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Submitted to:
Tetra Tech WEI Inc.Winnipeg, Manitoba
Submitted by:
Amec Foster Wheeler Environment & Infrastructure
A Division of Amec Foster Wheeler Americas Limited
440 Dovercourt Drive
Winnipeg, Manitoba
R3Y 1N4
Canada
File number: WX17918
30 September 2016
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page i
TABLE OF CONTENTSPAGE
4.1 Stratigraphy .........................................................................................................6
4.2 Auger Refusal ......................................................................................................9
4.3 Groundwater and Sloughing Conditions...............................................................9
6.1 Spatial Variability in Soil and Groundwater Conditions.......................................11
6.2 Foundation Strength and Deformability ..............................................................12
6.3 Artesian Groundwater ........................................................................................12
8.1 General Evaluation ............................................................................................14
8.2 Driven Steel Piles...............................................................................................14
8.2.1 General Discussion...............................................................................14
8.2.2 Geotechnical Bearing Resistance .........................................................15
8.2.2.1 Strength Limit State ..............................................................15
8.2.2.2 Service Limit State................................................................17
8.2.3 Tensile (Uplift) Resistance ....................................................................18
8.2.3.1 Strength Limit State ..............................................................18
8.2.3.2 Service Limit State................................................................19
8.2.4 Lateral Pile Resistance .........................................................................19
8.2.4.1 Strength Limit State ..............................................................19
8.2.4.2 Service Limit State................................................................21
8.2.4.3 Lateral Pile Analysis Results for Select Piles Sizes...............21
8.2.4.4 Inclined (Battered) Piles........................................................23
8.2.5 Minimum Embedment Depth ................................................................23
8.3 Pile Group Effects ..............................................................................................24
8.4 Artesian Groundwater Impact on Semi-Integral Abutments................................24
10.1.1 Earth Pressure Coefficients ..................................................................26
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page ii
10.1.2 Calculation of Earth Pressure Distribution and Surcharge Loads..........27
10.1.2.1 Moderate to Well Compacted Backfill Case ..........................27
10.1.2.2 Surcharge Loads ..................................................................27
10.1.3 Load Factors.........................................................................................28
11.1 Frost Penetration Depth.....................................................................................28
11.2 Pile Foundations ................................................................................................28
11.2.1 Frost Heave..........................................................................................28
11.2.2 Adfreeze Stresses ................................................................................29
12.1 Design Philosophy .............................................................................................29
12.1.1 Design Criteria......................................................................................30
12.1.2 Methodology and Model Geometry .......................................................30
12.1.3 Soil Stratigraphy and Soil parameters...................................................30
12.1.4 Piezometric Conditions and Creek Levels.............................................31
12.2 Slope Stability Results .......................................................................................32
12.2.1 Headslope Stability ...............................................................................32
12.2.2 Grouted Rip Rap Slopes.......................................................................33
12.2.3 Cofferdam and Creek Dewatering (Short Term Stability) ......................34
12.3 Slope Stability Conclusions and Recommendations for Detailed Design............35
14.1 Pavement Design Methodology .........................................................................36
14.2 Design Vehicle and Traffic .................................................................................36
14.3 Subgrade Resilient Modulus ..............................................................................37
14.4 Granular Base Course and Subbase Course Materials......................................37
14.5 Subgrade Preparation........................................................................................37
14.6 Asphalt Concrete Pavement (ACP) Alternative ..................................................39
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page iii
LIST OF TABLES
Table 4-1: Summary of Existing Pavement and Gravel Thickness ..............................................7Table 4-2: Summary of Atterberg Limit and Particle Size Analysis Results .................................8Table 4-3: Summary of Unconfined Compressive Strength Tests ...............................................8Table 4-4: Observed Slough and Groundwater Conditions During Drilling ..................................9Table 5-1: Recommended Geotechnical Resistance Factors for Driven Steel Piles ..................11Table 7-1: Factor of Safety Against Basal Heave versus Sturgeon Creek Water Level .............13Table 8-1: Recommended Parameters for Tensile Resistance of Driven Piles..........................19Table 8-2: LPile Input Parameters for Lateral Pile Analysis .......................................................20Table 8-3: Top of Pile and Grade Elevation LPile Configurations ..............................................22Table 10-1: Earth Pressure Coefficients and Soil Unit Weights .................................................26Table 12-1: Summary of Slope Stability Material Parameters....................................................31Table 12-2: Summary of Slope Stability Results for Overburden Phreatic Surface
Elevation 233.8 m, and Artesian Total Head of 234.4 m in the underlyingTill ......................................................................................................................32
Table 14-1: Summary of Granular Pavement Structure Materials..............................................37Table 14-2: ACP Pavement Alternative for 90% Reliability........................................................40
LIST OF FIGURES
Figure 1: Test Hole Location PlanFigure 2: Summary of Slotted Standpipe and Vibrating Wire Piezometer MeasurementsFigure 3: Lateral Pile Analysis Results for 610x12.7 mm Pier Pipe PilesFigure 4: Lateral Pile Analysis Results for HP360x132 mm Integral Abutment H-PilesFigure 5: Lateral Earth Pressure Induced by CompactionFigure 6: Lateral Pressures Due to Surcharge Point and Line LoadsFigure 7: Pavement Serviceability Loss versus Time for Swelling and Frost
LIST OF APPENDICES
Appendix A Preliminary Drawings for Bridge Option 1 and 2 Provided by Tetra Tech
Appendix B Test Holes Logs & Explanation of Terms and Symbols
Appendix C Pavement Core Photographs
Appendix D TH01 Photographic Core Log
Appendix E Additional Lateral Pile Analysis Results
Appendix F Slope Stability Analysis Results
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 4
As authorized by Mr. Vaibhav Banthia, P.Eng for Tetra Tech WEI Inc. (Tetra Tech), Amec
Foster Wheeler Environment and Infrastructure, a Division of Amec Foster Wheeler Americas
Limited (Amec Foster Wheeler) conducted a geotechnical investigation for the proposed
Saskatchewan Avenue at Sturgeon Creek Culvert Replacement. The purpose of the
geotechnical investigation was to establish the general soil and groundwater conditions at the
Site, and on that basis, provide geotechnical recommendations for Tetra Tech to use in
conceptual development and proposal of a minimum of two crossing alternatives to the City
of Winnipeg, followed by eventual detailed design of a final selected alternative. Structural
pavement cross-section alternatives for Saskatchewan Avenue were also to be provided.
The following geotechnical report pertains to preliminary design of selected crossing
alternative(s).
The proposed project consists of undertaking engineering, design, and contract administration
services for replacement of the existing two-cell concrete box culvert Saskatchewan Avenuecrossing of Sturgeon Creek. Replacement of the existing crossing is being undertaken for the
following reasons:
The existing culvert is considered to be in fair to poor condition with signs of concretedeterioration.
Overtopping of Saskatchewan Avenue on several occasions in the past as well as
sever scour damage on the downstream wing walls indicate the existing crossing ishydraulically deficient to handle the Sturgeon Creek flows in spring run-off and heavy
rainfall conditions.
It is anticipated flows on Sturgeon Creek will increase as future development occurs
upstream, which may result in increasingly unstable conditions for the existingSaskatchewan Avenue crossing as well as the CP Rail crossing located approximately
20 m north.
Further to replacement of the existing box culvert crossing, the project includes the following:
There is possibility of unstable creek banks, and in this regard, investigation of slopestability approximately 300 m upstream and downstream of Saskatchewan Avenue is
required to determine what measures, if any, are required to maintain stability.
There is a desire to extend the existing multi-use pathway; located south of the culverton the east bank of the creek; across Sturgeon Creek to Cavalier Avenue.
Reconstruction of Saskatchewan Avenue between Hamilton Avenue and Cavalier
Drive.
In a meeting held at Tetra Tech’s Winnipeg office on 26 June 2016, Tetra Tech presented the
City of Winnipeg with multiple bridge crossing alternatives to the City of Winnipeg. The
alternatives varied in the span configuration (i.e. single span versus 3 span) and alignment
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 5
(i.e. maintain existing alignment or realign Saskatchewan Avenue to the South). The meeting
concluded with Amec Foster Wheeler understanding the City of Winnipeg has selected the
following two options to carry further through preliminary design for final design selection.
Preliminary drawings for each of the two options are included in Appendix A, and the options
are summarized as follows:
Option 1: Realignment of Saskatchewan Avenue south of existing and construction ofa three span bridge with 5.5 horizontal to 1 vertical headslopes. Drawings indicate
integral abutments supported on vertical driven steel H-Piles and piers supported on
vertical driven pipe piles.
Option 2: Realignment of Saskatchewan Avenue south of existing and construction ofa single span bridge with 4 horizontal to 1 vertical headslopes. Drawings indicate semi-
integral abutments supported on vertical driven steel H-Piles (back row vertical, front
row inclined).
Prior to initiating drilling on any occasion, Amec Foster Wheeler notified public utility providers
(i.e. Manitoba Hydro, MTS, City of Winnipeg, etc.) of the intent to drill in order to clear public
utilities, and where required, met with said representatives on-site. Amec Foster Wheeler also
retained the services of a private utility locator to confirm clearance from privately owned
utilities near the test hole locations.
On 6 May through 11 May 2016, Amec Foster Wheeler supervised the drilling of a total of
sixteen test holes at the approximate locations illustrated in Figure 1. Test holes TH01 andTH02 comprised deep test holes advanced near the anticipated location of new bridge
abutments, south of the existing bridge abutments. TH03 through TH06 comprised test holes
advanced through the Sturgeon Creek slopes north and south of Saskatchewan Avenue to
just below the top of underlying till. The remaining ten test holes (RW-01 through RW-10)
comprised roadway (i.e. RW) holes advanced to about 2.0 m below grade to evaluated
subgrade conditions for pavement structures along Saskatchewan Avenue. All ‘TH’ test holes
were advanced by Maple Leaf Drilling Ltd. using a track mounted Mobile B54X drill rig
equipped with 125 mm diameter Solid Stem Augers & HQ Coring. All ‘RW’ test holes were
advanced by Maple Leaf Drilling Ltd. using a truck mounted CME55 drill rig equipped with
125 mm diameter Solid Stem Augers.
During drilling, Amec Foster Wheeler field personnel visually classified the observed soils
according to the Modified Unified Soil Classification System (MUSCS). Groundwater and
drilling conditions were also recorded at the time of drilling. Grab samples were collected at
selected depths from the auger cuttings, while relatively undisturbed Shelby tube samples
were also collected from TH03, TH04, and TH06 at selected depths ranging from about 3.0 m
to 6.1 m below grade. Split spoon samples of the till at depth were collected from each of
TH01 through TH06. Split spoon samples of the clay from about 1.5 m to 2.0 m below grade
were also obtained at each of the ‘RW’ holes. The in-situ relative consistency of cohesive
overburden was evaluated within all test holes using a pocket penetrometer, as well as within
the ‘RW’ test holes using standard penetration tests (SPT), where the number of blows to
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 6
drive a split spoon sampler 0.3 m into the soil was recorded. The relative consistency of the
underlying till within all ‘TH’ test holes was evaluated using SPT results. The recorded number
of blows is shown on the logs as the SPT (N) value.
All test holes were left open for approximately ten minutes after completion of drilling to
observe the short-term groundwater seepage and sloughing conditions. TH01 through TH06
were backfilled to grade using auger cuttings and bentonite. All ‘RW’ test holes were backfilled
to approximately 100 mm below existing pavement using auger cuttings, bentonite, and gravel
fill, and capped with 100 mm of asphalt cold patch.
Following completion of the field drilling program, a laboratory testing program was conducted
on selected soil samples obtained from the test holes. The laboratory testing program
completed consisted of moisture content determinations and four unconfined compressive
strength (UCS) tests completed in accordance with ASTM Standard D2166.
Detailed test hole logs summarizing the sampling, field testing, laboratory test results, and
subsurface conditions encountered at the test hole locations are presented in Appendix B.Actual depths noted on the test hole logs may vary by ± 0.3 m from those recorded due to the
method by which the soil cuttings are returned to the surface. Summaries of the terms and
symbols used on the test hole log and of the Modified Unified Soil Classification System are
also presented in Appendix B.
4.1 Stratigraphy
Consistent with the regional geology and anticipated conditions, the stratigraphy at the test
hole locations consisted of the following, in descending order from grade level:
Organic Clay (TH01 through TH06 only)
140 mm to 280 mm of Asphalt and/or Concrete Pavement underlain by gravel fill tobetween about 280 m and 380 m below pavement surface (RW holes only)
Clay (All test holes)
Glacial Till ( TH01 through TH06 only, encountered between 229.0 and 230.0 m)
Limestone Bedrock (TH01 only at elevation 217.7 m)
Generalized descriptions of each of the soil layers are provided below. Detailed descriptions
of the soil layers above are presented in the test hole logs in Appendix B.
Organic Clay
Grass surfaced organic clay was encountered at the surface of all test holes other than the
‘RW’ holes advanced through existing pavement along Saskatchewan Avenue. The organic
clay extended to between about 25 mm and 75 mm below grade (60 mm average), and was
generally described as silty with frequent rootlets, medium to high plastic, moist, firm, and dark
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 7
grey to black. It should be noted that the thickness of organic clay over the project extents
may vary from that observed at the test hole locations.
Existing Pavement
Existing pavement encountered along Saskatchewan Avenue consisted of a mix of pavement
structure including concrete only, asphalt only, or concrete pavement with asphalt overlay. A
summary of the pavement structure encountered at each test hole is provided in Table 4-1.
Photographs of each of the recovered pavement cores are presented in Appendix C.
Table 4-1: Summary of Existing Pavement and Gravel Thickness
Test Hole ID Pavement Structure Gravel Thickness (mm)
RW-01 200 mm Concrete 100
RW-0225 mm Asphalt
175 mm Concrete80
RW-03 175 mm Asphalt 200
RW-04 175 mm Asphalt 130
RW-05 175 mm Asphalt 200
RW-06 175 mm Asphalt 200
RW-0750 mm Asphalt
230 mm Concrete30
RW-0850 mm Asphalt
230 mm Concrete90
RW-0965 mm Asphalt
215 mm Concrete100
RW-1065 mm Asphalt
165 mm Concrete50
Clay
Consistent with expected geology within and surrounding Winnipeg, overburden beneath the
existing topsoil and pavement consisted of clay. The clay was generally described as silty,
high plastic, moist, stiff to very stiff becoming firm below about 1.5 m below existing grade,
and dark grey to grey.
Atterberg Limits and Particle Size Analyses by Hydrometer Method were undertaken on a
total of seven samples obtained from various ‘TH’ and ‘RW’ test holes, the results of which
are summarized in Table 4-2. Unconfined compression strength tests were completed on
three Shelby Tube samples collected from TH03, TH04, and TH06, the results of which are
summarize in Table 4-3.
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 8
Table 4-2: Summary of Atterberg Limit and Particle Size Analysis Results
Test HoleDepth
(m)
Elev.
(m)
Liquid
Limit
(%)
Plastic
Limit
(%)
Gravel
(%)
Sand
(%)
Silt
(%)
Clay
(%)
RW03 0.5 236.6 80 20 0.5 15.2 15.5 68.7
RW04 0.5 235.9 82 23 3.0 15.3 17.4 64.3
RW07 0.5 236.9 90 26 0.0 3.2 21.1 75.8
RW09 0.5 237.8 87 20 0.0 0.8 27.2 72.0
TH03 3.0 231.4 96 21 0.6 4.3 7.0 88.1
TH04 3.0 229.9 92 20 0.0 3.9 18.6 77.5
TH06 3.0 232.9 76 19 0.0 2.4 30.5 67.2
Table 4-3: Summary of Unconfined Compressive Strength Tests
Test HoleDepth
(m)
Elev.
(m)UCS (kPa)
Strain at 100%
of UCS
(%)
Bulk Density
(kg/m3)
Dry Density
(kg/m3)
TH03 3.0 231.4 58 3.7 1842 1259
TH04 3.0 229.9 65 6.6 1733 1134
TH06 3.0 232.9 106 2.0 1787 1212
Glacial Till
Till was encountered beneath the clay overburden at each of test hole TH01 through TH06 at
depths ranging from about 2.9 m to 6.4 m below existing grade; equivalent to between
approximate elevations 229.0 m and 230.0 m.
The till was generally described as loose to very loose within the upper 0.6 m to 1.0 m below
the surface of the till, transitioning to very dense below. Cemented till resulting in core recovery
was observed at TH01, and is shown in the photo log provided in Appendix D. The
composition of the till varied highly, both with depth within a single test hole and between the
test hole locations. In general, some till zones were dominated by a silt and sand matrix with
some gravel and trace clay. Other zones were essentially ‘clean’ with no fines, and were
dominated by ‘clean’ gravel with trace sand. Cobble and boulder sizes were also encounteredthrough the till, and even resulting in breakage of the coring bit at TH02 forcing the hole to be
abandoned.
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 9
Bedrock
Bedrock was explored at TH01 only. A photographic log of the bedrock core obtained at TH01
is provided in Appendix D.
In summary, the bedrock was described as medium strong, moderately weathered, dark grey
to grey limestone interbedded with frequent reddish grey mudstone layers 10 mm to 75 mm
thick, and abundant subrounded to rounded clasts. Rock quality designation (RQD) values
ranged from about 63% from 0 to 1.5 m below the surface of the bedrock, to about 94% from
about 1.5 m to 3.0 m below bedrock surface.
4.2 Auger Refusal
Auger refusal occurred in TH01 at approximate elevation 228.1 m (5.3 m below grade) and in
TH02 at approximate elevation 228.3 m (5.6 m below grade) prior to switching to HQ coring.
Test holes TH03 through TH06 achieved target termination depth and were not taken to auger
refusal.
4.3 Groundwater and Sloughing Conditions
Seepage and sloughing conditions were noted during drilling, and the depth to the
accumulated water level within the test holes was measured about ten minutes after drilling
at each test hole location. Recorded observations at test hole locations TH01 through TH06
are summarized in Table 4-4. Neither seepage nor groundwater was observed in any of the
‘RW’ test holes advanced through Saskatchewan Avenue to 2.0 m below grade.
Table 4-4: Observed Slough and Groundwater Conditions During Drilling
TestHole
ID
TestHole
Depth(m)
During Drilling Upon Completion
Sloughing
Zone
Seepage
Zone
Depth to
Slough (m)
Depth to
Groundwater (m)
TH01 18.7 Below 14.6 m
Flowing artesianconditions
(approx. 4 L/min.)encountered upondrilling to 15.5 m
N/A - Cased testhole for wet coring
-0.3 (i.e. aboveexisting grade)
TH02 9.0 Below 5.5 m Below 5.5 mN/A - Cased test
hole for wet coring3.0
TH03 6.6 Below 6.4 m Below 6.4 m 5.8 4.0
TH04 4.8 Below 4.6 m Below 4.6 m 4.3 1.8
TH05 3.5 None At 3.5 m 3.5 Trace at bottom
TH06 7.2 None None 7.0 Dry
Further to seepage and groundwater observation at the time of drilling, TH01, TH02, TH04,
and TH05 were instrumented with slotted standpipe piezometers in order to allow for short
term monitoring of groundwater levels within the till (TH04 and TH05) and bedrock (TH01 and
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 10
TH02). Vibrating Wire (VW) piezometers equipped with single channel data logger obtaining
measurements at 4 hour intervals since installation were installed within the clay overburden
at TH03 and TH06 in order to monitor porepressure within the Sturgeon Creek slopes
upstream and downstream of Saskatchewan Avenue. All recorded standpipe and VW
piezometer readings obtained through 29 June 2016 are illustrated in Figure 2.
In summary, seepage and groundwater observation during and post drilling indicate an
artesian bedrock condition with an estimated phreatic surface of up to elevation 238.0 m north
of Saskatchewan Avenue, and up to about 235.1 m south of Saskatchewan Avenue.
Per clause D5.9 of City of Winnipeg RFP 3-2016, the bridge shall be designed in accordance
with CAN/CSA-S6-06, the Canadian Highway Bridge Design Code (CHBDC), to be
structurally and functionally safe for the duration of a minimum 75 year design life.
Minimum requirements for the design of foundations and geotechnical systems are outlined
in Section 6 of the CHBDC. CHBDC employs Limit States Design principles, wherebygeotechnical resistance factors are applied to the ultimate geotechnical resistance to obtain a
factored geotechnical resistance for the specified limit state. The geotechnical resistance
factor to be applied in design is to be selected from those outlined in the CHBDC on the basis
of the degree of understanding of the site and prediction model for design. Based on the
regional geology, subsurface conditions encountered at the site, and Amec Foster Wheeler’s
experience and empirical knowledge of local foundation performance within the City of
Winnipeg, Amec Foster Wheeler recommends the geotechnical resistance factors outlined in
Table 5-1, selected from those outlined in the CHBDC. The geotechnical resistance factor to
be applied in design shall be selected in accordance with the Analysis Method / Predictive
model used in final selection of the pile driving equipment and development of driving criteria.
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 11
Table 5-1: Recommended Geotechnical Resistance Factors for Driven Steel Piles
Limit State Analysis Method / Predictive ModelResistance
Factor
Ultimate Limit State (ULS)
Compression, ϕgu
Static Analysis without driveability analysis (WEAP),
dynamic pile measurements, or load tests.0.40
Hammer selection and driving criteria established by
driveability analysis (WEAP), without dynamic pile
measurements or load testing
0.45
Driving criteria established by driveability analysis (WEAP)
with confirmation of hammer performance with dynamic pile
(PDA) measurements and signal matching (CAPWAP)
0.50
Tension, ϕgu
Static Analysis 0.30
Dynamic Testing with Signal Matching (CAPWAP) 0.40
Lateral, ϕgu All analysis methods 0.50
Serviceability Limit State (SLS)
Settlement or
Lateral Deflection,
ϕgs
Static Analysis 0.9
Based on subsurface soil and groundwater conditions encountered at the test hole location,
the following key geotechnical considerations have been identified and will need to be
considered throughout design:
6.1 Spatial Variability in Soil and Groundwater Conditions
The soil and groundwater conditions encountered during the investigation were characterized
based on conditions observed in small diameter test holes advanced at relatively wide
spacing. Subsurface conditions at locations that were not investigated could vary from the
conditions observed in the test holes. Spatial variability of sub-surface conditions should be
expected and allowed for in the design and construction processes.
The stratigraphy and soil encountered within the test holes advanced at the site are
considered typical of geologic conditions within the region and are consistent with anticipated
subsurface conditions. However the depth to bedrock was greater than anticipated and
resulted in greater than normal/average till thickness.
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 12
6.2 Foundation Strength and Deformability
The high plastic lacustrine clay overburden is highly deformable (both vertically and laterally)
under surcharge loads. Construction of embankment fills over the compressible native
lacustrine clay foundation soils will result in long term consolidation of the clay foundation
soils, and by extension, settlement of embankment fill. Lateral deformation of the
embankments and underlying native clay soils will also occur following loading, and may be
more pronounced below the edges of embankment areas. In this regard, selection and design
of foundations must consider the potential impact of new embankment fill on both vertical and
lateral soil deformations around the foundation.
Notwithstanding the above, based on comparisons of the preliminary top of deck elevation of
about 236.150 m to existing roadway and ditch elevations along Saskatchewan Avenue, fill
requirements are expected to be limited to less than 1 m over the project extents. This
includes any fill required to construct the headslopes and sideslopes at the proposed
abutments, for both Options 1 and 2. In this regard, long terms consolidation of foundationclays is not expected to be an issue pending confirmation of fill depths less than 1 m during
detailed design. If during detailed design fill depths greater than 1 m are identified, the impact
of these fill depths on foundation performance should be evaluated by Amec Foster Wheeler
during detailed design.
6.3 Artesian Groundwater
Observed groundwater levels within the till and bedrock and porewater pressures measured
within the clay identify phreatic surfaces above existing grade. In this regard, groundwater
conditions, at the site, in particular artesian groundwater pressure with the till and bedrock,
present a high risk for basal heave and excavation instability. The clay overburden is an
aquiclude, the removal and/or thinning of which could result in basal heave and development
of a flowing artesian condition, which would necessitate temporary depressurization of the
aquifer to complete construction and restore the aquiclude / stability. Excavations presenting
risks to the project include all temporary excavations, as well as any permanent excavations
or voids inclusive of annular voids around integral abutment piles (to allow abutment
movement due to thermal expansion and contraction). The impact of cofferdam construction
and any dewatering and/or subexcavation of the creek bed shall be considered throughout
design, particularly as it relates to constructability.
To the best achievable extent, design and construction planning should seek to keep all
temporary and permanent excavation requirements to a minimum; preferably zero. In
particular, design and construction planning should seek to eliminate subexcavation of theexisting creek bed as comparison of artesian pressures in the underlying till and bedrock
suggest that Sturgeon Creek, at this location, may be a zone of perpetual groundwater
discharge. The rate of groundwater discharge at the base of the creek is mitigated by head
loss through the clay layer between the base of the creek and underlying till.
Additional discussion and recommendations for determining the factory of safety against basal
heave is presented in Section 7.0.
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 13
Artesian groundwater pressures with the underlying till and bedrock require that care be taken
in selection and design of the required dewatering, and excavation depths (both temporary
and permanent) in order to mitigate the risk of basal heave and loss of stability. This includes
evaluating the impact of temporary dewatering on stability of the bed of Sturgeon Creek, as
well as evaluation for the potential of developing perpetual seepage and groundwater flow
along integral abutment piles where annular voids are provided to accommodate movement
of the abutments.
According to Canadian Foundation Engineering Manual (CFEM), the porewater pressure
acting at the bottom of an aquiclude (i.e. on the underside of the highly plastic overburden)
should not exceed 70% of the total weight of soil and groundwater above this depth. In other
words, a minimum factor of safety of 1.4 against basal heave is recommended, whereby the
factor of safety is determined as the ratio of total weight above the point of pressure to the
porewater pressure acting at that point.
For evaluation purposes, the base of the aquiclude may be assumed equivalent to the surface
of the glacial till. Till surface elevations varied at the test hole locations between about 229 m
to 230 m. Based on the water levels measured at the standpipe locations (see Figure 2),
porewater pressure acting on the underside of the aquiclude may be determined assuming a
phreatic surface (i.e. total head) of 234.5 m. This equates to a pressure head of 55 kPa acting
on the underside of the aquiclude for a till surface elevation of 229.0 m, and 45 kPa for a till
surface elevation of 230.0 m.
Assuming a creek bed elevation of 231.2 m and a bulk density of 16 kN/m3 for overburden
soil comprising the creek bed, the factor of safety against basal heave as a function of till
elevation and water level in the creek is outlined in Table 7-1.
Table 7-1: Factor of Safety Against Basal Heave versus Sturgeon Creek Water Level
Water Elevation in Sturgeon CreekFactor of Safety
Till Elev. 229.0 m Till Elev. 230.0 m
231.2
(Dewatered Channel)0.64 0.35
232.67
(Water Level 22 June 2016)1.33 1.04
233.37 (Q50) 1.66 1.37
234.52 (Q1) 2.19 1.91
In summary, the results indicate that dewatering of the creek will severely jeopardize the
stability of the base of the creek against basal heave. In this regard, localized depressurization
of the aquifer beneath the creek bed would be required to ensure a stable creek bed
throughout construction. The pumping/extraction rates required to lower the groundwater
pressure head are dependent on the transmissivity of the aquifer, Completion of a
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hydrogeological study and well pumping program was beyond the scope of this report;
however, would be undertaken during detailed design in support of a depressurization system
of the underlying aquifer.
Further to stability of the creek, determination of the required overburden pressure to resist
basal heave indicates permanent excavations for the annular voids around integral abutment
piles extending below elevation 233.9 m will require some form of seal to mitigate groundwater
flow through the void space. Recommendations for design of a seal are discussed in Section
8.4.
8.1 General Evaluation
Based on preliminary design discussions with Tetra Tech, Amec Foster Wheeler understands
driven steel piles are the preferred foundation alternative. Based on the subsurface conditions
at the test hole locations and the key geotechnical considerations discussed in Section 0,
Amec Foster Wheeler supports driven steel piles as the preferred pile foundation alternativeat this site. Where employed, it is expected that driven steel piles would be driven to ‘practical
refusal’ refusal within the underlying till and/or on bedrock.
Piles at this site are expected to develop greater than 90 percent of their resistance in
combined shaft friction and end bearing within the till. The depth of penetration into the till will
be restricted by end bearing development which in turn will be highly influenced by the
presence of cobbles at boulders at the pile tip. In this regard, the variability in achieved (i.e.
as-built) pile embedment depths is expected to be greater than normal. Further, cobbles and
boulders induce severe driving conditions and high localized driving stresses within the pile
sections. As such, steel pile sections are recommended due to their ductility. Driven pre-cast
pre-stressed concrete piles are susceptible to brittle failure and/or breakage at the pile tip and
are not recommended. The use of conventionally bored friction piles is not recommended for
support of bridge structures given the limited thickness and available resistance of the clay
overburden. Furthermore, extension of bored piles into the underlying till is not recommended
due to the artesian conditions and potential development of flowing groundwater conditions
discussed in Section 6.3.
Based on the discussion above, foundation recommendations presented in this report are
limited to driven steel piles. Foundation recommendations for alternate pile types can be
provided upon request.
8.2 Driven Steel Piles
8.2.1 General Discussion
Amec Foster Wheeler understood preliminary designs propose to use driven steel H-Piles for
support of the bridge abutments, where-as pipe piles are proposed for support of piers. The
following additional comments are provided with respect to driven steel piles:
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Amec Foster Wheeler anticipates steel piles could be driven to end bearing in theunderlying silt till and/or bedrock. It is anticipated that steel piles could be driven with
relative ease through clay embankment fill and overburden; however, severe driving
conditions associated with cobbles and boulders in the till and end bearing on rock
are expected. In this regard, the toes of all driven steel piles should be equipped with
am internally fitted, hardened (i.e. cast) steel driving shoe continuously welded or
fastened to the pile in order to reinforce the toe of the pile during driving.
Both H-Pile and Pipe piles are expected to develop a plugged toe condition (i.e. endbearing applied to full cross-sectional area of pile tip) with penetration into the
underlying till.
H-Piles are anticipated to penetrate the till to grater depths than pipe piles.
Due to artesian conditions within the underlying till, the interior of pipe piles shouldbe infilled with concrete to the underside of pile cap in order to prevent loss of pile
toe support associated with upward seepage, basal heave, piping at the pile.
Where a CSP (or other material) is provided to maintain a void around integralabutment, special precautions such as a grout seal between the pile and the CSP
will be required to prevent upward seepage and groundwater flow along the pile
induced by artesian pressure in the underlying till and bedrock.
8.2.2 Geotechnical Bearing Resistance
8.2.2.1 Strength Limit State
The ultimate geotechnical resistance of a steel pile driven to ‘practical refusal’ using an
appropriately sized hammer and driving force and energy shall be limited to a maximum given
by the structural column capacity of the steel pile section, which may be taken as 0.63FyAs
for severe driving conditions, where Fy is the yield strength of the steel, and As is the cross-
sectional steel area. The geotechnical resistance factor for determination of factored bearing
resistance should be selected in accordance with the recommendations in Section 5.0 in
consideration of pile load testing undertaken at the time of construction.
Additional recommendations for design and construction of driven steel piles are as follow:
The pile capacity given by the above pertains to the geotechnical resistance of a fullyembedded with zero unsupported pile length.
Pile cross sections must be structurally designed to withstand the design loads andthe driving forces during installation. Evaluation of the structural resistance of piles
shall be undertaken by the structural engineer, and should consider laterally
unsupported pile length. Examples of unsupported pile length include pile stick-up, or
any annular void space provided around any embedded portion of the pile to preclude
lateral resistance, such as is proposed by Tetra Tech for the integral abutment
alternative.
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The effect of corrosion and deterioration from environmental conditions shall beconsidered in selection of the required pile cross-section for long term pile capacity.
The potential for corrosion and anticipated corrosion rates should be investigated
during detailed design by a corrosion expert. Amec Foster Wheeler anticipates that
corrosion of steel piles will likely be addressed by sacrificial steel thickness, although
other alternatives may be adopted.
As a minimum, steel H-piles should meet the requirements ofCAN/CSA-G40.20/G40.21, Grade 350 W, and pipe piles should have a minimum yield
strength of 310 MPa (i.e. ASTM A252 Grade 3 steel). The toes of all H-piles shall be
equipped with a hardened (i.e. cast) steel driving shoe continuously welded or
fastened to the pile in order to reinforce the toe of the pile during driving.
Wave Equation and driveability analysis (i.e. WEAP) should be completed for each ofthe selected pile sections, prior to proceeding to construction and concurrent with
selection of the pile driving equipment, to confirm the ability of the proposed hammer
and appurtenances to drive the piles to the required design capacity and embedment
depth without damage. Similarly, the WEAP shall be extended to develop termination
criteria for use in pile installation monitoring.
An appropriately selected pile driving hammer and appurtenances shall be capable ofdriving the specified pile section to the design ultimate geotechnical resistance at a
termination criterion (or pile driving acceptance criterion) defined by a penetration
resistance of not less than 5 blows per 25 mm of penetration and no more than 15
blows per 25 mm of penetration. In order to mitigate the risk of damage to both the pile
and piling equipment, continuous driving of a pile at penetration resistances exceeding
15 blows per 25 mm of penetration should be avoided.
To reduce the potential for structural damage to the pile, maximum compression andtension driving stresses developed within the pile during installation shall be limited to
≤0.9Fy for steel piles. If WEAP analysis prior to construction predicts driving stresses
exceeding 0.9Fy, then the foundation design and/or selected pile section shall be
revised until the design pile resistance is achieved within the recommended driving
stress limit.
For an appropriately sized pile driving hammer, it is anticipated that piles could be
driven through the embankment fill and clay overburden with relative ease. However,
potentially highly variable pile capacity development (i.e. driving resistance) is
anticipated over the depth of penetration into the underlying till. In particular, potential
cobbles and boulders in the till could provide for difficult driving conditions and sudden
refusal. A contingency should be carried to allow for potentially highly variable pile
lengths across the site.
Excessive penetration resistance (i.e. values greater than 15 blows per 25 mm ofpenetration) is expected to develop shortly following contact of the pile toe with the
bedrock surface. It should also be noted that potential cobbles and boulders in the till
could provide for increased penetration resistance and sudden refusal. As indicated
above, in order to mitigate the risk of damage to both the pile and piling equipment,
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continuous driving of a pile at penetration resistances exceeding 15 blows per 25 mm
of penetration should be avoided.
If damage to a pile is observed during driving, driving should cease immediately, andthe pile capacity and extent of damage assessed by a qualified geotechnical engineer
and structural engineer. This may include estimating the pile capacity and assessing
damage by dynamic testing. Any piles that have been damaged, are excessively out
of plumb, or have refused prematurely may need to be replaced pending the results of
the review.
Dynamic (PDA) testing of a minimum given by the greater of 5% of the total pile countand one pile per pier and abutment substructure unit should be undertaken to validate
the pre-construction WEAP analysis and verify that pile driven to the established
termination criteria meet design requirements. Dynamic pile measurements (PDA)
should also be used to monitor for indications of pile damage.
Prior to the pile installation, the piles should be inspected to confirm that the materialspecifications are satisfied. The piles should be free from protrusions, which could
create voids in the soil around the pile during driving.
All piles driven within five pile diameters of a previously driven pile should have thepreviously driven pile monitored for heave and, where heave is observed, the heaved
pile should be re-driven. Piles that are re-driven should be advanced to at least the
original elevation.
As driven steel pile installations do not allow for direct confirmation of soil conditionsduring construction, and the piles cannot generally be visually inspected for damage
following installation, construction monitoring will be important in quality control and
quality assurance of pile installations, and to verify that the piles are installed in
accordance with design assumptions and the driving criteria are satisfied. Construction
monitoring of pile installations should be undertaken on a full time basis, and should
include, but not be limited to, confirmation that pile materials meet or exceed
specifications, confirmation of hammer and appurtenance operating conditions, and a
detailed driving record inclusive of penetration resistance (i.e. blow counts per unit
penetration) and blow rate or energy. Completed pile driving records should be
reviewed on a regular basis during pile driving by a qualified geotechnical engineer.
8.2.2.2 Service Limit State
The settlement of a single pile depends on the applied load, strength-deformation properties
of the foundation soils, load transfer mechanism, load distribution over the pile embedment
depth, and the relative proportions of the load carried by shaft friction and end-bearing. A pile
settlement limit value was not specified by the structural agent for use in developing
geotechnical resistance limits for the serviceability limit state design criterion.
The settlement of a single pile driven to refusal in the till or on the underlying bedrock (using
an appropriately sized hammer and driving energy) is expected to be governed primarily by
elastic shortening of the pile section under applied loads. It is further anticipated that the
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majority of settlement would occur during construction under the progressive application of
sustained dead loads.
Abutments piles will be required to extend through embankment fill. Consolidation of the
embankment fill and underlying compressible clay would induce negative skin friction or drag
load on piles driven to refusal in the till or on the bedrock, and would result in additional stress
development within the pile section and increased elastic shortening of the pile. Consequently,
it is important that the potential for drag loads be considered in addition to structural loads
when estimating the settlement performance of piles extending through fill.
The estimated total pile head settlement of a steel pile driven to practical refusal in accordance
with the recommendations of this report can be estimated as a function of elastic shortening
of the pile under the applied service load plus drag load. It is recommended that the settlement
performance of the pile be approximated assuming a fully end bearing pile (i.e. zero shaft
resistance) where the drag load is assumed to be applied at the top of the pile (rather than
along the pile shaft) as follows (CFEM 3, 1992):
where: S = Total pile head settlement (m)
b/100 = Approximate toe mobilization (m)
b = Pile diameter (O.D., in m)
es = The elastic compression of the pile (m)Q = Sum of applied unfactored working load and drag load (kN)
L = Pile length (m)
A = Cross-sectional area of the pile material (m2)
E = Elastic modulus of the pile material (kPa)
Recommendations for determining the drag load are discussed in Section 9.0.
8.2.3 Tensile (Uplift) Resistance
8.2.3.1 Strength Limit State
When tensile forces are present, the ultimate tensile resistance of driven piles should be
determined using unit shaft friction values outlined in Table 8-1. For pipe piles, only the
exterior surface area of the pile in contact with the soil should be used in the calculation of thefrictional resistance. In the case of steel H piles, the surface area should include the exterior
sides of the two flanges plus twice the depth of the web. Although not commonly employed
for the installation of driven steel piles, if pre-boring is required (i.e. for ground disturbance
clearance or contractor preference), shaft friction must be neglected over the depth of the pre-
bore for H-Piles, and over the depth of the pre-bore for pipe piles if the pre-bore exceeds 95
percent the pile diameter. If the pre-bore is limited to no more than 95 percent of the pile
diameter, no reduction in shaft friction is required for pipe piles. Shaft resistance should also
AEQLbbS es
100100δ
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be neglected over the portion of the piles that may be inserted within sleeves, such as has
been proposed by Tetra-Tech for the integral abutment alternative (Option 1).
Table 8-1: Recommended Parameters for Tensile Resistance of Driven Piles
Elevation Range Anticipated Soil TypeUltimate Unit
Shaft Friction
El. 233.0 to El. 229.0 Clay Fill / Clay 30*
El. 229.0 to El. 228.0 Loose Till 30
El. 228.0 to El. 222.0 Dense Till
Linearly Increasing with depth from:
30 kPa at El. 228.0 to
90 kPa at El. 222.0
Below El. 222.0 Dense Till 90 kPa
* The upper 2.4 m of the shaft, below final grade, is to be ignored in estimating shaft friction to account for the loss of
contact between the soil and pile interface, such as may result from seasonal frost, moisture changes, etc.
The geotechnical resistance factor for determination of factored bearing resistance should be
selected in accordance with the recommendations in Section 5.0 for pile design using Static
Analysis.
8.2.3.2 Service Limit State
The upward displacement of a pile in tension can be estimated in the same manner as
estimating the settlement of a friction pile in compression. In this regard, the upward
displacement of a driven steel pile under in tension under a maximum factored load given by
the recommendations in Section 8.2.3.2 is expected to be in the range of 0.05 to 0.2 percent
of the shaft diameter, plus elongation of the pile due to the applied tension load.
8.2.4 Lateral Pile Resistance
8.2.4.1 Strength Limit State
Piles resist laterally applied loads by deflecting until the necessary resistance is mobilized in
the adjacent soils. The lateral capacity depends upon the properties of the soil and pile
materials, pile size, fixity at the top of the piles, depth of embedment, load distribution along
the pile, and tolerable deflections. Where lateral pile resistance is required, it is recommended
that the nominal horizontal load resistance of piles be estimated using procedures that
consider soil-structure interaction, such as the method of non-linear p-y curves, whereby
horizontal resistance is estimated based on both the non-linear strength-deformation
characteristics of the soil stratum surrounding the pile and the structural properties of the pile.
Based on Reese and others (1984), the soil reaction (p) is related to the shaft deflection (y)
for various depths below the ground surface. In general, p-y curves are nonlinear and depend
on several parameters, including depth, shaft diameter and soil strength.
Based on conditions observed within the appended test hole logs, the stratigraphy and soil
parameters outlined in Table 8-2 are considered suitably representative of the average
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subsurface conditions expected to influence the lateral behaviour of driven steel piles at the
site. Revisions to the stratigraphy outlined below may be required by design and construction
requirements (i.e. at abutments where sand or gravel may be used as opposed to clay), and
as such, should be reviewed by the design engineering during design. Furthermore, Amec
Foster Wheeler should be notified of any deviations from the stratigraphy outlined in Table
8-2 for review.
Table 8-2: LPile Input Parameters for Lateral Pile Analysis
Elevation Range (m)Soil
Type
Soil
Model
Effective
Unit
Weight
(kN/m3)
Friction
Angle
(°)
Undrained
Shear
Strength
(kPa)
E50
(%)
p-y
subgrade
modulus, k
(kPa/m)
Final grade to:
underside of rip-rap
Rip-
Rap
Ignore strength contribution provided by rip-rap in the event of movement
and formation of a gap at the rip-rap / pile interface.
Lesser of: final grade
above existing or
underside of rip-rap to:
El. 234.0
Clay
Fill
Soft Clay
(Matlock)17.5 n/a 30 0.0088 Default
Lesser of: final grade
below El. 234.0 m,
underside of rip-rap
layer, or El. 234.0 m to:
El. 229.0
Clay
Fill /
Clay
Soft Clay
(Matlock)7.7 n/a 30 0.0088 Default
El. 229.0 to El. 228.0Loose
TillAPI Sand 10.2 30 n/a n/a 16,000
El. 228.0 to El. 217.0Dense
TillAPI Sand 12.2 35 n/a n/a 24,000
The recommended ‘soil model’ outlined in Table 8-2 for analysis of the lateral load-
deformation of piles have been presented for the condition of intimate contact between the
pile and surrounding soil. Lateral soil resistance shall be neglected if the soil providing
resistance is, or is likely to become soft, loose, removed due to scouring, or disturbed, or if
the contact between the soil and wall is not tight. In this regard, lateral resistance should be
neglected over the depth of seasonal frost penetration, or in the case of semi-integral
abutments, over the length of the annular void maintained around the piles to allow movement
of the abutment (i.e. in response to thermal expansion and contraction of the bridge deck).
Boundary conditions for LPile analysis should further be selected in accordance with the
functional and performance requirements, and whether or not movement of the pile head is
load controlled (i.e. prediction of pile head deflection in response to applied loads), or
deflection controlled (i.e. evaluation of shear and bending moments within the pile in response
to known movement of the pile head).
Where piles are required to provide horizontal resistance, evaluation of the nominal horizontal
load resistance of a pile or pile group at the strength limit state requires that a criterion (or
criteria) defining the strength limit state of piles under horizontal load be defined. The nominal
horizontal resistance of a pile may be defined in accordance with one of the three following
theoretical scenarios (CFEM, 2006):
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1. The capacity of the soil may be exceeded, resulting in large horizontal movements of
the piles and failure of the foundation;
2. The bending moment (M) and/or shear (V) may generate excessive bending or shear
stresses in the pile material, resulting in structural failure of the piles; or
3. The deflection of the pile head may be too large to be compatible with the
superstructure (i.e. the deflection that can be tolerated at the foundation/
superstructure interface prior to initiating the ultimate limit state in the superstructure).
Case 1 is approximately similar to loss of fixity at the pile toe, and mobilization of plastic
deformation of the soil along the full embedment length of the pile. For this determination, the
loads applied to the pile are factored, and a soil resistance factor of 0.5 is used.
For Case 2, the method of p-y curves can be used to evaluate the bending moment and shear
generated in a given pile configuration under an applied load scenario. The nominal horizontal
resistance can then be determined as the maximum load that can be applied to the pile priorto exceeding the bending moment resistance or shear resistance of the pile.
For Case 3, the lateral deflection predicted by LPile under an applied loading condition should
be compared for compatibility with the maximum horizontal movement that can be tolerated
by the superstructure at the strength limit state. Note that the level of deflection that can be
tolerated at the strength limit state may be larger than the level of deflection defining the
serviceability limit state. Where the deflection that can be tolerated by the pile exceeds the
maximum tolerable level of deflection at the strength limit state, the nominal horizontal
resistance of the pile should be limited to the horizontal load scenario corresponding to the
maximum level deflection that can be tolerated by the superstructure.
8.2.4.2 Service Limit State
The horizontal movement of pile foundations shall be estimated using the method of non-
linear p-y curves discussed in Section 8.2.4.1. Tolerable horizontal movement of piles shall
be established on the basis of confirming compatible movements of structural components for
the loading condition being considered.
8.2.4.3 Lateral Pile Analysis Results for Select Piles Sizes
At the request of Tetra Tech, Amec Foster Wheeler undertook lateral pile analysis (using LPile
produced by ENSOFT Inc.) of the 610x12.7 mm pier pipe piles (concrete filled) and of the
HP360x132 abutment piles proposed for the three span bridge with integral abutment
alternative (Option 1). With respect to the H Piles, Amec Foster Wheeler understood analysis
was only required for flexure about the weak axis. LPile analyses were not required for Option
2 (Single Span Bridge) which employed a front row of inclined driven steel H-Piles to resist
lateral foundation loads.
Lateral loading analyses were performed for both fixed-head and free head conditions
assuming static, sustained loading conditions. With respect to the 610x12.7 mm pipe piles,
lateral displacement of the piers and piles is expected to be governed by applied loads (i.e.
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load controlled boundary condition). In this regard, the purpose of LPile modelling was to
predict lateral pile movement in response to applied loads. Modelling of the free head
condition for the 610x12.7 mm pipe piles consisted of predicting the lateral deflection of the
pile for a specified combination of lateral load, vertical load, and moment applied to the pile
head. Modelling of the fixed head condition consisted of predicting the lateral deflection of the
pile for a specified combination of lateral and vertical force applied to the pile head while
maintain zero rotation of the pile head.
With respect to the HP360-132 integral abutment piles, the magnitude of lateral pile
displacement is restricted by thermal expansion and contraction of the structure (i.e.
displacement controlled boundary condition). Integral abutments seek to allow this movement
to occur with minimal soil resistance, and in this regard, the purpose of LPile analysis was to
determine internal shear and bending moments for selection of an adequate pile section. Tetra
Tech also requested p-y curve definitions for modelling of the H-Piles in structural analysis
software. Modelling of the free head condition for the HP360x132 piles for integral abutmentconsisted of predicting soil resistance over the pile length and internal shear and bending
moments within the pile in response to specified lateral pile head displacements of up to 30
mm and zero moment applied at the pile head. Modelling of the fixed head condition for the
HP360x132 piles for integral abutment consisted of predicting soil resistance over the pile
length and internal shear and bending moments within the pile in response to specified lateral
pile head displacements of up to 30 mm while maintaining zero rotation of the pile head.
Top of pile and final grade configurations used to define the LPile models are summarized in
Table 8-3 and were developed from the Preliminary Drawing for Option 1 in Appendix A.
Although piles are embedded into pile caps, the top of pile elevations were taken as the
underside of pile cap such that top of pile lateral deflections correspond to lateral movement
at the underside of the pile cap. A modulus of elasticity (E) of 200 GPa was used to define the
stiffness of steel piles.
Table 8-3: Top of Pile and Grade Elevation LPile Configurations
Pile Type Pile LocationElevation
Top of PileElevation
Top of Final Grade
610x12.7 Pipe
(Concrete Filled)Pier 232.9 m 231.9 m
HP360x132
(weak axis)
Integral abutment
(annular void to 3 m
below top of pile)
232.4 m1 234.4 m
1. LPile does not support pile heads below existing grade. In this regard, a top of pile elevation of 234.4 m was
used in LPile, and horizontal movement at elevation 232.4 m was estimated from the lateral deflection versus
depth plot at a depth of 2.0 m.
Graphical results summarizing lateral deflection of the pile top (bottom graph) and maximum
bending moment in the pile (top graph) as a function of applied static lateral load at the pile
head and zero applied moment are presented in Figure 3 for the concrete filled pipe piles, and
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Figure 4 for the H-piles. These charts can be used as design charts to estimate lateral
deflection at the underside of pile cap. At the request of Tetra-Tech, p-y curves and additional
graphical results output by LPile for the HP360x132 piles are included in Appendix E, and
summarize the following:
Page 1: p-y curve definitions
Page 2, Top Left Figure: deflection versus pile length
Page 2, Top Right Figure: bending moment versus pile length
Page 2, Middle Left Figure: shear force versus pile length
Page 2, Bottom Left Figure: mobilized pile stiffness versus pile length
Page 2, Bottom Right Figure: load intensity p versus lateral deflection
Amec Foster Wheeler recommends that all piles required to provide lateral resistance should
be embedded a minimum of 2 m into underlying till (i.e. elevation 227.0 m assuming till at
229.0 m) in order to maintain pile toe fixity. This minimum embedment depth shall be included
with the blow count criteria in developing pile acceptance criteria for pile required to provide
lateral resistance. Where this minimum embedment depth into till is not achieved, thenevaluation of pile specific monitoring logs should be undertaken to evaluate the lateral pile
resistance achieved and the impact on foundation performance. This may include drilling of a
test holes next to the pile(s) to confirm soil conditions at the afflicted pile location(s).
8.2.4.4 Inclined (Battered) Piles
Where fill depths are limited to less than 2 m and final design grades remain sufficiently similar
to existing grades, such as is anticipated for this project, significant soil settlement relative to
pile movement is not anticipated at the abutments. In this regard, it is anticipated inclined piles
will be acceptable. Inclined piles may be designed using the recommendations outlined in
Section 8.2.1 to determine the axial capacity of the pile, and horizontal and vertical capacity
determined from the angle of inclination. Inclined piles should be sloped no shallower than
1H:4V.
8.2.5 Minimum Embedment Depth
The minimum required embedment depth of a pile shall be taken as the greater of the
following, as determined in accordance with the recommendation outlined in the previous
sections of this report:
The embedment depth required to provide the required compressive (downwardacting) load resistance (See Section 8.2.1).
The embedment depth required to provide the required tensile (uplift) load resistance,
either due to foundation loads or due to adfreeze forces and/or frost heave pressure
on the underside of foundation elements (See Sections 8.2.3 and 11.2).
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The embedment depth required to provide the required fixity for lateral loadresistance, inclusive of any additional embedment depth required to extend below
the depth of scouring (See Section 8.2.4).
Notwithstanding the minimum embedment depths required to attain the design compressive,
tensile, and lateral resistance requirements, individual pile capacities should be confirmed
during driving based on either WEAP correlations between capacity and blow count/driving
energy, or through PDA monitoring and CAPWAP interpretations.
8.3 Pile Group Effects
Based on review of the preliminary design bridge profiles provided by Tetra Tech (see
Appendix A), Amec Foster Wheeler understood abutments and piers (if required) will be
supported on two rows of driven steel piles.
Generally, piles will behave individually in compression (i.e. Group efficiency η = 1.0) when a
minimum centre-to-centre spacing of fives pile diameters is provided between adjacent piles,
and will behave individually laterally when the center-to-center spacing is greater than five pile
diameters in either the direction transverse to loading (side-by-side), or the direction parallel
to loading (in-line). However, for circumstances in which the final pile layout places piles closer
than the spacing outlined above, interaction between the piles could occur and should be
reviewed by Amec Foster Wheeler during detailed design.
Possible interactive effects include changes to the efficiency of individual piles forming the
pile group, as well as stress overlap. In addition, construction related issues such as possible
heave of piles during driving of adjacent piles become a factor with closely spaced piles, and
needs to be verified with suitable quality control during construction.
8.4 Artesian Groundwater Impact on Semi-Integral Abutments
Amec Foster Wheeler understands that annular voids will be provided around integral
abutment piles to mitigate stresses within the abutments in response to thermal expansion
and contraction of the bridge superstructure. It is further understood that this voids will extend
approximately 3 m below the underside of pile cap, where the underside of pile cap elevation
is approximately 232.9 m for the integral Bridge Option 1. This places the base of the void at
approximate elevation 229.9 m, well with the range of potential development of perpetual
groundwater flow along the piles due to artesian conditions within the underlying till and
bedrock. The potential for developing flowing groundwater conditions through the proposed
annular voids can be mitigated by placing a seal at the base void that will resist artesian
groundwater pressure acting on the underside of the seal. The seal material itself also has to
be resistant to piping.
Discussions during preliminary design meetings with Tetra Tech identified a concrete seal
bonded to the surface of the pile as well as the internal wall of the void medium, likely to
consist of CSP. The concrete seal should be designed to be a minimum of 600 mm thick, oradditional thickness as required such that the sum of the weight of the concrete seal plus the
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 25
bond stress between the seal and the steel pile is equal to 1.4 times the artesian pressure
acting on the underside of the tremie seal. The assumed weight of the concrete may be taken
as 22 kN/m3, and the bond stress between the concrete and steel pile may be taken as
250 kPa. The pressure acting on the underside of the seal should be determined in
accordance with Section 7.0.
Tension in the concrete seal due to bending moments induced by artesian pressure acting on
the underside of the seal should be checked by the structural engineer. The pile and annular
sleeve should be treated as points of support for the tremie seal, and the tremie seal should
be treated as an unreinforced concrete beam.
In accordance with Section 6.11.4.10 of the CHBDC, the following two limit states shall be
considered when settlement of the surrounding ground occurs relative to a pile:
a) ULS of the pile at the neutral plane; and
b) SLS deformations at the pile top
Further in accordance with Article 6.11.4.10 of the CHBDC, unfactored permanent loads
associated with life cycle settlement of the surrounding ground shall be used when predicting
the neutral plane location. Transient loads shall not be included in the prediction of the location
of the neutral plane or settlement.
For clarity, the term ‘downdrag load’ as it is used in the CHBDC is defined in this report as the
drag load transferred to foundation elements by the downward movement of the soil relative
to the foundation element (i.e. pile). It is the integration of negative shaft friction transferred to
the foundation over the zone where soil moves down relative to the foundation element.
Downdrag load does not reduce the geotechnical resistance of a pile; but it must be
considered in combination with structural loads in order to verify that the stress in the pile at
the neutral plane (i.e. the point where the stress in the pile is greatest) does not jeopardize
pile integrity.
With respect to driven steel piles for the bridge structure at this Site, steel piles driven to
practical refusal within the very dense underlying till or on bedrock will develop the majority of
their resistance (i.e. greater than 90 percent) in shaft friction through the till combined with
end-bearing at the pile toe. Final design grades should be reviewed for fill thickness; however,
where fills do not raise final grade above elevation 237 m, it is anticipated that the surcharge
loading due to fill placement will not induce significant settlement within the underlying till. In
this regard, steel piles driven to practical refusal in within the underlying till will not be
susceptible to downdrag (i.e. settlement). Rather, the piles will be subjected to drag load
induced by negative shaft friction over the length of pile in contact with the clay overburden
and overlying embankment fill, where present.
Conservatively, the underside of the drag load zone may be assumed at elevation 229.0 m
(i.e. near the top of the till), and the drag load can be determined using a negative unit shaft
friction of 50 kPa integrated over the length of pile in contact with soil above this elevation.
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 26
For pipe piles, negative shaft friction should only be applied over the exterior surface area of
the pile in contact with the soil. In the case of steel H piles, the surface area should include
the exterior sides of the two flanges plus twice the depth of the web.
10.1.1 Earth Pressure Coefficients
The determination of lateral earth pressures will be required for the design of abutment
wingwalls and other substructures. Table 10-1 provides recommended earth pressure
coefficients for the active, passive and “at rest” earth pressure cases, and total unit weights
for various soil backfill types assuming horizontal grades and a vertical wall. The earth
pressure coefficients should be reviewed during detailed design for sloping grades and wall
faces.
Table 10-1: Earth Pressure Coefficients and Soil Unit Weights
Soil Type
Active
Pressure
Coefficient
Ka
“At Rest”
Pressure
Coefficient
Ko
Passive
Pressure
Coefficient
Kp
Total
Soil
Unit
Weight
(kN/m3)
Friction
Angle
(deg)
Between
Soil and
Concrete
Granular
Fill
Well Compacted (35°) 0.27 0.431 2.46 22 23
Moderately Compacted (30°) 0.33 0.501 2.00 21 20
Cohesive
Fill
Well Compacted (20°) 0.49 0.661 1.36 18 16
Moderately Compacted (15°) 0.59 0.741 1.13 17 12
1. In the case of unyielding walls exposed to frost penetration above the groundwater table, it is recommended that Ko = 1.0,
be used to account for lateral frost pressures1
The passive earth pressure coefficients provided in Table 10-1 include a reduction factor of
1.5 to account for the partial mobilization of passive resistance that is consistent with the smallwall displacements expected under operational conditions. Relatively large wall
displacements would be necessary to realize full passive resistances. To determine the
factored resistance, the resistance factor (Φ) of 0.5 should be applied to the passive earth
pressure.
Where sub-drainage will not be provided behind a wall, buoyant soil unit weights should be
used, and a hydrostatic pressure component will need to be included in the design. Buoyant
soil unit weights are determined by subtracting the unit weight of water (10 kN/m3) from the
given total unit weights. The recommended design groundwater level may be taken as
elevation 235 m.
1 As per Canadian Foundation Engineering Manual, 3rd Edition, P. 429, an earth pressure coefficient K=1 should be used incombination with insulation for highly frost susceptible soils.
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 27
The “at rest” (Ko) earth pressure should be used in the case of unyielding walls. To attain
active earth pressure (Ka) conditions, the displacement at the top of a cantilevered wall should
be at least 0.01 times the height of the wall. In the case of unyielding walls exposed to frost
penetration above the groundwater table, it is recommended that Ko = 1.0, be used to account
for lateral frost pressures. However, where lateral frost pressures need to be considered in
the design, the site-specific configurations of the walls or sub-structures should be reviewed
by qualified geotechnical personnel to explore alternatives in reducing the frost pressures
10.1.2 Calculation of Earth Pressure Distribution and Surcharge Loads
The magnitude and distribution of the lateral earth pressures on below grade structures will
depend on such factors as the rigidity of the below grade structure; the degree of compaction
of the backfill against the structure; the backfill soil type; the slope angle at the structure/soil
interface; and the subsurface drainage and groundwater conditions over the height of the
structure. It is anticipated that a sloped excavation will be implemented for construction of
below grade foundation structures, which will necessitate the placement of backfill behindbelow grade structure walls. The magnitude and distribution of the lateral earth pressures (P)
on below grade structures will depend on the degree of compaction of the backfill. In addition
to earth pressures, lateral stresses generated by any applicable surcharge loads also need to
be evaluated in the design. Recommended earth pressure distributions for light to moderate
and moderate to well compacted backfill cases, as well as for line or point surcharge loads,
are discussed in the following sections. The recommended earth pressure distributions are
for preliminary design only, and their applicability should be reviewed during detailed design.
10.1.2.1 Moderate to Well Compacted Backfill Case
Where subgrade support on the surface of the retained soil behind a wall is required, the
backfill against the wall will need to be compacted to at least 95 percent Standard Proctor
maximum dry density. In this case, the design earth pressure distribution should adopt a
combined trapezoidal/triangular distribution as shown in Figure 5 attached to account for the
induced lateral pressures due to compaction. Figure 5 also provides the relationships to be
used in the calculation of the compaction induced earth pressures, and tabulated loads (P)
generated by typical compactors. The earth pressure coefficients to be used in the calculation
of the lateral pressures should be those applicable to the backfill types given in Table 10-1.
10.1.2.2 Surcharge Loads
In addition to earth pressures, lateral stresses generated by surcharge loads, such as point
loads from heavy trucks, also need to be evaluated in the design. For line or point surcharge
loads, the lateral pressures should be determined using the relationships given in Figure 6. Inthe case of uniformly distributed surcharge loads, such as those due to the fluid contents
beneath a tank base (for design of the concrete ring beam), or those acting on the surface of
the retained soil, the induced lateral earth pressure may be determined by multiplying the
surcharge load by the appropriate earth pressure coefficient.
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 28
10.1.3 Load Factors
For the Limit States Design procedure for walls, the following Load Factors should be applied
to loads calculated from the pressure distributions given above.
For earth loads acting on walls, a Load Factor of 1.25 is recommended for sustainedloads.
For hydrostatic loads acting on walls, a Load Factor of 1.1 is recommended.
For live surcharge loads acting on walls, the Load Factor of 1.5 should be used.
The above load factors should be applied to loads lead to instability of the walls.
11.1 Frost Penetration Depth
The upper stratigraphy at the Site is considered moderately to highly frost susceptible in the
presence of a free supply of water, and as such, frost effects should be considered for
foundations or surface structures sensitive to movement. Based on historical temperature
data for the area, a design frost penetration of 2.4 m below final grade is recommended in
areas that will not have regular snow or vegetative ground cover. It should be noted that this
recommended frost penetration depth extends both vertically and laterally behind final surface
(i.e. extends 2.4 m behind the headwall).
11.2 Pile Foundations
Frost forces applied to pile foundations include adfreeze pressures acting along the pile shafts
within the depth of frost penetration. If pile caps are used and extend beyond the perimeter of
the underlying pile, then frost heave forces acting on the undersides of the pile caps, as well
as any connecting supports (i.e. lateral tie between the piles) will also need to be considered.
11.2.1 Frost Heave
To reduce the potential of frost heave pressures, a void-forming product should be installed
beneath the underside of the pile caps and any other structural element located within the
depth of frost penetration above the groundwater table. The recommended minimum
thickness of the void should be 150 mm. Alternatively, a compressible material may be used
in lieu of a void forming material, and the uplift pressures may be taken as the crushing
strength of the compressible medium. It is recommended that a frost heave of 150 mm be
assumed in determining the required thickness for the void-filler and the associated uplift
pressures associated with the thickness used.
The finished grade adjacent to each pile cap should be capped with well compacted clay and
sloped away so that the surface runoff is not allowed to infiltrate and collect in the void space
or saturate the compressible medium. If the soil layer within which the underside of the pile
cap is located is free draining (i.e. sand or gravel) such that infiltration cannot accumulate
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 29
within a void or saturation of the compressible medium, then a clay cap at finished grade may
not be required.
The use of void-forming product below the groundwater is unfeasible. In instances where
groundwater is located within the recommended depth of frost penetration, the underside of
foundation elements such as pile caps should extend below the depth of frost penetration to
mitigate frost heave development on the underside of the foundation element.
11.2.2 Adfreeze Stresses
Resistance to adfreeze and frost heave forces will be provided by the sustained vertical loads
on the foundation, the buoyant weight of the foundation and dead weight of the structure, and
the soil uplift resistance component provided by the length of the piles extending below the
depth of frost penetration. In the case of driven steel piles, the adfreeze force acting on the
pile may be determined assuming an unfactored unit adfreeze stress of 65 kPa applied to the
exterior surface of the pile and supported foundation elements (i.e. pile caps) located within
the zone of frost penetration. A load factor of 1.25 should be applied to obtain the factoredadfreeze stress.
Adfreeze stresses along the sides of pile caps and buried substructures can be reduced by
the installation of a ‘bond-break’ or ‘friction reducer’ within the zone of frost penetration.
Friction reducers could consist of a system of poly wrapped sono-tubes. A smooth
geosynthetic liner material, fixed to the shaft of the pile or to the sides of the pile cap would
also be a suitable bond-break.
In the case of straight shaft piles supporting lightly loaded unheated facilities, the piles should
be embedded a minimum of 7 m below final grade in order to provide sufficient frictional
resistance against potential uplift due to adfreeze stresses.
12.1 Design Philosophy
Slope stability analyses were completed for the proposed 4H:1V headslope configuration for
Bridge Option 1 and the proposed 5H:1V headslope configuration for Bridge Option 2. Slope
stability analyses were also completed for three cross-sections (Cross Section A, B, C)
selected by Tetra-Tech through the proposed grouted slope protection to be constructed
between the proposed bridge alignment and the existing railway. Specifically, slope stability
analyses of the headslopes and grouted slope protection were completed to determine the
Factor of Safety (FS) against rotational failure to satisfy minimum target factor of safety levels
for both long term conditions (i.e., when all construction induced pore water pressures are
fully dissipated), and short term conditions (i.e. a dewatered creek as is expected forconstruction of slope protection works). Minimum factor of safety targets are discussed in
Section 12.1.1.
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 30
12.1.1 Design Criteria
Limit equilibrium stability analyses were performed to confirm that; in keeping with the
standard local practice and city of Winnipeg Water Way permit requirements; the following
minimum target factor of safety would be achieved:
A minimum factor of safety of 1.5 for long term stability under ‘normal’ seasonalgroundwater and creek levels; and
A minimum factor of safety of 1.3 for short term stability under ‘short term’ conditions.Examples of short term conditions include spring flooding followed by rapid drawdown
to normal summer creek level, and temporary dewatering of the creek for construction.
Creek levels and groundwater conditions are discussed in Section 12.1.4.
12.1.2 Methodology and Model Geometry
All slope stability analyses were conducted using SLOPE/W, a limit equilibrium software
package developed by Geo-Slope International. Slope stability models for each of the two
proposed headslope configurations and each of the three cross-sections taken through the
grouted slope protection are shown in Appendix F. The models were developed by Amec
Foster Wheeler from the elevations and the bridge and slope configurations shown in the
preliminary plans provided by Tetra Tech, included in Appendix A.
12.1.3 Soil Stratigraphy and Soil parameters
Soil conditions along the project site were discussed in Section 4.0. In summary, Amec Foster
Wheeler idealized the stratigraphy into individual layers by soil type, as illustrated in the slope
stability outputs included in Appendix F. Table 12-1 summarizes the material strengthparameters and unit weights assigned for each of the soil types/layers. The parameters
presented in Table 12-1are based on Amec Foster Wheeler’s previous experience with similar
soils in the Winnipeg area. Other than erosion and scour failures south of the existing bridge
structure, there were limited signs of any slope movement. In this regard, post peak shear
strengths have been assumed for all clay soils, which is in keeping with accepted practice
The Morgenstern-Price method with a half sine variation of inter-slice forces was used for all
analyses.
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 31
Table 12-1: Summary of Slope Stability Material Parameters
LayerUnit Weight,
γ (kN/m3)
Pore Pressure
Coefficient ru
Effective Stress Parameters
Cohesion, c’
(kPa)
Friction Angle,
ϕ’ (°)
Clay Fill 17.5 NA 2 20
Native Brown Clay 17.5 NA 5 15
Native Grey Clay 17.5 NA 5 15
Glacial Silt Till 20 NA 1 35
Rip-Rap (Unbound
Aggregate)20 NA 0 35
Grouted Slope
Protection (Cracked)23 NA 0 35
12.1.4 Piezometric Conditions and Creek Levels
Groundwater conditions observed at the test hole locations is discussed in Section 4.3. Based
on the vibrating wire measurements obtained at the site, groundwater levels within the till and
clay overburden are considered to be governed by artesian pressure within the underlying
bedrock. In summary, porewater pressures were modelled using a phreatic surface; applied
to each soil layers; defined using Cartesian coordinates. Based on the VW measurements, a
phreatic surface of 234.4 m was assigned to the till stratum, and was assumed constant
throughout the year. Porewater pressure with the clay and clay fill were modelled based on
the VW measurement, and were defined using a minimum phreatic surface elevation taken
as equal to the water level in the creek, and phreatic surface elevations above the creek level
taken as the lower elevation given by the following:
Elevation 233.8 m, or
Underside of rip-rap or grouted slope protection.
Porewater pressure within the clay fill and clay established from the summer VW readings
were conservatively assumed to remain constant throughout the year. Notwithstanding,
groundwater monitoring data should continue to be collected and reviewed during the detailed
design phase to confirm that assumed groundwater conditions.
With respect to water levels within the Creek, the preliminary design drawings provided byTetra Tech (Appendix A) outlined a Q50% (i.e. 1 in 2 year) water level of 233.370, and a Q1%
(i.e. 1 in 100 year) water level of 234.520. The existing water level surveyed on 22 June 2016
was approximately 232.67 m. The Q50% level defines the Creek levels for the ‘Normal’
groundwater and creek configuration; while the Q1% defines the creek level for the 1in 100
year flood condition. A dewatered creek defines the ‘Extreme’ model condition.
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 32
12.2 Slope Stability Results
Table 12-2 presents the resulting Factor of Safety (FS’s) results for ‘normal’ for long term
stability of the headslopes and cross-sections taken through the proposed grouted slope
protection discussed in Section 12.1.2.
Table 12-2: Summary of Slope Stability Results for Overburden Phreatic Surface
Elevation 233.8 m, and Artesian Total Head of 234.4 m in the underlying Till
Model and Slope
Configuration
General Location of
Potential Slip
Surface (PSS)
Critical Factor of Safety (FOS) for Various Creek Levels
Empty
Channel
232.67
(22 June 2016)
233.37
(Q50)
234.52
(Q1)
(Extreme
Condition)
(Existing
Condition)
(Normal
Condition)
(Flood
Condition)
Bridge Option 1 –
5H:1V Headslope
Configuration
All PSS 1.31 1.69 1.87 2.20
Bridge Option 2 –
4H:1V Headslope
Configuration
All PSS 1.11 1.45 1.62 1.93
Grouted Slope
Protection – Section AAll PSS 1.01 1.90 NC NC
Grouted Slope
Protection – Section B
PSS originating at
top of railway –
Composite Slope
1.13 1.22 1.34 1.44
PSS originating at
crest of rip-rap /
base of rail
embankment
1.36 1.54 1.81 2.18
Grouted Slope
Protection – Section C
PSS originating at
top of railway –
Composite Slope
1.06 1.13 1.20 1.28
PSS originating at
crest of rip-rap /
base of rail
embankment
1.16 1.32 1.47 1.75
Design Criteria (Minimum FOS Requirement) 1.3 N/A 1.5 1.3
NC = not completed. FOS result will be greater than 1.5 by inspection.
12.2.1 Headslope Stability
With respect to Bridge Option 1 and a 5H:1V headslope configuration, slope stability results
outlined in Table 12-2 indicate that a 5H:1V headslope configuration will achieve the design
criteria and minimum factor of safety requirements set forth in Section 12.1.1 withoutrequirement for additional slope improvement (or stabilization) measures; provided that the
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
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headslopes and full width of the channel bottom are protected with 675 mm thick rip-rap
erosion protection.
With respect Bridge Option 2 and a 4H:1V headslope configuration, slope stability results
outlined in Table 12-2 indicate that the factor of safety of a 4H:1V headslope configuration is
highly susceptible to fluctuation in the water level of Sturgeon Creek. In fact, factor of safety
results are less than 1.5 for the water level surveyed on the 22 June 2016. In this regard, use
of a 4H:1V headslope configuration could necessitate additional slope improvement (or
stabilization) measures. Shear keys and/or stone columns comprise the slope improvement
measures most commonly employed in Winnipeg. Critical slip surfaces illustrated within the
slope stability outputs in Appendix F are deep seated and extend along the top of the till. In
this regard, shear keys and stone columns would have to extend to slightly below the surface
of the till in order to intersect potential slip surface along the surface of the till and achieve
minimum factors of safety targets. However, given artesian conditions within the underlying
till and the associated risk of developing a flowing artesian conditions and/or basal instability(See Section 6.3), the use of shear keys and/or stone columns is not recommended at this
site. If required, alternative slope improvement measures that could be considered through
detailed design include anchored sheet pile walls embedded to toe fixity in the underlying till.
Based on the discussion above, and giving consideration to both design and construction
costs as well as potential construction risks associated with artesian conditions within the
underlying till, unreinforced headslopes steeper than 5H:1V are not recommended.
12.2.2 Grouted Rip Rap Slopes
Due to the complexity of transitioning slopes over short distances as confined to the space
between the existing railway embankment and the proposed bridge structure, 3D modelling
of the stability of slopes between the structures is recommended for detailed design. 2D slope
stability analyses are not considered suitably representative of the real interaction between
transitioning slopes. Notwithstanding, 2D slope stability was undertaken to qualitatively
assess the range within which 3D slope stability results could be expected to fall.
Two dimensional slope stability analyses were undertaken on three cross-sections taken
through the proposed grouted rip-rap slopes. The proposed rip-rap configuration and the
locations of the cross-sections are shown on Tetra-Tech drawing SK3 included in Appendix
A. It should be noted that cross-sections B and C are composite sections, whereby the bottom
of the sections originate at the creek; the sections extend perpendicular with the slope to the
toe of the existing railway embankment (i.e. above the top of the proposed rip-rap and
approximate invert of the existing ditch running parallel with the railway); and then the sectionschange direction to extend up to the top of the existing railway embankment.
Slope stability analysis results are summarized in Table 12-2 for various Sturgeon Creek water
elevations. In summary, slope stability analyses indicate slip surfaces originating at the crest
of rip-rap / base of rail embankment (i.e. prior to the inflection in the composite section), are
stable under normal groundwater conditions and flood conditions. These same slip surfaces
are inherently unstable for the extreme dewatered creek condition expected to only occur
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
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17918 Geotechnical Report_Preliminary Design_final.docx Page 34
during construction. In this regard, slope stabilization measures will be required to maintain
slope stability during construction. It is envisaged analyses for detailed design will explore
anchored sheet piling installed at the toe of the slopes (i.e. at the crest of the creek) to fixity
in the underlying till to intersect slip surfaces extending into the creek bed.
Notwithstanding the stability of slip surface originating at the crest of rip-rap / base of rail
embankment, the slope stability result in Table 12-2 for composite Section B does not meet
the target FOS for the ‘normal’ creek level. Furthermore, the slope stability results in Table
12-2 for composite Section C do not meet the target FOS values for either the ‘normal’ or
‘flood’ creek levels. Notwithstanding, 2D representation of a three dimensional cross-sections
is not an accurate representation of real stress conditions. Due to the complexity of
transitioning slopes over short distances as confined to the space between the existing railway
embankment and the proposed bridge structure, 3D slope stability modelling is recommended
for detailed design. It is anticipated that 3D slope modelling, combined with implementation
of anchored sheet piling at the toe of the slopes to support temporary dewatering of the creek,will support the proposed Grouted Rip Rap Slope configuration.
12.2.3 Cofferdam and Creek Dewatering (Short Term Stability)
Amec Foster Wheeler understood Tetra-Tech proposes to dewater the creek (using coffer
dams) in order to evaluate the existing structure between the existing railway and the existing
Saskatchewan Avenue Bridge; as well as to undertake slope works required for the final
bridge design. It should be noted that dewatering of the creek, and in particular removing soil
in the creek, will negatively impact slope stability and potentially jeopardize the existing railway
embankment.
The factor of safety of the proposed headslopes and proposed rip-rap configuration for a
dewatered creek condition; assuming no further subexcavation of the creek bottom or slopes;
may be taken as the results outlined in Table 12-2 for the “Empty Channel” condition. In
summary, slope stability results indicate that sideslopes steeper than 5H:1V, including the
grouted rip-rap slopes, will all fail to meet the minimum target FOS of 1.3 for an ‘Extreme’
Condition given by a dewatered creek condition. In this regard, slope stabilization measures
will be required to meet the minimum target factor of safety for slope stability during
construction. It is envisaged analyses for detailed design will explore anchored sheet piling
installed at the toe of the slopes (i.e. at the crest of the creek) to fixity in the underlying till to
intersect slip surfaces extending into the creek bed.
Slope stability analyses to satisfy short term construction activities (i.e. dewatering of the creek
and subexcavation of the creek bed) shall by undertaken during detailed design once the finalbridge alternative and temporary construction requirements have been determined. Due to
the complexity of transitioning slopes over short distances as confined to the space between
the existing railway embankment and the proposed bridge structure, 3D slope stability
modelling is recommended for detailed design.
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 35
12.3 Slope Stability Conclusions and Recommendations for Detailed Design
Based on the slope stability analyses and results discussed in Sections 12.1 and 12.2,
conclusions and recommendations for detailed design are as follows:
The stability of both existing slopes and proposed design slopes is very sensitive tochanges in the water level of Sturgeon Creek.
A 5H:1V headslope configuration will achieve the design criteria and minimum factorof safety requirements set forth in Section 12.1.1 for all potential creek levels,
including a dewatered creek condition (neglecting basal stability issues due to
artesian pressure).
Although 4H:1V headslopes are marginally stable for the ‘normal’ condition definedby a Creek Level of 233.37 m, the Factor of Safety for the existing creek level of
232.67 m at the time of this investigation falls below the minimum target of 1.5 for the
normal condition. Based on interpolation of the FOS results for Creek Levels 232.67
and 233.37, a minimum creek level of about 232.9 m would need to be assured to
meet the minimum target FOS of 1.5 for ‘Normal’ Conditions. Furthermore, 4H:1V
headslopes do not achieve the minimum target FOS of 1.3 for the dewatered creek
condition.
Giving consideration to bullets 2 and 3 above, as well as consideration to design andconstruction costs as well as potential construction risks associated with artesian
conditions within the underlying till, Amec Foster Wheeler recommends a maximum
(i.e. steepest) headslope configuration of 5H:1V.
Excluding concerns related to basal heave discussed in Sections 6.3 and 7.0,dewatering of Sturgeon Creek will temporarily reduce the stability of both existing
slopes and proposed construction slopes to less than the minimum recommended
target of 1.3. In this regard, slope stabilization of will be required for the construction
condition (dewatered creek).
It is envisaged slope stabilization considerations for detailed design will exploreanchored sheet piling installed at the toe of the slopes (i.e. at the crest of the creek)
in order to intersect potential slip surface extending into the creek bed. Specifically,
the sheet piles will be embedded to fixity in the underlying till, and if required,
anchored near the top in order to mitigate lateral deflection of the sheet pile. Due to
artesian pressure within the underlying till and the risk of basal heave and
development of a flowing artesian condition, construction of conventional shear keys
or rock filled caissons to stabilize slopes is not recommended unless
depressurization of the aquifer were undertaken.
Due to the complexity of transitioning slopes over short distances as confined to the
space between the existing railway embankment and the proposed bridge structure,
3D slope stability modelling is recommended for detailed design of the grouted rip-rap configuration. It is anticipated that 3D slope modelling, combined with
implementation of anchored sheet piling at the toe of the slopes to support temporary
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 36
dewatering of the creek, will support the proposed Grouted Rip Rap Slope
configuration.
Where concrete elements outlined in this report and all other concrete in contact with the local
soil will be subjected in service to weathering, sulphate attack, a corrosive environment, or
saturated conditions, the concrete should be designed, specified, and constructed in
accordance with concrete exposure classifications outlined in the latest edition of CSA
standard A23.1, Concrete Materials and Methods of Concrete Construction. In addition, all
concrete must be supplied in accordance with current Manitoba and National Building Code
requirements.
Based on significant data gathered through previous work in southern Manitoba, water soluble
sulphate concentrations in the soil are typically in the range of 0.2% to 2.0%. As such, the
degree of sulphate exposure at the site may be considered as ‘severe’ in accordance with
current CSA standards, and the use of sulphate resistance cement (Type HS or HSb) isrecommended for concrete in contact with the local soil. Furthermore, air entrainment should
be incorporated into any concrete elements that are exposed to freeze-thaw to enhance its
durability.
It should be recognized that there may be structural and other considerations, which may
necessitate additional requirements for subsurface concrete mix design.
14.1 Pavement Design Methodology
Asphalt Concrete Pavement designs for Manitoba Avenue have been developed using the
1993 AASHTO Guide for Design of Pavement Structures.
14.2 Design Vehicle and Traffic
Based on discussions with Tetra Tech, Amec Foster Wheeler understood that pavement
design were to be developed assuming an initial two-way average annual daily traffic (AADT)
of 4700 vehicles per day. In terms of reducing the Two-Way AADT to a single lane design
volume, further discussions with Tetra Tech indicated a direction distribution factor of 0.5 and
a lane distribution factor of 1.0.
In terms of design vehicles to determine pavement loading, discussions with Tetra Tech
indicated truck volumes shall be taken as 2.5% of the AADT, and that the design truck shall
consist of a 5-axle single steer tandem truck tractor/tandem trailer combination (or 3S2 truck)
with a gross vehicle weight of about 40,000 kg (88,185 lbs). Using axle load factors from
AASHTO, such a truck has a single pass ESAL equivalency of about 3.56 ESALs per pass
for ACP. The remaining 97.5% of traffic volume was assumed to be passenger vehicles with
a gross vehicle weight of about 1,800 kg (4,000 lbs).
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 37
14.3 Subgrade Resilient Modulus
Pavement sections have been designed using a nominal effective subgrade resilient modulus
(Mr) of about 30 MPa (4500 psi), or an approximately equivalent California Bearing Ratio of
3.0 percent. The above subgrade resilient modulus has been determined for the soil
conditions observed at the test hole locations, and has been calculated accounting for the
climatic region and associated seasonal variations in subgrade strength using the
serviceability method outlined in Section 2.3.1 of the 1993 AASHTO Guide for Design of
Pavement Structures.
14.4 Granular Base Course and Subbase Course Materials
Table 14-1 summarizes key parameters for granular base course and granular subbase
course materials used in pavement design. Granular base course and granular subbase
course materials were selected in accordance with locally available aggregates. Resilient
moduli as shown were selected based on correlation with layer coefficients commonly used
in local practice and design of pavements using the 1993 AASHTO Guide for Design ofPavement Structures. Anticipated maximum dry unit weights and optimum moisture contents
have been specified based on Amec Foster Wheeler Winnipeg’s database of maximum dry
density results.
Table 14-1: Summary of Granular Pavement Structure Materials
Layer Material Resilient Modulus(MPa / psi)
Maximum DryUnit Weight(kg/m3 / pcf)
Optimum MoistureContent (%)
Granular BaseCourse &
Drainable StableBase
20 mm minuscrushed limestone 207 / 30,000 2,200 / 137 8.3
Granular SubbaseCourse Layer 1
50 mm maxcrushed limestone 172 / 25,000 2,300 / 144 6.3
Granular SubbaseCourse Layer 1
150 mm minuslimestone 172 / 25,000 2,000 / 125 6.3
14.5 Subgrade Preparation
Typical construction recommendations for subgrade preparation and pavement construction
provided locally are as follows:
1. Excavate to design subgrade elevation, which should be taken as the top of
the pavement minus the pavement thickness and the recommended
minimum base course and subbase gravel structure for the specified
pavement structure. Further excavation should be conducted as required to
remove organic or otherwise unsuitable soils.
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 38
2. Stripping and excavation to subgrade design elevation should be completed
in such a manner as to minimize disturbance of the subgrade. In this regard,
Amec Foster Wheeler recommends that excavation be completed using a
backhoe equipped with a smooth bladed bucket operating from the edge of
the excavation. Further, no construction equipment should be allowed on the
exposed subgrade until an assessment of the subgrade has been completed
by knowledgeable and experienced geotechnical personnel.
3. Once the final subgrade elevation has been achieved, an assessment of the
subgrade shall be completed in order to identify any localized loose, ‘weak’,
or soft areas prior to trafficking the subgrade and/or prior to fill operations.
Ground conditions permitting, assessment of the subgrade should consist of
proof-rolling the subgrade with multiple passes of a fully loaded tandem.
Notwithstanding, the ability of a subgrade to support proof-roll loads is
subject to change throughout construction as a result of changing moistureconditions, and in this regard, proof-rolling may not be possible. The exposed
subgrade and feasibility of proof-rolling of the subgrade should be visually
evaluated by qualified geotechnical personnel throughout stripping and
subgrade preparation operations.
4. Loose, ‘weak’, or soft areas identified either visually or by proof-rolling should
be sub-excavated below design subgrade as required to achieve a
competent subgrade stratum up to a maximum of 400 mm below grade, and
replaced with engineered fill material, as directed by the engineer at the time
of construction. Where silt remains at the subgrade elevation, specific
backfilling methods and procedures and use of select fill materials (such as
100 mm down crushed limestone underlain by a geotextile separator) may
be required.
5. Protect the exposed subgrade from frost, desiccation (drying), and
inundation (wetting) both during and following construction. To reduce
accumulation of surface runoff and softening of the subgrade, rough grades
should be designed to reduce the potential for ponding of water on the
surface and to provide positive drainage towards the perimeter of the
subgrade area and/or collection areas as quickly as possible, both during
and following subgrade preparation.
6. Depending on disturbance and protection of the subgrade, exposed
subgrades that are highly disturbed (i.e. rutted), or desiccated or inundated
outside of the acceptable range of the optimum moisture content (i.e. more
than 3 percent below or 3 percent above OMC), should be re-conditioned
and re-compacted prior to fill placement. If required, re-conditioning of the
subgrade should consist of scarifying the subgrade to a minimum of 200 mm
below grade, moisture conditioning dried or wetted soil to between OMC and
3 percent above OMC, and compacted to a minimum of 95 percent of
standard Proctor maximum dry density (SPMDD). If excavation to subgrade
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 39
minimizes disturbance of the subgrade and the subgrade is stable and within
an acceptable moisture state, then scarification and re-compaction of the
subgrade is not required.
7. Fill materials, if required between the subgrade elevation and the underside
of the pavement structure, should consist of additional granular subbase. The
fill material should be placed in 150 mm thick lifts and uniformly compacted
to 98% of SPMDD. Alternatively, suitable approved clay fill could be used,
provided that it is free of deleterious materials, moisture conditioned to wet
of optimum (preferably two to five percent wet of optimum) and compacted
uniformly to 98 % of SPMDD.
8. The 50 mm minus granular subbase should be placed in maximum 200 mm
thick lifts (or reduced lift thicknesses as governed by the compactive abilities
of the compaction equipment) and uniformly compacted to a minimum of 98
percent of SPMDD at ± 3 percent of OMC to the bottom of the base coursedesign elevation.
9. The granular base course should be placed in maximum 200 mm thick lifts
(or reduced lift thicknesses as governed by the compactive abilities of the
compaction equipment) and uniformly compacted to a minimum 100 percent
of SPMDD at ± 3 percent of OMC to the bottom of the asphalt design
elevation.
10. Qualified geotechnical personnel should monitor the quality and placement
of gravel and the compaction of the gravel should be monitored by field
density testing at regular frequencies. The density of each lift should be
tested to confirm that adequate compaction has been achieved before
placing the next lift.
11. Asphalt should be compacted to a minimum 98 percent of a 75 blow Marshall
Density.
14.6 Asphalt Concrete Pavement (ACP) Alternative
Table 14-2 presents a recommended HMA pavement sections for a Target Reliability level of
90. The first pavement section does not include any consideration to loss of serviceability level
resulting from swelling and freezing of the subgrade soils. The sections have been developed
using the design methodology, traffic loading, and subgrade and base course design
properties outlined in Sections 1.1 through 1.5. Based on discussions with Tetra Tech and
their experience with City of Winnipeg preferences for the Ness Avenue pavement structure,additional assumptions required for selection of other HMA pavement design inputs were as
follows:
Design Life: 25 years
Annual Traffic Growth: 0.5%
Design Serviceability Level: 4.2
Terminal Serviceability Level: 2.5
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 40
Standard Deviation: 0.45
Structural Layer Coefficient for Asphalt Concrete Pavement: 0.40
Structural Layer Coefficient for Base Course: 0.14
Structural Layer Coefficient for Granular Sub-base: 0.10
Table 14-2: ACP Pavement Alternative for 90% Reliability
Serviceability Loss Due to Swelling and Frost Heave Included? No YesLayer Name Thickness (mm) Thickness (mm)
Surface Asphalt Layer – PG58-40 50 50Asphalt Layer # 2 (Two-Lifts) – PG58-34 90 9520 mm Crushed Limestone Base Course 75 75
50 mm MAX Crushed Limestone Sub-base 150 15050 mm MAX, 100 mm minus, or 150 minus Crushed Sub-base 360 515
Non-Woven Geotextile Fabric (At Subgrade) Yes YesTotal Structure Depth 725 885Structural Number (SN) 120.1 137.7
The estimated loss of serviceability versus pavement age is presented in Figure 7, and has
been estimated based on an estimated swell rate constant of 0.11 for the anticipated subgrade
conditions, and an estimate potential vertical rise of about 46 mm (1.8 inches) over a 4.6 m
thick swelling zone (i.e. given by the depth of the active zone). From the figure, a serviceability
loss of 0.62 is observed over the duration of a 25 year design life. If loss of serviceability due
to swelling and frost heave is not considered over the design life of the pavement, than it isestimated that the terminal serviceability level of 2.5 would occur at about year 14 as opposed
to the 25 year design life. If swelling and frost heave is considered, than the thicker pavement
structure in Table 14-2 would be required to satisfy the target 25 year design life.
All engineering design recommendations presented in this report are based on the
assumption that an adequate level of testing and monitoring will be provided during
construction and that all construction will be carried out by a suitably qualified contractor
experienced in foundation and earthworks construction. An adequate level of testing and
monitoring is considered to be:
for earthworks: full-time monitoring and compaction testing.
for deep foundations: design review and full time monitoring during construction.
Project File No. WX17918 Amec Foster Wheeler
Preliminary Design Geotechnical Report Environment & Infrastructure
Saskatchewan Avenue over Sturgeon Creek Culvert Replacement
Winnipeg, Manitoba
30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 41
for concrete construction: testing of plastic and hardened concrete in accordancewith the latest editions of CSA A23.1 and A23.2; and
review of concrete supplier’s mix designs for conformance with prescribed and/or
performance concrete specifications.
Amec Foster Wheeler requests the opportunity to review the design drawings, and the
installation of the foundations, to confirm that the geotechnical recommendations have been
correctly interpreted. Amec Foster Wheeler would be pleased to provide any further
information that may be needed during design and to advise on the geotechnical aspects of
specifications for inclusion in contract documents.
The findings and recommendations presented in this report were based on geotechnical
evaluation of the subsurface conditions and limited groundwater data observed during the site
investigation described in this report and based on the bridge configurations provided to Amec
Foster Wheeler. If conditions other than those reported in this report are noted during
subsequent phases of the project, or if the assumptions stated herein are not in keeping with
the design, this office should be notified immediately in order that the recommendations can
be verified and revised as required. Recommendations presented herein may not be valid if
an adequate level of inspection is not provided during construction, or if relevant building coderequirements are not met.
Soil conditions, by their nature, can be highly variable across a site. The placement of fill and
prior construction activities on a site can contribute to the variability especially in near surface
soil conditions. A contingency should always be included in any construction budget to allow
for the possibility of variation in soil conditions, which may result in modification of the design
and construction procedures.
This report has been prepared for the exclusive use of Tetra Tech WEI Inc., and their agents,
for specific application to the project described in this report. The data and recommendations
provided herein should not be used for any other purpose, or by any other parties, without
review and written advice from Amec Foster Wheeler. Any use that a third party makes of this
report, or any reliance or decisions made based on this report, are the responsibility of those
parties. Amec Foster Wheeler accepts no responsibility for damages suffered by a third party
as a result of decisions made or actions based on this report.
Project File No. WX17918 Amec Foster Wheeler Preliminary Design Geotechnical Report Environment & Infrastructure Saskatchewan Avenue over Sturgeon Creek Culvert Replacement Winnipeg, Manitoba 30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 43
FIGURES
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20.0
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90.0
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Q=1800 kN, M=0 kN•m, Fixed Head Q=1800 kN, M=0 kN•m, Free Head
TETRA TECH WEI INC.PROJECT:
WX17918
DATE:
AUGUST 2016
DRAWING:
FIGURE 3
SASKATCHEWAN AVENUE OVER STURGEON CREEK LATERAL RESPONSE OF DN600x12.7mm PIPE PILES
CONCRETE FILLED - FIXED & FREE HEAD
0
20
40
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120
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180
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LATERAL RESPONSE OF HP360x132 INTEGRAL ABUTMENT PILES(Annular void to 3 m below underside of pile cap)
0.0
10.0
20.0
30.0
40.0
50.0
60.0
70.0
80.0
90.0
100.0
0 5 10 15 20 25 30 35
Max
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(kN
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Lateral Deflection at Underside of Pile Cap (mm)
Q=1630 kN, S= 0, Fixed Head Q=1630 kN, M= 0, Free Head
TETRA TECH WEI INC.PROJECT:
WX17918
DATE:
AUGUST 2016
DRAWING:
FIGURE 4
SASKATCHEWAN AVENUE OVER STURGEON CREEKLATERAL RESPONSE OF HP360x132 INTEGRAL ABUTMENT PILES
BENDING ABOUT WEAK AXIS - FIXED & FREE HEAD
FOR Zc <- Z d-<
FOR Z > d
P (ROLLER LOAD) = DEAD WT. OF ROLLER + CENTRIFUGAL FORCE
WIDTH OF ROLLER
(SEE TEXT OF REPORT)
= SOIL UNIT WEIGHT(SEE TEXT OF REPORT)
EARTH PRESSURE COEFFICIENTS
K = Ko ("AT REST") OR Ka (ACTIVE CASE)
TYPICAL COMPACTOR LOADS (P)
CompactorLOAD (P)
kN/mCompactor
LOAD (P)kN/m
Bowmag TSE 31 Bowmag BW122PD 36
Bowmag 60S 32 Bowmag 142PDB 47
Bowmag 65S 22 Bowmag 172PDB 93
Bowmag 75S 33 Dynapac LR100 42
Bowmag 90S 39 Dynapac 2100V 93
Bowmag 75AD 20 Dynapac CA121D 53
Bowmag 100AD 20 Dynapac CA121PD 54
Bowmag 120AD 34 Dynapac CA151 80
Bowmag 130AD 36 Dynapac CA151D 80
Bowmag BW122D 30 Dynapac CA151PD 96
EARTH PRESSURE DISTRIBUTION
TYPICAL VALUES GIVEN IN TABLE
SASKATCHEWAN AVENUE OVER STURGEON CREEKLATERAL EARTH PRESSURE INDUCED BY COMPACTION
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FIGURE 5
SEPTEMBER 2016
Amec Foster Wheeler
TETRA TECH WEI INC.DATE:
WX17718PROJECT No.:
PROJECT No.:
VA
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PRESSURES FROM LINE LOAD(BOUSSINESQ EQUATION MODIFIED BY EXPERIMENT)
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hP
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0.64 QL
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FOR m > 0.4:
LINE LOAD QL FOR m ≤ 0.4:
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SASKATCHEWAN AVENUE OVER STURGEON CREEKLATERAL PRESSURE DUE TO
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FIGURE 6
SEPTEMBER 2016
Amec Foster Wheeler
TETRA TECH WEI INC.DATE:
WX17718PROJECT No.:
PROJECT No.:
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201
6
Project File No. WX17918 Amec Foster Wheeler Preliminary Design Geotechnical Report Environment & Infrastructure Saskatchewan Avenue over Sturgeon Creek Culvert Replacement Winnipeg, Manitoba 30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 50
APPENDIX A
Project File No. WX17918 Amec Foster Wheeler Preliminary Design Geotechnical Report Environment & Infrastructure Saskatchewan Avenue over Sturgeon Creek Culvert Replacement Winnipeg, Manitoba 30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 51
APPENDIX B
Project File No. WX17918 Amec Foster Wheeler Preliminary Design Geotechnical Report Environment & Infrastructure Saskatchewan Avenue over Sturgeon Creek Culvert Replacement Winnipeg, Manitoba 30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 52
APPENDIX C
Photo 1: RW01 - Westbound lane on Saskatchewan Ave., ~1.0 m from curb
Photo 2: RW02 - Eastbound lane on Saskatchewan Ave., ~1.0 m from curb
C
Photo 3: RW03 - Westbound lane on Saskatchewan Ave., ~1.0 m from curb
Photo 4: RW04 - Eastbound lane on Saskatchewan Ave., ~1.0 m from curb
C
Photo 5: RW05 - Westbound lane on Saskatchewan Ave., ~1.0 m from curb
Photo 6: RW06 - Eastbound lane on Saskatchewan Ave., ~1.0 m from curb
C
Photo 7: RW07 - Westbound lane on Saskatchewan Ave., ~1.0 m from curb
Photo 8: RW08 - Eastbound lane on Saskatchewan Ave., ~1.0 m from curb
C
Photo 9: RW09 - Westbound lane on Saskatchewan Ave., ~1.0 m from curb
Photo 10: RW10 - Eastbound lane on Saskatchewan Ave., ~1.0 m from curb
C
Project File No. WX17918 Amec Foster Wheeler Preliminary Design Geotechnical Report Environment & Infrastructure Saskatchewan Avenue over Sturgeon Creek Culvert Replacement Winnipeg, Manitoba 30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 53
APPENDIX D
12.6
m14
.2m
9.6m
15.7
m
17.2
m
11.1
m9.
1m
18.7
m17
.2m
D
Project File No. WX17918 Amec Foster Wheeler Preliminary Design Geotechnical Report Environment & Infrastructure Saskatchewan Avenue over Sturgeon Creek Culvert Replacement Winnipeg, Manitoba 30 September 2016
17918 Geotechnical Report_Preliminary Design_final.docx Page 54
APPENDIX E
To
p o
f P
ileG
rad
ey,
m0.
0000
00.
0000
70.
0004
40.
0008
80.
0013
20.
0017
50.
0021
90.
0026
30.
0030
70.
0035
10.
0039
50.
0043
90.
0048
30.
0052
60.
0140
40.
0263
20.
0351
0p,
kN
/m0.
0000
00.
0000
00.
0000
00.
0000
00.
0000
00.
0000
00.
0000
00.
0000
00.
0000
00.
0000
00.
0000
00.
0000
00.
0000
00.
0000
00.
0000
00.
0000
00.
0000
0y,
m0.
0000
00.
0000
20.
0001
50.
0004
90.
0011
70.
0022
90.
0039
50.
0062
80.
0093
70.
0133
40.
0183
00.
0243
60.
0316
30.
0402
10.
0502
30.
0617
80.
0656
4p,
kN
/m0.
0000
06.
3180
012
.636
0018
.954
0025
.272
0031
.590
0037
.908
0044
.226
0050
.544
0056
.862
0063
.180
0069
.498
0075
.816
0082
.134
0088
.452
0094
.770
0094
.770
00y,
m0.
0000
00.
0000
20.
0001
50.
0004
90.
0011
70.
0022
90.
0039
50.
0062
80.
0093
70.
0133
40.
0183
00.
0243
60.
0316
30.
0402
10.
0502
30.
0617
80.
0656
4p,
kN
/m0.
0000
06.
3180
012
.636
0018
.954
0025
.272
0031
.590
0037
.908
0044
.226
0050
.544
0056
.862
0063
.180
0069
.498
0075
.816
0082
.134
0088
.452
0094
.770
0094
.770
00y,
m0.
0000
00.
0000
20.
0001
50.
0004
90.
0011
70.
0022
90.
0039
50.
0062
80.
0093
70.
0133
40.
0183
00.
0243
60.
0316
30.
0402
10.
0502
30.
0617
80.
0656
4p,
kN
/m0.
0000
06.
3180
012
.636
0018
.954
0025
.272
0031
.590
0037
.908
0044
.226
0050
.544
0056
.862
0063
.180
0069
.498
0075
.816
0082
.134
0088
.452
0094
.770
0094
.770
00y,
m0.
0000
00.
0009
40.
0018
70.
0028
10.
0037
40.
0046
80.
0056
10.
0065
50.
0074
80.
0084
20.
0093
50.
0102
90.
0112
20.
0121
60.
0130
90.
0140
30.
0149
6p,
kN
/m0.
0000
010
9.32
001
210.
7204
229
8.42
152
369.
8016
442
5.07
911
466.
2602
049
6.06
282
517.
1812
153
1.92
346
542.
1074
154
9.09
168
553.
8577
855
7.09
916
559.
2985
056
0.78
845
561.
7967
4y,
m0.
0000
00.
0009
50.
0018
90.
0028
40.
0037
80.
0047
30.
0056
70.
0066
20.
0075
60.
0085
10.
0094
50.
0104
00.
0113
40.
0122
90.
0132
40.
0141
80.
0151
3p,
kN
/m0.
0000
011
9.47
508
230.
2948
832
6.14
280
404.
1536
446
4.56
600
509.
5725
354
2.14
360
565.
2237
458
1.33
544
592.
4654
260
0.09
847
605.
3073
160
8.84
979
611.
2534
361
2.88
179
613.
9837
5y,
m0.
0000
00.
0011
10.
0022
20.
0033
30.
0044
40.
0055
50.
0066
60.
0077
70.
0088
80.
0099
90.
0111
00.
0122
10.
0133
20.
0144
40.
0155
50.
0166
60.
0177
7p,
kN
/m0.
0000
023
6.81
433
456.
4728
364
6.45
521
801.
0822
992
0.82
704
1010
.035
5210
74.5
9539
1120
.343
0811
52.2
7846
1174
.339
4411
89.4
6910
1199
.793
6612
06.8
1529
1211
.579
6012
14.8
0720
1216
.991
42y,
m0.
0000
00.
0011
90.
0023
90.
0035
80.
0047
80.
0059
70.
0071
60.
0083
60.
0095
50.
0107
40.
0119
40.
0131
30.
0143
30.
0155
20.
0167
10.
0179
10.
0191
0p,
kN
/m0.
0000
028
2.88
428
545.
2752
377
2.21
686
956.
9251
610
99.9
6510
1206
.528
2312
83.6
4760
1338
.295
0613
76.4
4316
1402
.795
8914
20.8
6888
1433
.201
9914
41.5
8961
1447
.280
7714
51.1
3626
1453
.745
41y,
m0.
0000
00.
0012
00.
0024
00.
0036
00.
0048
00.
0060
00.
0071
90.
0083
90.
0095
90.
0107
90.
0119
90.
0131
90.
0143
90.
0155
90.
0167
90.
0179
90.
0191
8p,
kN
/m0.
0000
031
2.53
396
602.
4266
485
3.15
448
1057
.222
4312
15.2
5468
1332
.986
9114
18.1
8932
1478
.564
4915
20.7
1097
1549
.825
7915
69.7
9304
1583
.418
8115
92.6
8555
1598
.973
2116
03.2
3281
1606
.115
439
11
68
79
810
46
57
5.4
7.4
Dep
th B
elo
w:
p-y
cu
rve
def
init
ion
0 to
<3
2 to
<7
35
0.00
00.
010
0.02
00.
030
0.04
00.
050
Load Intensity, p (kN/m)
Late
ral D
efle
ctio
n,y
(m)
p-y
CU
RV
ES
FO
R O
F H
P36
0x13
2 IN
TE
GR
AL
AB
UT
ME
NT
PIL
ES