the embankment on top of the stone columns, the embankment failed with large cracks (asshown in Figure 2) when the fill height reached 3.2m which is 2.3m lower than the required fill height of 5.5m.

Table 3 : Methods for Estimation of Ultimate Bearing Capacity of Stone Columns

Mode of Failures References

Bulging Greenwood (1970); Vesic (1972); Datye & Nagaraju (1975); Hughes andWithers (1974); Madhav et al. (1979).

General Shear Madhav & Vitkar (1978); Wong (1975); Barksdale and Bachus (1983).

Our review indicates, the design by the Specialist Contractor only used Priebe’s methods (1995) to check on the stability and settlement of the subsoils treated with stone columns. There was no evidence of separate calculations using other methods to check on the bulging and general shear failures of the stone

columns when determining the ultimate bearing capacity; these failure modes are not sufficiently coveredin Priebe’s method. Table 3 lists some of the methodologies available for bulging and general shear

failure check.

From the investigation by the Authors using disturbed strength of the subsoil on the methods listed in

Table 3, the results show that generally bulging failure is not a concern but general shear failure is grosslyinadequate.

Figure 5 – (a) Stresses on Stone Column . (b) Comparison of Different Methods (after Madhav &

Miura, 1994)



Most of the methods listed in Table 3 are reproduced in a graph by Madhav & Miura (1994) together withtheir proposed method as shown in Figure 5. It is observed that there is a large range of possible ultimate bearing capacity when using different methods and this tends to cause confusion to design engineers.

Therefore, it is recommended that when using stone columns in very soft ground (e.g. su < 15kPa) or as

remedial measures for reconstruction of failed embankments, attention shall be given to probable failuredue to general shear. In addition, load tests and close monitoring of the instrumentation should be carriedout to verify the design.

2.1 Failures of Embankment B

Embankment B is located about 2km away from Embankment A. It was initially treated with prefabricated vertical drains. Cracks were observed at the embankment after reaching the surcharge levelwith fill height of 3.9m and immediate action was taken to lower down the embankment height to Finish

Road Level (FRL) which is 1.5m lower. The embankment was observed for 2 months and since no further cracks developed, the Consultant agreed to refill the embankment to surcharge level. Slip failure occurred

during the filling of the surcharge. After the 1st

Failure, the Contractor decided to use stone columns as

remedial measures to strengthen the subsoil so that the embankment can be reconstructed. However, theembankment supported by stone columns failed again after reaching the fill height of 3.9m.

The subsoil at Embankment B area generally consists of organic soil with a thickness of about 4m.Underlying the organic soil is a layer of very soft to soft silty Clay with thickness of about 10m follow bystiff to very stiff clayey Silt. Similar to 2

ndfailure of Embankment A, the stone columns for Embankment

B were also being design using Priebe’s method (1995) only without other separate checks on the bulging

and general shear failures as listed in Table 3. The investigation carried out by the Authors indicate thatthe stone columns bearing capacity against general shear failure is grossly inadequate; resulting to the

failure of the embankment.

2.3 Lessons Learned from the Embankment Failures

The 1st

failure of the Embankment A treated with vacuum preloading method and prefabricated verticaldrains was monitored but no action was taken to review the monitoring results and prompt for preventive

action. The failure could have been prevented if the Contractor or Consultant had reviewed themonitoring results regularly and taken the necessary preventive action.

The failures of Embankment A and Embankment B treated with stone columns were mainly due to

inadequate design. The Authors are of the opinion that when designing stone columns to treat very softground (e.g. su <15kPa) or as remedial measures for an embankment, attention should be given to probablegeneral shear failure instead of over relying on single method. For remedial measure, it is also importantto determine the representative “disturbed” (remoulded and regaining of strength through thixotropy

effects) strength of the subsoil to be used in the analyses. In addition, load test shall be carried out on

stone columns to verify the design assumptions as there are large differences among methods of analysis.In brief, further works are necessary before a reliable unified and comprehensive design method isavailable for stone columns supporting embankment on very soft ground.

When stone columns are used to treat very soft ground, it is recommended that observational method(Peck, 1969) be used with proper instrumentation and closer monitoring to prevent failure if there is a

slight doubt on the design methodology. Many embankments on very soft ground treated with stonecolumns have been successfully constructed with the help of observational method.

Failures of embankment due to design are commonly caused by the following inadequacies :-

(A) Settlement Analysis(B) Stability of Embankment



(A) Settlement Analysis

It is very important to evaluate both the magnitude and rate of settlements of the subsoil

supporting an embankment. This is to ensure the settlement in the long term will not affect theserviceability and safety of the embankment.

In carrying out stability analysis, it isimportant to correctly estimate the

magnitude of settlement duringconstruction so that the correct thickness

of the fill can be incorporated in thedesign to ensure stability. An iterative process is required in the estimation of settlement because the extra fill (higher

pressure) that is required to compensatefor settlement will lead to additionalmagnitude of settlement.

The three main settlements need to be evaluated are :- Initial Settlement

- Primary Consolidation Settlement- Secondary Compression

(B) Stability of Embankment

It is necessary to design the embankment with consideration for different potential failure surfacesnamely circular and non-circular as shown in Figure 6. The thickness, unit weight and strength of the fill need to be properly determined. Minimum design surcharge loading of 10kPa is required

for embankment design to represent traffic and unexpected loading during construction.Generally in practice, the factor of safety (FOS) for temporary stage (construction stage) using

undrained strength analysis should be 1.2 or higher and the long term FOS for effective stressanalysis of embankment is usually 1.4 or higher.

3. FAILURE OF BRIDGE FOUNDATION AND APPROACH EMBANKMENT

One case history of bridge failure investigated by the Authors is presented in this paper with the causes of failure and lessons learned illustrated. Although only one case history is presented, a few other casehistories of bridge failures investigated by the Authors are quite similar in nature and were mostly induced

by bearing capacity

and stability of the

embankment (Gue,1988). Theinvestigations alsoclearly show thatconstruction methodsemployed at the site

also have significantinfluence on the

success of the project.These failures could

be prevented if thedesign consultant and

the contractor havetaken adequate care in

Figure 6 – Circular & Non-Circular Slip Failure Surfaces

Figure7 – Overview of Partially Completed Bridge after Failure



Abutment

Pier II

Abutment

Pier I

Figure 8 – Layout of Piers and Abutments

I

II

geotechnical considerations inthe analysis, design andconstruction.

The bridge failure presented in

this paper is a project of anaccess road with prestressedconcrete beams over a river in

Sarawak and the failure occurredduring construction. The

proposed heights of theapproach embankments on bothsides of the abutments wereabout 5m with side slopes of 1v(vertical) to 1.5h(horizontal).

Figure 7 shows the partiallycompleted bridge after failure

and removal of fill embankment.The layout of the proposed bridge is shown in Figure 8

The approach embankmentswere constructed over 25m thick

of soft coastal and riverinealluvium clay underlain by

dense silty Sand and very stiff silty clay. The soft alluvium

generally has SPT N’ value of zero and average moisturecontent of more than 70% whichis near its Liquid Limit. Figure9 shows the subsoil profile of the site.

The approach embankmentswere supported by 200x200mmsquare reinforced concrete (RC) piles and cast with individual

pilecaps. In addition, 6m lengthwood piles were also added

between the RC piles for further support of the embankment fill.More wood piles were alsoinstalled on the banks of theriver trying to stabilize the

lateral displacement of the soft alluvium. The abutments and piers were supported by 400mm diameter spun piles driven to set in the hard layer of more than 30m deep.

A deep seated slip failure occurred at the approach embankment with a sheer drop at about 25m behindAbutment II. It happened when the fill reached about 3m high. Figure 10 shows the sheer drop after removal of some of the fill behind the abutment. Abutment II has tilted away from the river with a

magnitude of about 550mm at the top of the abutment at the time of the site inspection by the Authors who

were carrying out the geotechnical investigation of the failure. The tilt translates into an angular distortionof 1/6. Due to the excessive angular distortion, the integrity of the spun piles driven to set into the stiffer

Figure 9 – Subsoil Condition after Failure



Opening between

bridge decks

Tilt from

Vertical

Figure 11 – Tilted Abutment and Observed Gap between Bridge Decks

Opening between

bridge decks at

Pier II

Sheer Drop

Pilecaps

Figure 10 – Sheer Drop at about 25m behind the Tilted Abutment

stratum has also been affected as it exceeds the normal structural failure threshold of about 1/75. Due tothe tilt of the Abutment II away from Pier II, a gap of about 300mm wide was observed between the two

bridge decks at the pier. Figure 11 shows the photograph of the tilt at the Abutment II and the gap between two bridge decks at Pier II. The failure also caused Pier II to tilt slightly. Figure 12 shows the

schematic diagram of the possible slip plane relative to the deformed structures.

These observations infer that the slip failure of the Approach Embankment near Abutment II is deepseated and is consistent with the depth of the soft alluvium. The cause of the rotational slip failure was dueto the weak subsoil unable to support the weight of the approach embankment.

The use of the RC piles and wood piles offered little lateral resistance and instead, extended the rotational

slip deeper into the soft subsoil. At the pier, the bridge deck, being simply supported and fixed to theAbutment II via bearing pads, moved along with the displacement of the Abutment.

At the start of the construction, the supervisor of the contractor had observed that their workers could notwalk on the riverbanks without their feet sinking into the soft subsoil to a depth of about a foot. This



observation infers that the upper subsoil had an undrained shear strength of about 10kPa. As a quick preliminary check using simplified bearing capacity equation stated in Section 3.1, the ultimate bearingcapacity was about 50 to 60kPa. The estimated maximum height of fill that could be supported at failureis about 3m which is consistent with the observed failure when the embankment reached 3m high.

Therefore, if the designer and contractor had carried out simple bearing capacity checks, failure could

have been prevented.

The results of the additional subsurface investigation after the failure show that the undrained shear

strength from the vane shear tests range from 18kPa to 51kPa with remoulded strength of 7kPa to 12kPa.The higher su obtained from the S.I. carried out after failure is due to the gain in strength from the imposed

embankment fill over time.

3.1 Lessons Learned from Bridge Failures

The failures were caused by the following factors :-- Inadequacy of geotechnical design for the approach embankments and abutments.

- Lack of understanding of the subsoil condition and awareness on the possible problems/failures that could happen during construction.

- Lack of construction control and site supervision by the Consultant.

As highlighted in Section 2.3 of this paper, embankment stability shall be checked for both possiblecircular and non-circular (wedge) failure surfaces using limit equilibrium method. It is wrong to assume

that as long as the structural design of an abutment has considered both vertical and lateral earth pressures behind the abutment, slip failure would not occur. Figure 12 is a good example of abutment instabilitywith deep seated failure seriously affecting the stability of the abutment. The most critical condition thatof an embankment on soft ground is during filling where the stability of an embankment should beanalysed based on undrained shear strength (su) of the subsoil. Sufficient in-situ field vane shear tests

should be carried out to provide representative moderately conservative undrained shear strength profile of the subsoil for stability analysis.

Figure 12 – Schematic of Slip Failure



A quick preliminary check on the stability of the embankment is possible using simplified bearing

capacity equation below :qallow = (su. Nc / FOS)

where :

qallow = allowable bearing pressure = (γ fill.H + 10) (kN/m2

)γ fill = bulk unit weight of the compacted fill (kN/m

3)

H = allowable height of embankment (m)su = undrained shear strength of the subsoil (kPa) Nc = 5 (suggested by Authors for ease of hand calculation)FOS = Factor of Safety (e.g. minimum of 1.2 for short term using moderately

conservative su) Note : The 10kPa allowance in the qallow is to cater for the minimum vehicle load.

It is also important to check for the loading on the abutment piles from lateral soil pressure imposed by theembankment fill behind an abutment. This is to prevent failure of the pile group supporting the abutment.

Methods that can be used are Tschebotarioff (1973), Stewart, et al. (1994), Springman (1989), DeBeer

and Wallays (1972). For more complicated structures, Finite Element Method (FEM) should also be used.

When constructing bridges on very soft ground, design consultant, consultant’s site engineer(s) andcontractor should have some fundamental geotechnical knowledge which include understanding of thesubsoil condition and awareness on the possible problems or failures that could happen during

construction. A good example is shown in Section 3.0 where the contractor were aware that their workerscould not walk on the very soft riverbank and could have used the simple bearing capacity equation to

check on the allowable height of the fill that the subsoil can support. Quite often, failures were due to badtemporary works that were never considered in the design.

One serious problem that usually occurs for bridge project is the temporary fill placed by the contractor toform a temporary platform to facilitate their piling or other construction works. If not careful, slip failure

in subsoil could occur by the load imposed by the temporary fill. Therefore, it is recommended that thedesign consultant should consider the possible construction method to be used by the contractor and

designed for it or check the stability when the method statement is submitted for approval or record. Thedesign consultant shall also ensure that during construction, the contractor must carry out works accordingto the approved method statement to prevent failure. Finally, proper full-time site supervision by theconsultant’s representatives who have adequate site experience and knowledge are also very important to

prevent failure due to temporary works and ensure permanent works are constructed according to thedrawings and specifications.

Another common problem caused by temporary fill over soft ground is the failure to remove the

temporary fill after construction. The temporary fill would cause the compressible subsoil to settle withtime (consolidation settlement). If temporary fill area has piles, then the piles will be subjected to down

drag (negative skin friction) due to the settling subsoil and reduce the capacity of the piles. If the downdrag is not catered for in the design, the piles will have lower allowable capacity and larger settlementcausing distortion to the structures. Therefore, the design consultant shall ensure the removal of temporary fill after construction by the contractor or to design the piles to accommodate negative skinfriction.

4. EXCAVATION FAILURE

Excavation in soft ground can be carried out either through open cut or using retaining wall systemdepending on the site constraint, depth of excavation, groundwater conditions and type of subsoil. Usually

failures of retaining wall system for excavation in soft ground can be divided into four major modes of failures :-



Figure 13 – Failure of Temporary Sheet Pile Wall

Movement of Sheet Pile

- Inadequacy of Penetration Depth of wall or Support(Resisting Force)

- Base Heave

- Hydraulic Failure- Slip Failure

This paper presents a case history of failure of the temporary sheet pile wall

near Port Klang. The site was a flatmarine deposits with the original groundlevel at grade with the access road. Thesite is on a former residential lot with adetached house and planted with various

fruit trees.

Whereq = Surcharge Load

~ 10kPa (minimum)

Q = q x D (no prop)= q x r (with prop)

W = Total Weigh of Soil

= γ HD (no Prop)

= γ Hr (with Prop)

r = D + s

L = Total Arc Length of Soil Resistance

= πD (no prop)

= πr – 2s (simplified, with prop)

Case 1 : No Prop

FOS =

2)( DQW

D L su

⋅+

⋅⋅

Case 2 : With Prop

FOS =

2)( r QW

r L su

⋅+

⋅⋅

Note :

The required FOS is 1.2 where the verticalshear resistance along the retained ground

shallower than the excavations is ignored.(Kohsaka & Ishizuka, 1995).

Case 1 : No Prop

Case 2 : With Pro

Figure 14 – Base Heave Check based on Equilibrium of Moments



Figure 15 – Stability of Sheet Pile Penetration Depth

The results of the boreholes drilled at site for design indicate that the thickness of the soft marine clayvaries from 18m to 20m follows by 25m to 32m of soft to stiff clay with an intermittent layer of dense tovery dense sand of less than 3m thick. The top layer of marine clay has an undrained shear strength (su) of about 10kPa near the surface and slowly increases with depth.

The project involved the construction of a high rise building with a level of basement carpark. Reinforcedconcrete (RC) piles were driven from the original ground level. After the installation of the reinforcedconcrete piles, excavation was carried out for the construction of the basement and pilecaps. A 12m sheet

pile (FSC III) acting as temporary cofferdam was installed to facilitate the basement and pilecapsexcavation.

The sheet pile wall was stable when the excavation reached the proposed basement level of 2.5 m.However, when the excavation for pilecaps in front of the sheet pile, reached 3.5m to 4m, the sheet pilemoved excessively towards the excavation site and the base of the excavation also heaved up. Theexcessive movement of the soil pushed and moved the installed RC piles. Some of the piles moved

laterally for more than a meter thus damaging the integrity of the piles. Figure 13 shows the condition of the site after the movement of the sheet pile wall.

The Authors carried out analysesagainst base heave failure and alsoadequacy of the penetration depth

using simplified gross-pressuremethod. Figure 14 shows the

method for quick evaluation of baseheave which take into consideration

of embedded length of the wall andcan be modified for varying su. The

method is commonly used in Japan(Kohsaka & Ishizuka, 1995) and theAuthors have used it successfully for many projects in Malaysia. Therequired Factor of Safety usingmoderately conservative strength isnot less than 1.2 as the vertical shear

resistance along the retained groundshallower than the excavation isignored. If the vertical shear resistance along the retained ground

is considered, then the FOS of notless than 1.4 should be adopted. The

back-analyses carried out by theAuthors indicate that FOS against base heave is more than 1.2 andtherefore it is not the cause of thefailure. Hydraulic and slip failures

have also been checked and found to be acceptable.

The Authors also carried out back-analyses for the adequacy of the sheet pile penetration depth using thesimplified gross pressure method. The results are presented in Figure 15 which indicate that the criticaldepth of excavation using 12m deep sheet pile in the above site ranges from 2.75m to 3.5m depending

whether it is propped and with 10kPa surcharge. The results confirm that the 12m penetration depth of the

sheet pile is not adequate to support an excavation depth exceeding 3.5m with props at this site.

0.0 0.5 1.0 1.5 2.0 2.5 3.0 3.5 4.0 4.5 5.0 5.5 6.0

Depth of Excavation (m)

0.25

0.50

0.75

1.00

1.25

1.50

1.75

2.00

F a c t o r o f S a f e t y ( F O S )

Unpropped (10kPa Surcharge)

Unpropped (No Surcharge)

Propped (10kPa Surcharge)

Propped (No Surcharge)

Note: Prop assumed at 2.5m below retained level

Critical depth ≈ 2.75m to 3.5m



4.1 Lessons Learned from Excavation Failure

Many case histories investigated by the Authors indicate that excavation failures are usually caused by thefollowing factors :-

- Inadequacy of geotechnical design for various modes of failures listed in Section 4 above.- Lack of construction control and site supervision by Consultant such as over-excavation

(e.g. excavate deeper than designed depth) and uncontrolled surcharging at retained soil

(e.g. stacking of excavated materials or other materials behind the wall at the retainedside).

When designing retaining structures for excavation, it is necessary to check the following ultimate limitstates of the wall :-

1) Overall Stability : to provide sufficient embedment depth to prevent overturning of the wall and overall slope stability.

2) Basal Failure : the wall penetration depth must be sufficient to prevent basal

failure in front of the wall after excavation to the proposedformation level.

3) Hydraulic Failure : the penetration of the wall must be sufficient to avoid piping or ‘blow out’ in front of the wall after excavation to the proposedformation level.

Gue & Tan (1998) summarises many manual methods to design walls to prevent all modes of failures

listed above. Normally manual methods are slightly on the conservative side. In order to further optimisethe design of retaining wall system for excavation, it is recommended to use Finite Element Method

(FEM) with proper input of representative soil parameters, groundwater conditions, and also theconstruction sequence.

As excavation is a complicated soil-structure interaction problem, especially for deep excavation withmultiple levels of support, FEM method if used properly is the most suitable. The main advantage of

FEM method is its ability to predict wall and ground deformations and allows sensitivity analyses to becarried out for value engineering. The recent Rankine Lecture by Potts (2003) provides good account of

the numerical method with case histories. The FEM method is particularly useful for excavations in theurban area with many nearby buildings and services.

Another important factor to consider when selecting and designing retaining wall system for excavation

adjacent to properties sensitive to ground settlement is the lowering of groundwater at the retained soil dueto excavation. Every meter of drop of groundwater level is equivalent to about 10kPa of surcharging onthe subsoil below the original groundwater level, hence causing it to consolidate and settle. Rechargewells should be considered and used if settlement of adjacent ground due to lowering of groundwater level

likely to cause distortion and damage to adjacent properties.

5. CONCLUSIONS

The case histories of failures related to geotechnical works on soft ground have clearly indicated that thesefailures are generally quite similar in nature and are avoidable. These failures are usually man-made andcaused by failing to comply with one or a combination of factors which include planning, analysis, design,

construction control and supervision. Observational method should be used to compare design predictionwith field performance to ensure safety.

From the 55 cases of failures investigated by the Authors over the recent four years, 50% of them are dueto inadequacy in design. In order to prevent similar failures, it is important for design consultant,



consultant’s site representatives and contractor to have some fundamental geotechnical knowledge so thatany inconsistencies at site can be spotted and precautionary actions taken before failure occurs. Proper full-time site supervision by the consultant’s representatives with adequate experiences, knowledge is amust. It is also the obligation of the consultant to properly brief the supervising team on the design and

construction requirements.

REFERENCES

Aboshi, H., Ichimoto, E., Enoki, M. and Haraka, K. (1979). “The Compozer – A Method to ImproveCharacteristics of Soft Clays by Inclusion of Large Diameter Sand Columns”. Proc. Intl. Conf. OnSoil Reinforcement : Reinforced Earth and other Techniques, Paris, Vol. 1, p.211-216.

Barksdale, R.D. and Bachus, R.C. (1983). “Design and Construction of Stone Columns”. Report No.FHWA/RD-83/026, National Technical Information Service, Springfield, Virginia, USA.

Bergado, D.T. and Lam, F.L. (1987). “Full Scale Load Test of Granular Piles with Different Densitites andDifferent Proportions of Gravel and Sand in Soft Bangkok Clay”. Soils and Foundations, Vol. 27,

No. 1, p. 86-93.

Bergado, D.T., Alfaro, M.C. and Chai, J.C. (1991). “The Granular Pile : Its Present State and FutureProspects for Improvement of Soft Bangkok Clay”. Geotechnical Engineering, 22 , p. 143-175.

Datye, K.R. and Nagaraju, S.S. (1981). “Design Approach and Field Control for Stone Columns”. Proc.10th Intl. Conf. On Soil Mech. Found. Eng’g., Stockholm.

DeBeer, E.E. and Wallays, M (1972). “Forces Induced in Piles by Unsymmetrical Surcharges on the SoilAround the Piles”. Proc. 5th ECSMFE. Madrid, Vol. 1, p. 325-332

Gue, S. S. & Tan, Y. C. (2003), “Prevention of Failure of Bridge Foundation and Approach Embankmenton Soft Ground”. Lead Speaker, 2nd International Conference on Soft Soil Engineering andTechnology, Putrajaya, Selangor, July, 2003.

Gue, S. S. & Tan, Y. C. (2001), “Geotechnical Solutions for High Speed Track Embankment – A Brief Overview”, Technical Seminar Talk, PWI Annual Convention, Permanent Way Institution, Beserah,Kuantan, Pahang, 28th – 29th September, 2001.

Gue, S. S., Tan, Y. C. & Liew, S. S. & Ng, H. B. (2001), “Failure of an Embankment Treated withVacuum Preloading Method”, Proceedings of the XVth International Conference on Soil Mechanicsand Geotechnical Engineering, Istanbul, Turkey, Vol. 3, p. 1749 – 1751.

Gue, S.S. and Tan, Y.C (2000). “Subsurface Investigation and Interpretation of Test Results for Foundation Design in Soft Clay”. Seminar on Ground Improvement – Soft Clay (SOGISC-2000),23-24 Aug. 2000, UTM, Kuala Lumpur.

Gue, S.S. (1999). “Selection of Geotechnical Parameters for Design”. Short Course on Soil Investigationand Design for Slope (SCOFG99), 11 & 12th May, 1999, Kuala Lumpur.

Gue,S.S. and Tan, Y.C. (1998) “Design and Construction Considerations for Deep Excavation”. Lecture.IEM Northern Branch, 26 September, 1998. Penang.

Gue, S.S. (1988). “An Investigation into Geotechnical Failures of a Bridge Project”. IEM/RRIM JointEngineering Symposium, 20th Aug 1988, Johor Bahru, Malaysia.

Greenwood, D.A. (1970). “Mechanical Improvement of Soils Below Ground Surface”. Proc. GroundEng’g. Conference, Institute of Civil Eng’g., p. 9-20.

Hughes, J.M.O. and Withers, N.J. (1974). “Reinforcing soft cohesive soil with stone columns”. GroundEngineering. Vol. 7, No. 3, p. 42-49.

Hughes, J.M.O., Withers, N.J. and Greenwood, D.A. (1975). “A Field Trial of Reinforcing Effects of

Stone Columns in Soil”. Geotechnique, Vol. 25, No. 1, p. 31-44.



Kohsaka, N. & Ishizuka, K. (1995). “Stability of Bottom during Deep Excavations – A Survey onJapanese Codes”. Proc. of the Intl. Symposium on Underground Construction in Soft Ground,(additional paper),New Delhi, India, p. 323-328.

Madhav, M.R. and Miura, N. (1994). “Stone Columns”, Proc. 13th Intl. Conf. Soil Mech. Found. Eng’g., New Delhi. India. 4 , pp. 163-164.

Madhav, M.R., Iyengar, N.G.R., Vitkar, R.P. and Nandia, A. (1979). “Increased Bearing Capacity andReduced Settlements due to Inclusions in Soil:. Proc. Intl. Conf. On Soil Reinforcement : ReinforcedEarth and Other Techniques, Vol. 2, p.239-333.

Madhav, M.R. and Vitkar, R.P. (1978). “Strip Footing on Weak Clay Stabilized with Granular Trench or Pile”. Canadian Geotechnical Journal, Vol. 15, No. 4, p.605-609.

Peck, R.B. (1969). “Advantages and limitations of the observational method in applied soil mechanics.”Geotechnique. Vol. 19, No.2, pp.171-187.

Potts, D.M. (2003). “Numerical Analysis : A Virtual Dream or Practical Reality ?”. Geotechnique. Vol53, No 6, p.535-573.

Priebe, H.J. (1995). “The Design of Vibro Replacement”. Ground Engineering, December , pp. 31-37.

Springman, S.M. (1989). “Lateral Loading on Piles due to Simulated Embankment Construction.” PhDthesis. Cambridge University.

Stewart, D.P., Jewell, R.J. and Randolph, M.F. (1994). “Design of Piled Bridge Abutments on Soft Clayfor Loading from Lateral Soil Movements”. Geotechnique, Vol. 44, No. 2, p. 277-296.

Tan, Y. C. and Liew, S. S. (2000). “Observational Method to Prevent Failure of Embankment Treatedwith Vaccum Preloading”. Seminar on Failures Related to Geotechnical Works, IEM, Kuala Lumpur

Tan, Y.C. and Gue, S.S. (2000). “Embankment over Soft Clay – Design and Construction Control”.Seminar on Geotechnical Engineering, 22 & 23 September 2000, Penang.

Tschebotarioff, G.P. (1973). “Foundations, Retaining and Earth Structures”. McGraw-Hill. New York.

Wong, H.Y. (1975). “Vibroflotation – Its Effect on Weak Cohesive Soils”. Civil Engineering (London), No. 82, p. 44-67.

Vesic, A.S. (1972). “Expansion of Cavities in Infinite Soil Mass.” Journal of Soil Mech. And Found.Eng’g Div., ASCE, Vol. 98, No. SM3, p. 265-290.

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Prevention Failure in Soft Ground