Spring Thesis_ Final Report_REPORT ONLY_Blast Proof

8/20/2019 Spring Thesis_ Final Report_REPORT ONLY_Blast Proof

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Rebecca Dick

Structural Option

Dr. Memari

Spring 2012 Senior Thesis

April 4, 2012

Blast Design and Analysis

National Business Park- Building 300

300 Sentinel Drive

Annapolis Junction, MD, 20701


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Spring Thesis Rebecca Dick | Structural Option

Advisor: Dr. Memari Page | 2

Acknowledgments

A very special thank you Dr. John O. Hallquist, President, Livermore Software Technology Corporation

(LSTC), and Marsha J. Victory, President, FEA Information Inc, for providing me access to the LS-DYNA

finite element software for free for my senior thesis and to Gunther Blankenhorn and Todd P Slavik for

the invaluable technical support and troubleshooting they provided.Additional thanks to Michael Baker, Jr., Inc. for the use of National Business Park- Building 300,

especially Mr. Erik Spicker.

Special Thanks to:

My Family:

Thank you to my dad for always answering my questions for the last five years. Thank you to my mom

for listening to me complain all the time and putting up with my breakdowns. Thank you to my brother

for giving me academic competition.

Penn State Faculty:

Dr. Memari

Dr. Hanagan

Dr. Parfitt

Ryan Solnosky

My Friends:

Brian Rose

Dave Tran

Mike Kostick

Ryan Blatz

Liz Kimble

Rob Livorio

RJ Fazio

The past five years have been a blast and I couldn’t have made it through without you guys. You’ve kept

me as close to sane as any AE can be through all the crazy projects, late night study sessions, and post-

exam celebrations. I will always love you guys!


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Table of ContentsAcknowledgments ......................................................................................................................................... 2

Special Thanks to: ................................................................................................................................. 2

Executive Summary ....................................................................................................................................... 6

National Business Park- Building 300 ............................................................................................................ 8

Structural Background .................................................................................................................................. 9

Foundations .............................................................................................................................................. 9

Floors....................................................................................................................................................... 12

Load Path ............................................................................................................................................ 12

Columns .................................................................................................................................................. 13

Lateral System ......................................................................................................................................... 14

Load Path ............................................................................................................................................ 15

Impact of Lateral System on Calculations ........................................................................................... 16

Design Codes ............................................................................................................................................... 17

Material Properties ..................................................................................................................................... 18

Original Design ........................................................................................................................................ 18

Reinforcement: ................................................................................................................................... 18

Structural Steel: .................................................................................................................................. 18

Metal Deck: ......................................................................................................................................... 18

Concrete: ............................................................................................................................................. 18

Blast Redesign ......................................................................................................................................... 19

Reinforcement: ................................................................................................................................... 19

Structural Steel: .................................................................................................................................. 19

Metal Deck: ......................................................................................................................................... 19

Concrete: ............................................................................................................................................. 19

Gravity Loads............................................................................................................................................... 20

Dead Load ............................................................................................................................................... 20

Live Load ................................................................................................................................................. 20

Snow Load ............................................................................................................................................... 20

Structural Proposal ..................................................................................................................................... 21

Blast Design ................................................................................................................................................. 22

Procedure .................................................................................................................................................... 23


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Blast Load Determination ........................................................................................................................... 24

Column Design ............................................................................................................................................ 28

Beam Design ............................................................................................................................................... 30

Girder Design .............................................................................................................................................. 32

Slab Design .................................................................................................................................................. 33

Moment Connection Design ....................................................................................................................... 36

Final Design ................................................................................................................................................. 39

LS-DYNA Modeling ...................................................................................................................................... 40

Comparison of Results ................................................................................................................................ 44

Additional Comments and Conclusion ........................................................................................................ 48

Lateral System Implications .................................................................................................................... 48

Effects on Foundations ........................................................................................................................... 48

MAE Requirements ..................................................................................................................................... 49

Breadth Topic I: Site Redesign .................................................................................................................... 50

Breadth Topic II: Façade Redesign and Heat Transfer ................................................................................ 54

Façade Design ......................................................................................................................................... 55

Heat Transfer .......................................................................................................................................... 57

Cost Estimates ............................................................................................................................................. 61

Conclusion ................................................................................................................................................... 63


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Executive Summary

National Business Park- Building 300 is a seven story office building located in Annapolis Junction,

Maryland. It was designed in 2007 and construction was completed in 2009. The structure of NBP-300

is composed of a composite steel system and utilizes four eccentrically braced frames for the lateral

system located at the core of the building.

National Business Park- Building 300 was not designed to resist blast, and therefore, were an attack

made on the building, heavy structural damage would likely result. For this reason, it will be assumed

that NBP-300 will be redesigned as a high risk building for terroristic threats, and as such would be

required to meet certain criteria for blast loading. Designing for interior blast has many challenges, as

well as many solutions. There is no clear “cookie-cutter” method to designing for blast, which makes it

one of the most intriguing and interesting new topics of study. Blast design isn’t new per say, but in the

realm of structural design, it has only be in consideration for the last 50 years or so. Although

progressive collapse is a major issue when designing for blast, the scope of this redesign will be limitedto strength factors only due to time restraints, and redundancy of design will be ignored.

A specific situation has been created for the design. It can be assumed that the greatest threat in terms

of explosives will be a briefcase sizes device detonated in the interior of the lobby on the second floor.

The UFC -340-02 documents outlining blast requirements were used to identify threat levels, design site

security, and find blast loads. The blast load from this situation was calculated and a typical bay located

at the lobby location was designed to withstand the blast loading found. Additionally, the façade of

NBP-300 was redesigned to allow venting of the interior during the blast. The large percentage of glass

currently on the façade could potentially be harmful to occupants when the façade fails.

LS-DYNA Blast Modeling software was used to analyze a critical portion of the original and redesigned

structure. A comparative study was completed where the assumed explosive for this situation was

detonated in the original LS-DYNA model and the redesigned LS-DYNA model. The intent was to

minimize structural damage as much as economically possible through this redesign. By comparing the

original structure to the redesign it became apparent that the redesign was highly conservative, but

overall, withstood the blast load and incurred very little damage.

In addition to the blast design and analysis, a site redesign was completed. Also, a façade breadth was

completed, including a façade redesign with anchored “blowout panel” and a heat transfer study of the

new façade.

MAE coursework was incorporated into several aspects of the redesign for NBP-300, including building

modeling techniques (AE 597A), a moment connection design (AE534), and a façade design (AE542).


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Going into this thesis design, the following goals were set forth and achieved:

Design a typical beam, girder, column, moment connection, and floor system

Use a finite element analysis software (LS-DYNA Blast) to model the effects of blast in theoriginal structure and redesigned structure, and compare the results

Show that designs using hand calculations are overly conservative

Redesign the site of NBP-300 to mitigate large exterior blasts

Redesign the façade of NBP-300 to allow venting of the interior during an interior explosion

Calculate heat transfer through the new façade and determine if the new design is acceptable


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National Business Park- Building 300

National Business Park- Building 300 is an office building located in an industrial park in Annapolis

Junction, Maryland. Owned by COPT, the seven story office building, which has an additional

mechanical penthouse on the roof, was designed in 2007 and construction was completed in 2009. As

part of an industrial park, the architecture was intended to complement the existing offices in the

surrounding area. Shown below in Figures 1 and 2 is the building footprint of National Business Park-

Building 300 and a satellite view of the building, respectively.

NBP-300 is a 212,019 gross square foot, 116.92’ tall, composite steel framed building that utilizes braced

frames for its lateral system. Foundations consist of strip and spread footings, but since the first story is

only a partial floor plan, and is sub-grade at a few locations, footings are located below both the first

and second stories at their respective locations. The façade system is composed of a glass and precast

concrete curtain wall that is tied into the structural system at each floor level.

Figure 1: NBP-300 Footprint on Site Figure 2: Satellite View of NBP-300 provided by

Courtesy of Baker and Associates ©Google Maps, 2011


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Structural Background

Foundations

Since the first floor is only a partial floor and the grading on the site is such that the second floor is at

ground level at the main entrance, foundations occur below both the first and second floors at the

respective areas where no preceding floor exists below. Figure 3, on the following page, outlines the

area of the partial first floor. Figure 4, also on the following page, provides more detail by showing a

section cut in the North-South direction, which shows the partial first floor and the slab on grade and

foundations on the second floor. Additionally, two side of the first floor will be supported by basement

walls. These walls are reinforced with #7 bars at 10” on center, each face, vertically, and #5 bars at 12”

on center, each face, horizontally. These walls will see lateral pressure from soil, which was noted to be

60 psf per foot of depth in the structural notes.

The foundations of NBP-300 consist of strip and spread footings with a 5” slab-on-grade on the partial

first floor and on the second floor where it does not sit over the first floor, and were designed to meetthe suggestions set forth in the geotechnical report prepared by Hillis-Carnes Engineering Associates,

Inc. 6x6 W2.0xW2.0 Welded Wire Fabric was used in all slabs-on-grade. A bearing pressure of 3.5 ksi

was required for all foundations. 4” of compacted backfill meeting AASHTO 57 course aggregate was

also a requirement below the slabs-on-grade.

The strip and spread footings are located around the perimeter of the partial first floor and at the

perimeter locations on the second floor that contain the slab on grade. Additionally, interior columns

are supported by spread footings. These range in size from 8’ square to 19’ square, depending on the

location and the load. The depths of the spread foundations also vary between 24” to 46”. A

5’-0”x5’-0” footing at piers is typical, reinforced with #5@12” on center, each way.

Additionally, the mechanical room on the first floor is sunken 3’ below the standard floor level which

required the foundations to be stepped down at the transition locations.


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Figure 3: Outline of Partial First Floor Plan on Foundations, Courtesy of Baker and Associates

Figure 4: Building section showing partial first floor and slab on grade at second floor,

Courtesy of Baker and Associates


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Floors

NBP-300 was constructed using a composite steel system. For the floors, this entailed the use of

cambered composite beams and girders. A 3 ½” lightweight concrete topping (117 pcf) on 3”x20 gage

metal floor deck, bonding type, was used on each elevated floor for the composite slab, reinforced with

6x6 W2.0xW2.0 welded wire mesh. Additional reinforcement was required over filler beams and girders

along column lines. This additional reinforcement consists of #3 bars at 18” on center, 8’-0” long.

Composite action is obtained through the use of ¾” diameter, 5 ½” long shear studs spaced equally

along beams. See Figure 5, below, for a typical section cut through the composite floor system.

Floor beams vary in size and vary per floor, but are cambered 0”, ¾”, or 1”, depending on the location.

The exterior girders also vary in size, but are not cambered. Beams typically span 35’ at 10’ on-center,

while girders typically span 30” at 35’ on-center.

Load Path

Gravity loads on each floor are transferred from the composite slab to the beams. The load is

then transferred to the girders, which then transfer the load to the columns. The gravity loadsterminate at the foundations at the second level or at the partial first level.

Figure 5: Typical Composite Floor System,



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Columns

Wide flange columns are used throughout NBP-300. Although the weight of the columns differs

throughout the building and across floor levels, most wide flanges are W14 and all conform to ASTM A-

992 Grade 50 steel requirements. Splices occur at levels 4 and 6 for most columns. This information is

presented in a column schedule, of which a partial view can be seen in Figure 6, below. At the

penthouse level, square HSS posts are used around the perimeter of the enclosure to support the

penthouse cladding.

Base plates are used to connect the steel columns to the foundations or shear walls located on the first

floor. The base plates vary in size, but conform to ASTM A-36 type steel.

Figure 6: Partial Column Schedule,



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Lateral System

The main lateral system used in NBP-300 consists of four different eccentrically braced frames. These

occur at column lines “D”, “E”, “4”, and “11”, each spanning about one full bay (See Figure 9 on the

following page). This provides two lateral bracing frames in each direction. The frame at column line

“D” begins on the first floor, while the other three frames begin on the second floor, supported by

concrete shear walls on the first level. Two wind bracing frames are shown below, in Figures 7 and 8, to

show the difference in height. The shear walls at the base of the three shorter frames are 1’-2” thick

and are reinforced with #7 bars at 10” on center, each way on each face of the wall.

HSS tubes are used for the bracing, and vary in size from frame to frame, but conform to ASTM A-500

Grade C steel and have a yield strength of 46 ksi. All connections in the braced frames are slip-critical.

Shear studs at wind bracing beams occur at 12” on center, and are ¾” in diameter and 5 ½” in length. At

the base plates, an 8x4x3/4 angle at 1’-6” in length is used on each side of the gusset plate to resist

100% of the design shear. Additionally, 7/8” diameter by 8” long shear studs are used at the shear wall

connections for the three elevated braced frames.

Figures 7 and 8: Wind Braced Frames, Courtesy of Baker and Associates


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Figure 9: Location of Braced Frames at Core of building, Courtesy of Baker and Associates

Load Path

As wind lateral forces are applied to NBP-300, they are transferred from the façade to the

composite floor system through bearing connections. From there, the load is transferred from

the floor system to the four braced frames, which travel through the height of the building to

foundations and shear walls on the first and second floor. Where braced frames connect to

shear walls, the load is then resisted by the shear wall and ultimately transferred to the

foundations below.

Ground motion due to seismic loads is resisted by the foundations, first floor shear walls, andbraced frames that run the height of the building. As each floor is seismically loaded, the force

is transferred to the braced frames, which transfers the load to the foundations and shear walls

at the base of the building.


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Impact of Lateral System on Calculations

The lateral system used in the design of NBP-300 consists of four eccentrically braced frames. In

ASCE 7-05, this system provides a Response Modification Coefficient of 7, but since the

structural documents provided for NBP-300 specifically state that seismic resistance was not acontrolling factor in design (see Figure 10 below), an R value of 3 was used instead, to allow

comparison between the calculations presented in this document and the design criteria of NBP-

300. This assumption effected the seismic calculations, namely the value of the seismic

response coefficient. By using R=3, a higher base shear was calculated, making this a

conservative assumption for design. In regards to detailing, this system may be more

complicated and expensive. Eccentrically braced frames are used more in regions with high

seismic activity, but by using an R=3, the stability will be affected during design and will have to

be compensated through other detailing methods.

Figure 10: Response Modification Coefficient Assumption,



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Design Codes

National Business Park- Building 300 was designed in 2007 and constructed in 2009. At this point in

time, the following codes were used to complete the design of the project:

International Building Code (IBC) 2003 (with local amendments for Anne Arundel County,

Maryland)

The Life Safety Code, 2006 (NFPA 101)

ASCE 7-02

AISC, 13th Edition

ACI 318-02

AISI Specification for the Design of Light Gage Cold-Formed Structural Steel Members and the

Steel Deck Institute’s Design Requirements

Specifications for Masonry Structures, ACI 530.1/ASCE 6/TMS 602-92

Structural Welding Code, 2006, D1.1

In the redesign, the codes listed below were used to complete the analysis of National Business Park-

Building 300.

International Building Code (IBC) 2009

ASCE 7-05

AISC, 13th Edition

ACI 318-05

Vulcraft 2008 Decking Manual

UFC-340-02

AISC Seismic Design Manual


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Material Properties

Original Design

Reinforcement: Reinforcing Bars ASTM A615, Grade 60

Welded Wire Fabric (WWF) ASTM A-185

Reinforcing Bar Mats ASTM A184

Lap Splices ACI 318

Structural Steel:

Grade 50: ASTM A992, Grade 50, Fy= 50 ksi

HSS Pipes: ASTM A500, Grade C, Fy = 46ksi

Steel Tubes: ASTM A500, Grade B, Fy = 46ksi

All other steel: ASTM A 36, Fy = 36 ksi

Bolts: ASTM A325, with threads included in shear planes, ¾”

or 7/8” diameter

Shear Studs: ¾” diameter x 5” long, uniformly spaced at 24”

maximum

Metal Deck:

Floors: 3”x20 gage, bonding type

Roof: 3”x22 gage

Concrete:

Minimum Concrete Compressive Strengths (f'c)

Member 28 Day Strength (psi)

Elevated Slabs 3500

Slab-on-Grade 3500

Walls, Piers, and Grade Beams 4000

Interior Concrete Topping 3500

Concrete Exposed to Freezing 4500Grout 3000

All Other Concrete 3000

Table 1: Minimum Concrete Compressive Strengths


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Blast Redesign

The following material properties were used in the redesign:

Reinforcement: Reinforcing Bars ASTM A615, Grade 60

Welded Wire Fabric (WWF) ASTM A-185

Lap Splices ACI 318

Structural Steel:

Grade 50: ASTM A992, Grade 50, Fy= 50 ksi

All other steel: ASTM A572, Grade 50, Fy = 50 ksi

Bolts: ASTM A325, with threads included in shear planes, 1.5”

diameter

Metal Deck: Floors: 3VLI16, composite

Concrete:

Minimum Concrete Compressive Strengths (f'c)

Member 28 Day Strength (psi)

Elevated Slabs 4000

Slab-on-Grade 4000

Walls, Piers, and Grade Beams 4000

Interior Concrete Topping 4000

Concrete Exposed to Freezing 4000

Table 2: Minimum Concrete Compressive Strengths


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Gravity LoadsUsing IBC 200, ASCE 7-05, and estimation, the dead, live, and snow loads were calculated for NBP-300.

Tables 2, 3, and 4 that follow show the loads used for the analysis presented in this paper compared to

the loads used in the original design of NBP-300.

Dead Load

Superimposed Dead Loads

Area Design Load

Floors 15 psf

Roofs 15 psf

MEP 20 psf

Table 3: Superimposed Dead Loads

Live Load

Live Loads

Area Design Load ASCE 7-05 Load

Floors (including partition load) 100 psf 80 + 20 psf

Mechanical Room 125 psf -

Elevator Machine Room 150 psf -

Penthouse Floor 150 psf -

Stairs 100 psf 100 psf

Slab-on-Grade 150 psf -

Screen Enclosure and Roof Area 60 psf 60 psf

Table 4: Live Loads

Note: (-) signifies that there was no recommended value from ASCE 7-05 for the specified loading

condition and the design load was assumed to be correct.

Snow Load

Snow LoadsLoad Type Design Load ASCE 7-05 Load

Roof Snow Load 20 17.5

Drift Load Not available 81.94

Table 5: Snow Loads


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Structural Proposal

National Business Park-300 was designed initially as a typical office building. For the purpose of this

thesis, it will be assumed that National Business Park- Building 300 will be designed as a Federal office

building with the potential to be a target of terrorism. This means that, to protect the life and safety of

the occupants of NBP-300, it will be required to be designed for blast loading.

The first step in designing for blast is to determine the type and level of threat expected. It was decided

that the site would be redesigned and the security of the site increased to eliminate a vehicular

explosion from being the controlling blast load. It was then determined that the controlling load would

be a small-device, and that it most likely would be an interior explosion.

Using these specific design constraints, a typical bay of National Business Park- Building 300 was

redesigned to withstand the blast load. This includes the design of a column, a beam, a girder, and the

floor slab. A typical moment connection will also be discussed. Due to time constraints, progressive

collapse and a lateral system analysis were not completed, but would be investigated in depth if given

more time.

The intent of this redesign is to show a difference in the amount of structural damage between a

building designed for blast and a building not designed for blast, and to determine the effectiveness of

minimizing structural damage using blast design and analysis. This thesis will also look at the

effectiveness of modeling only critical sections of a building designed for blast. This will be achieved

through the use of LS-DYNA Blast Modeling software, a finite element analysis program. Comparisons

between the two structures and their reactions to an explosive event will be discussed in detail, and

through design and analysis, it will be shown that the linear elastic method is highly conservative.


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Blast Design

Until the 1960’s, blast design was reserved for facilities where accidental or chemical explosions could

occur. Blast design was not considered for ordinary structures. In the 1960’s, though, design guides for

blast started to emerge, mostly for buildings with the potential for chemical explosions. In today’s

world, terrorism is an unfortunate reality, and designing for the safety of the occupants in high risk

structures has become more important. After the Oklahoma City bombing in 1995, blast design gained

significant popularity as a design consideration for life safety. Since 9-11, blast design has become a

well-sought after design not only for federal and military building, but other high risk buildings such as

hospitals, banks, and international business buildings, to name a few, and is an issue of life safety.

Blast design has to take into consideration both impact loads from the initial wave front of the blast, and

additional time-dependent pressures, which can occur due to thermal effects behind the wave front.

Reflected pressures must be taken into account as well as ventilation to relieve built up pressures in

confined explosions. In some cases, shrapnel from the explosive container may act as a projectile, andcould cause serious damage, such as breaching.

Much of blast design is based on the members behaving plastically. Therefore, there are strict limits on

their ductility ratios and support rotations. The ductility ratio is “an approximate measure of plastic

strain based on the assumption that the curvature in the maximum moment regions increase

proportionally with deflection after yielding and plane sections remain plane,” as stated in the

“Handbook for Blast Resistant Design of Buildings.” The limit on the ductility ratio makes sure the

member’s plastic deflections are within the allowable range. The limit on the support rotation

maintains that tension membrane action that may develop in a member will be limited so as not to

cause connection failures.

Due to the unpredictable nature of blast design, often times it is overly conservative. Sometime,

though, this over conservative nature leads to unreasonable designs or costs. For that reason, structural

elements are allowed to be damaged, as long as collapse is prevented.

The principles of blast design were researched throughout this redesign and will be presented through

calculations, diagrams, tables, and figures throughout this thesis.


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Figure 11: UFC-340-02 Cover Page

Procedure

The first step in designing for blast is to decide what the maximum threat on the building will be during

the life of the building. As stated in earlier section, in this case, it was determined that an interior blast

from a briefcase sized explosive would be the design scenario. To eliminate the possibility of any other

type of blast controlling, both building setback and site control had to be determined. This information

will be presented in the Site Redesign Breadth.

Next, the blast load was determined by using

engineering judgment to estimate the type of

explosives in the blast. In this scenario, C-4 was used to

estimate the blast load. By following the UFC-340-02

(seen to the right in Figure 11) prescribed technique,

blast pressures were determined for the specific

situation set forth in the problem statement.

After determining blast loads, the member designs

were completed. A typical column, an interior beam,

the exterior girder, and the floor slab were designed for

the determined blast pressures. Plasticity was

considered in each design. Additionally, a moment

connection design was investigated.

After the hand calculations were completed for the

member designs, LS-DYNA was used to build a partial

model of NBP-300. A 3 bay by 3 bay by 3 story model of

NBP-300 was built to demonstrate several findings.

Firstly, it was a goal of this thesis to show that since the

stress in structural members decreases significantly as distance from the bay of detonation increases it

is not necessary to show an entire model. The blast forces may not even be significant in bays that are

two or three bays from the initiation of the blast in some cases. Additionally, due to the sophistication

of the LS-DYNA program and the tedious nature of the input, it was deemed not in the scope of this

thesis to build the entire model.

All hand calculations, Excel spreadsheets, tables and figures used in the design process will be presented

in the appendices of this document, and will be referenced in the detailed explanations in each section.


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Figure 12: 3D Blast Wave Projections

Blast Load Determination

Blast loads are pressure waves caused by the rapid

release of energy during a chemical reaction. The

wave propagation is spherical in nature and

dissipates with distance from the blast initiation,

which can be seen in Figure 12 to the right. Although

the spherical nature of a blast wave will cause

differential pressures along the length of certain

elements, such as columns, it can be assumed to be

linear for simplification of design and analysis.

Generally, the design process using a single degreeof freedom method is conservative, so this assumption will still provide a valid design.

The first step of the actual design process in blast design requires that the blast event load be

determined. As C-4 was assumed to be the explosive compound being used in this analysis/redesign, a

determination of a reasonable weight of the explosive needed to be calculated. Since C-4 has properties

similar to clay, it was assumed that about 50 pounds of C-4 could fit inside a large briefcase or small

rolling suitcase. Since NBP-300 is an office building, it was assumed that the explosives would have to

be camouflaged to fit into the everyday environment of the workspace, hence a briefcase or a small

business suitcase.

The weight of C-4 then needed to be converted into equivalent TNT weight. Not a lot of research has

been done with explosives other than TNT to date, so all the tables and charts in the UFC used in blast

design were designed around data collected with TNT testing. Much is known about the explosive

power of this compound, but not of many others. Therefore, the weight of actual explosives assumed in

this scenario had to be converted into equivalent pounds of TNT. The conversion was performed by

dividing the weight of C-4 by the weight conversion factor found in Table 6.1 of “The Handbook for Blast

Resistant Design of Buildings,” and an excerpt from the table can be seen in Figure 13 on the following

page. The equivalent mass for pressure was used since design pressures would be the critical values for

the structural design.


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Figure 13: C-4 Equivalency Conversion Factors, from

“The Handbook for Blast Resistant Design of Buildings”

Figure 14: Simplified Floor Plan of Lobby Area

with members to be designed highlighted

The equivalent weight was found to be 36.5 pounds of TNT.

Next, the scaled distance needed to be found, which is representative of the front of the pressure wave.

The equation was used to find the scaled distance, which equaled to 4.5 ft/lb1/3

assuming a

standoff distance of 15 feet since that would be the average distance of the explosive to the closest

structural elements in the lobby. See Figure 14 below for a simplified floor plan of the lobby area. From

this information the peak incident overpressure value of 50 psi was found using Figure 6.6 of “The

Handbook for Blast Resistant Design of Buildings”, which can be found in Appendix B.


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From these values the peak reflected pressure was found using the UFC figures. Interpolation between

Figures 2-100 and 2-97 provided a peak reflected pressure of 117 psi. Peak reflected pressure is

important in confined explosions because as the wave front encounters barriers, some of the pressures

are reflected off these materials, thereby increasing the pressure created by the blast. This is a critical

component for determining design loads.

Additionally, the value was found similarly by interpolation between Figures 2-149 and 2-146to give a value of 81 psi-ms/lb1/3. These values were then used to find the effective duration of the blast

from the UFC. An effective duration of 4.6ms was calculated.

This is only half of the battle, though. Since it is assumed, for the purpose of this thesis, that the

explosion will be a confined event, the gas pressure must also be calculated to account for the pressure

build up behind the initial front that occurs during an explosion due to the heat released during the

explosion. Gas pressure loading density was calculated by dividing the equivalent weight of TNT by the

free volume of the space. Then, Figure 2-152 of the UFC was used to determine the peak gas pressure.Additionally, the scaled gas impulse was found using Figure 7-15 of “The Handbook for Blast Resistant

Design of Buildings.”

No fragmentation was accounted for in the determination of blast loads because it was assumed that

debris would be negligible. As stated previously, the explosives would likely be contained in a briefcase

or suitcase, which is assumed to disintegrate from tremendous heat and pressure at the initiation of

blast.

A summary of the pressure and impulse design values are summarized in Table 6 on the next page and a

diagram of the bilinear pulse blast loading expected can be seen in Figure 15 on the next page. The full

calculations are provided in Appendix B.


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Figure 15: Peak Reflected Pressure and Peak

Gas Pressure vs. Time

Table 6: Summary of Blast Load Data


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Column Design

In blast design, there are many additional factors that affect the design process as compared to a

simpler design for service loads only. Certain ductility requirements must be met. Additionally, there

are both dynamic increase factors and overstrength factors that have to be considered during design.

The dynamic increase factors account for the increased yield and tensile strengths of steel due to rapid

loading.

The overstrength factor accounts for the fact that steel actually has a higher yield stress than specified

by the AISC. The Department of Defense Explosives Safety Board allows the use of this higher yield

stress in design and analysis.

There is an additional load combination to be checked, also: 1.0B+1.0D+0.25L. Since the controlling

load combination in the original design was 1.2D+1.6L for the gravity system, this was again used as the

controlling gravity load. Both load combinations were checked to determine the overall controlling

design load for the column.

In the original design, all connections were shear connections. As stated previously, the lateral system

was composed of four eccentrically braced frames, so the columns in the lobby took no lateral load as

they were not part of the original lateral system. Although the redesign requires all connections to be

moment connections, the controlling load combination for blast does not require other lateral loads to

be accounted for due to the probability of having maximum blast load simultaneously with maximum

wind or earthquake loads being very minimal. Therefore, the only lateral load applied to the column

was the blast load.

Another assumption made was that the column would behave plastically during the blast loading. Thisallows the use of higher allowable strength values described previously.

To begin the design, several more assumptions had to be made. Firstly, the column was assumed to be

pinned-pinned. This assumption follows the procedure set forth in the UFC to allow for more simplified

calculations in the linear elastic analysis procedure. Additionally, due to the orientation of the columns

in the bay, it was assumed that only the flange would be loaded during the blast. Realistically, the load

would be distributed to both the web and the flange as it would originate at about a 45 angle to the

flange face.

Finally, it was assumed that the 117 psi peak reflected pressure would be the controlling pressure in this

design. The blast pressures were determined assuming that the explosion would detonate in the center

of the bay at floor level. As stated previously, the relationship between pressure and distance is not

linear, and the closer the blast is to a structural element, the higher the blast and impulse on that

member. Although the space will be fully vented and the reflected pressure is unlikely to ever reach 117

psi, the possibility of the blast being detonated much closer to the column exists. This would inflict the

most stress and have the highest probability of damaging the structure, so the 117 psi was used to

account for the possibility of this situation. This may be conservative, but column failure could result in


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Figure 16: 3D model of typical bay with column called out

progressive collapse of the structure or “pancaking” of floors from above, so it is important to design the

column to withstand a direct blast load.

Once these assumptions were

established, the distributed and axial

loads on the column could be calculated.Then the maximum moment and shear

values were calculated. An unbraced

length of 15’ was used. The dynamic

yield strength was calculated by

multiplying the 50 ksi yield strength by

both the dynamic increase factor for yield

and the overstrength factor to get 71 ksi

yield strength. A preliminary section was

selected based on the required section

modulus, calculated by dividing themaximum moment by the yield strength,

and the requirement that bf /2tf ≤7. A

W14x159 was found to be the smallest

section that met the required 8.64 in3

section modulus and the compact section

criteria.

The W14x159 was then analyzed to

determine if it met all the requirements for blast. This included checking the d/tw ratio, web shear, the

plastic section modulus, plastic moment, plastic axial load, slenderness ratio, allowable axial stress,allowable moment, and the Euler buckling load. Modified interaction equations for blast loading and

plastic behavior were used to account for the combined axial and bending. These calculations can be

found in Appendix C.

It was determined that the member met all the design requirements, and so a W14x159 was used as the

typical column in the final design. Figure 16 above shows a typical column for the redesign.

Additionally, after designing a typical beam and girder, the column was again checked to make sure it

had the capacity for the transferred elastic moment from the moment connections in each frame. It

was determined to be adequate and a W14x159 was used as the column members in the LS-DYNA

model, which will be discussed further in later sections.


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Figure 17: 3D model of typical bay with beams called out

Beam Design

As for the column design, some assumptions

had to be made before jumping right into

the design process. The beam was assumed

to be simply supported with a maximum

support rotation of 2. Additionally, it was

assumed that the blast load would only act

over the width of the flange of the beam.

This system was not designed as a composite

system, so tributary uplift between beams

would not be an issue. Had the system been

designed as a composite system, the uplift

on the floor slab from the blast would be

transferred to the beams, and the beams

would have to be designed to withstand the

larger blast load. Finally, it was assumed

that the pressure-time loading on the beam

would be the most critical. The reason for

this is that the longer duration will have an

overall larger effect on the structure.

Additionally, since the probability of the space reaching 117 psi is minimal it can be considered not to be

the most critical condition.

Two different bay configurations were designed. Initially it was believed that using three infill beams asopposed to two would allow a significant decrease in the weight of the system. Since only the flange of

the beams were assumed to be taking the blast load (not a tributary width of floor slab transferring the

load to the beam as with composite systems), it was believed that a beam with a smaller flange width

would carry less blast load and could therefore be designed as a lighter member. A second design was

also performed using the original bay configuration with two infill beams. In the end, it was realized

that the larger members obtained in the second design with the original beam layout was a much lighter

system even though the beams had to withstand a larger blast load. Figure 17, above, shows the beams

that were designed and the final bay layout using two infill beams equally spaced.

Due to the strict rotational limitations, many iterations of the design process had to be completed. Afterperforming several iterations by hand, an Excel spreadsheet was set up to calculate most of the values,

with the exception of a few values that had to be read from tables in the UFC. An example hand

calculation of the first iteration can be found in Appendix and the Excel spreadsheets for the two

different bay layouts can be found in Appendix D.

To begin the design, both service load combinations and blast load combinations were checked to find

the controlling combination. Blast was found to be the controlling load, again using the combination


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1.0D+0.25L+1.0B. This equation accounts for the direction of the loading, in this case uplift from blast

but gravity loads from the superimposed dead loads, estimated self-weights, and live loads. From the

controlling distributed load, the moment was calculated. Plastic moment capacity was used to

determine the required section modulus for the beam. Additionally, the local buckling requirements

were determined. From these requirements, an initial member was chosen. Plastic moment was

determined to make sure it was less than the plastic moment capacity of the chosen member. The

natural period of vibration was calculated to find the ductility ratio of the beam. Additionally, the strain

rate of the beam was calculated to determine if the DIF originally estimated was accurate. The ductility

ratio and rotation of the member were calculated and compared to the criteria established at the

beginning of the design. Finally, ultimate shear capacity was compared to maximum support shear using

the dynamic shear yield stress.

Several iterations of this process were completed in order to determine that a W 27x161 was the

smallest beam that worked. The ductility ratio of this member was found to be 1.0, which means that

the DIF assumed at the beginning of the design was adequate. This means that the maximum deflection

is expected to be the equivalent elastic deflection. Also, lateral bracing was checked, and was calculatedto be required at every 3 feet along the length of the beam. This requirement will be met by the slab on

deck floor system that will produce an unbraced length of zero across the length of the beam.

Additionally, rebound of the member was checked. Rebound is the reversal of load and deflection that

occurs after the blast wave has passed. The member must have a rebound resistance greater that that

specified in Figure 5-13 of the UFC. This allows the member to remain elastic during the rebound phase.

It was determined that the member had sufficient rebound resistance.

Live load and total load deflections were also checked under service loads only. The beam was found to

be adequate in live load and total load deflections.


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Figure 18: 3D model of typical bay with girders called out

Girder Design

The girder design followed the same design

principles and procedure as the beam design,

differing only in their distribution of loading.

The interior girder was designed using point

loads from the reaction values from the

second scenario of the beam design. This

meant that there were two point loads

equally spaced along the length of the beam

with combined uplift and gravity, and a

distributed uplift blast load acting on the

bottom flange of the girder. Again, time-

pressure loading was determined to be the

more critical design pressure as it occurs

over a much longer period of time, relatively,

than the peak incident pressure. This design

may be conservative for the exterior beam,

but will be used for simplicity of design.

Again, plastic moment capacity was used to determine the required section modulus for the beam.

Additionally, the local buckling requirements were determined. From these requirements, an initial

member was chosen. A W30 x116 was chosen as a trial member. The following checks were performed

on the member to find if it met the blast requirements:

Plastic moment capacity must be greater than member plastic moment.

The strain rate of the girder was calculated to determine if the DIF originally estimated was

accurate.

The ductility ratio and rotation of the member were calculated and compared to the criteria

established at the beginning of the design.

Finally, ultimate shear capacity was compared to maximum support shear using the dynamic

shear yield stress.

Rebound resistance of the member must be greater than the required resistance.

Since the W30x116 met the specified requirements, it was chosen as the most economical member for

the design. An additional check for deflection under gravity loads only was performed. Both live load

deflection and dead load deflection were found to be less than the allowable deflections. The final

member chosen for the design was the W30x116. Figure 18, above, locates the girders that were

designed.

Hand calculations for the girder design can be found in Appendix E.


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Slab Design

Initially, it was intended to design a slab on deck for the redesign. After a few initial calculations,

though, it became evident that the capacity of such a slab would not be enough to withstand the time-

pressure loading caused by the explosion. At this point, it was decided that a two-way slab on shored

form deck could be designed to satisfy the requirements instead.

It was chosen to design the slab as a Type II Slab as defined by the UFC. This means that rebound

reinforcement will not be required because damage will be allowed and accepted in the slab if rebound

does occur. Therefore, rebound calculations were not performed on the slab, as they were unnecessary.

The slab, needing to withstand a significant uplift force without intermediate supports, was designed

with two-way reinforcement at the top. Being the more critical force on the slab, this reinforcement

was also used on the bottom of the slab to provide the same amount of resistance if the blast were from

above. Calculations for the two-way slab on shored deck form can be found in Appendix F. The design

yielded a 16.5” concrete slab on shored form deck held in place by shear studs. This was not designed

as a composite system, though, so the shear studs are intended only for holding the slab in place. Figure

19 on the next page shows a detailed section of the slab.

Although this slab design provides a solution, there are some problems. This slab is 10” deeper than the

original system. This means that the ceiling height on each floor may have to be decreased in order to

fit all the mechanical, electrical, and telecom equipment, especially since the girders and beams

remained at nearly the same depth as designed originally. Reducing the ceiling height to less than 9’

may make the space feel very crowded and cramped, which would be unpleasant for the occupants.

The other solution would be to increase the building height by 10” per floor, which would result in an

overall building height increase of 70”. This would add significant cost to the building, and would likely

be upsetting to the owner.

The system also does not provide the required lateral bracing for the beams. Although lateral bracing

can be provided by other means than a floor slab, it will add additional cost to the design. Utilizing the

floor slab as lateral bracing for the beams is the most common method of providing stability to beams

due to the economy of the design. Therefore, this would be an inefficient design in that respect, as well.

In addition to the issue of system depth, there are the issues of constructability and reasonable

solutions. A slab with a depth of 16.5” seems highly conservative for a floor system for day to day loads.

There is a lot of extra weight and cost to this system as compared to a traditional slab on deck. Manytimes, it is not economical to design slabs for blast. There are rarely solutions that are both cost

effective and provide complete safety to the occupants. For this reason, it is suggested that the

redesign utilize a composite slab on deck system. Damage will most certainly occur, including but not

limited to cracking and spalling of concrete and possible breach. As stated previously, damage is

allowed as long as collapse does not occur.


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After further research into slab design, it was determined that since the space would be vented at 5 psi,

this pressure could be used to design a floor system. This allowed a second look into slab on deck

systems. A composite deck with slab was designed, and it was found that a 3VLI16 with a 4.5” topping

would provide adequate strength for the gravity system. These calculations can also be found in

Appendix F. This gives a total system depth of 7.5”, which is only a 1” increase in system depth from the

original. A 1” decrease in the ceiling cavity depth should not be an issue with coordinating the MEP.

Additionally, this system will be much less costly, both in materials cost and labor. Overall, this system

provides an alternative solution if damage is not considered to be an issue.

Additionally, further investigation into Aluminum Foam Composite Sandwich Panels , lightweight

sandwich panels composed of aluminum foam cores, could be utilized. This innovative material takes

advantage of the high stiffness and high compressive strains of aluminum foams. Large deformations of

the sandwich panels give them the ability to absorb high quantities of energy. Placed underneath the

slab on deck system, this material could possibly provide the necessary energy dissipation to mitigate

structural failures thinner floor systems. Cost data was not looked into for this system, so no

recommendations can be made on the economy of the system as a whole. With more time, researchinto this new discovery would have been completed, and a design capitalizing on this technology may

have been possible.


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Figure 19: Two-Way Slab on Shored Form Deck Detail

Figure 20: 3VLI16 Slab on Deck Diagram taken from Vulcraft

2008 Cut Sheet


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Moment Connection Design

A typical moment connection at the exterior of the building between a beam and a column was

designed as part of the MAE requirement for thesis. Figure 21 above shows the location of the moment

connection at the exterior of the building. This is important to note because the columns at the interior

may need to be upsized to meet the strong column- weak beam requirements. With more time, an

interior connection between two beams and a column would have been investigated.

In this design, the moment connection must allow the beam to develop its full plasticity. That said, the

overstrength factors that are present in the beam calculations are not accounted for in the connection.

This is because it is necessary for the connection to be designed to develop the full strength of the

connecting members.

It was decided to use seismically prequalified connections in this design. This is because seismic

detailing has many of the same requirements as blast design. Additionally, the design can be

significantly simplified using the prequalified connections specified in the AISC Seismic Design Manual.

The column must have a higher plastic moment capacity than the beam in both situations, providing a

strong column- weak beam connection. In this type of framing, the intention is to have failure occur in

the beam before the column, since column failure could lead to pancaking of floors and ultimately

building collapse. This also helps distribute the inelastic deformations to the rest of the structure.

Another similarity is that member rotation is limited to .04 radians. Since the beam was designed for

this limit, the prequalified connections could be utilized for the design. Finally, seismic design using

prequalified connections ensures that ductile limit states will control, which is the objective in blast

design.

Figure 21: 3D model of typical bay with connection location called out


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One difference to note between seismic connection design and blast connection design is that there are

different compact section requirements. The AISC Seismic Design Manual specifies the limiting ratios for

seismically compact sections. These differ somewhat from the blast requirements for compact sections.

The blast requirements were met in the design of the column, and therefore engineering judgment was

used to follow those requirements rather than the seismic requirements. Additionally, the seismically

compact limiting ratios only apply to members of seismic load resisting systems, which this is not.

The design of the connection followed the requirements for Special Moment Frames, which are defined

as frames “expected to withstand significant inelastic deformations…” in the AISC Seismic Design

Manual since the members in blast design are expected to withstand the large inelastic deformations

that may occur as a result of blast loading.

The guidelines for prequalified connections design was used, which can be found in the specifications of

the AISC Seismic Design Manual. The beam limitations in section 5.3.1 of the specification for the AISC

Seismic Design Manual were met in accordance with the blast design requirements. Additionally, all

column limitations in section 5.3.2 were met. Additional requirements in Table 6.1 were metthroughout the design process.

Finally, the beam-column relationship requirements of sections 9.6 and 5.4 of the specification were

met. The overstrength factor used in seismic design for this calculation should not be used in blast

calculations. It was determined that the plastic capacity of the column was greater than the plastic

capacity of the beam, and therefore met the requirements for a strong column- weak beam connection.

To begin the actual design, a prequalified connection was chosen. Based on the large expected plastic

moments, an Eight-Bolt Stiffened connection was chosen, or 8ES. Next, a “dog-bone,” otherwise known

as a reduced beam section, was added to the beam to ensure that a plastic hinge occurs in the beam

and not at the connection. The reduced beam section was sized according to AISC design requirements.

The reduced beam section modulus, reduced beam shear, and the expected plastic capacity upon yield

were then calculated.

Next, the end plate and bolts were designed. It was found that shear yield in the plate controlled the

plate design. The plate was determined to be a 2.25” thick plate. The bolts were determined to be 1.5”

A325N bolts.

On the column side, it was determined that stiffeners would be required. Web crippling was

determined to be the controlling limit state, and stiffener plates were designed accordingly. It was

determined that a 1.5” thick stiffener plate was required at the top and bottom beam flange locations.

The same stiffener was used in both locations to account for load reversal during rebound of the beam.

Full penetration groove welds were determined to be required at all welded locations. It is noted that

this is quite costly, but it is also understood that this is the unfortunate downside to blast design.

Figure 22 on the following page shows a side and top view detail of the exterior moment connection.


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Figure 22: 8ES Moment Connection Detail


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Figure 23: Partial Floor Plan of Final Design with Members Labeled

Final Design

Figure 23 below shows a portion a floor plan of the redesigned structure with members labeled. This

was used for the LS-DYNA Redesign model, which will be discussed in later sections.


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LS-DYNA Modeling

LS-DYNA Blast is a finite element analysis software used for modeling and analyzing blast events.

Although not created specifically for use in the building industry, it can be used to model a building

structure to determine member reactions during a blast event. In this thesis, LS-DYNA computer

modeling and analysis will be used to meet the AE597A MAE requirements.

A partial model of NBP-300 was created in LS-DYNA to compare the results using a blast modeling

software to the hand calculations performed in the first half of the structural depth. A partial model, as

opposed to the model of the entire building was used for several reasons. Firstly, blast loads decrease

significantly with increasing distance from the origin. By building a partial model, it can be shown that

members as close as one to two bays away from the blast really don’t see much stress. Therefore, it can

be accurately assumed that a partial model will provide adequate stress-strain data. A full model would

be required for a lateral analysis though.

Second, each member had to be created individually in a text file as opposed to the user interface. This

made the process of building the model very tedious and time consuming. The scope of this thesis did

not allow for that in-depth of a modeling study.

Finally, since LS-DYNA is a finite element analysis software full of dynamic analysis equations and coding,

the run time for large, complicated models increases drastically as compared to smaller less complicated

models. In order to run several iterations, this time issue had to be taken into account.

The model was created using a base keycard file. This file has preset “cards” which are recognized

commands for the LS-DYNA programming. This is where the blast load is defined using the

*LOAD_BLAST card. The equivalent weight of TNT, the (x,y,z) location of the blast initiation, and thedelay time between the start of the model and the blast initiation were identified to define the blast.

This keyword takes into account incident pressures and reflected pressures.

The first member, a W14x159 was created in the user interface using a block mesh command (BMESH).

Figure 24 on the next page shows an example of the mesh for the beam. Initially a rectangular section

was created using five indices in the x-direction, 5 indices in the y-direction, and 6 indices in the z-

direction. The distances between indices in each direction were entered to develop the outline of the

cross section of a W-flange. Finally, material was removed from the square on either side of the

outlined web to create the W-flange shape. This created the “I” shape used for wide flanges. Then the

dimensions for the web and flanges were defined and a height was given to the column.


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Each member was created using this process. In the cases where beams framed into girders, copes were

created by adding more indices in the x-direction and z-direction, and then subtracting out the space

where the cope was required. A standard 3” cope was used and every location required for simplicity.

Once one member was created in the user interface, the file was saved, and then the text file for that

column was opened. From here, each member was created using the format from the first member.

The model was checked periodically during the text input to make sure that the geometry created was

accurate. A sample of this text is available in Appendix H.

The floor slab was created using the block mesh command again. Individual quadrants were created

and restrained at interfaces with one another. Cutouts were created at the column corners. This can

also be seen in the sample text file.

Fixed end moment connections at the ends of the beams and girders were modeled using the

*CONTACT_TIED_SURFACE_TO_SURFACE_CONSTRAINED_OFFSET option.

Once the model was built and the *LOAD_BLAST card was defined, the model was run. The blast was

defined to detonate at 2.5ms into the run time of the model. This process was repeated for the

redesigned structure using the blast members.

Figures 25 through 28 on the following pages show several isometric snapshots from the LS-DYNA

output model, including the frame, the frame with slab, and a connection detail.

Figure 24: View of Block Mesh used in LS-DYNA with Indices

Labeled in “I” and “j” directions


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Figure 25: Isometric view of structural elements of partial LS-

DYNA Model (slab not shown)

Figure 26: Isometric view of structural elements of partial LS-

DYNA Model (slab shown)


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Figure 27: Isometric view of underside of structural elements

of partial LS-DYNA Model (slab shown)

Figure 28: Close-up of Connection in LS-DYNA model


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Comparison of Results

The results of the LS-DYNA model revealed a lot about the blast design process. It showed that the

stresses are lower than expected in the redesign, and that the hand calculations may be too

conservative. Plastic strain was never reached in the redesign, so the members should not yield.

Spikes in stresses and strains occur as pressures are reflected back into the structure, causing the

stresses in the members to build. This is more noticeable in the redesigned structure since it would be

stiffer, and therefore would have a less flexible response to the blast load.

Von Mises stresses were compared to the allowable stresses for the structural steel in dynamic loading.

Von Mises stresses are calculated using the following equation:

This defines a stress plane based on stresses in the 1st, 2nd, and 3rd principle stresses. Within this plane,

stresses are below allowable. Outside the plane, yield occurs. The Von Mises stresses obtained from

the LS-DYNA output should be below the dynamic yield stresses of the structural steel, and were

compared to determine the effectiveness of the redesign. In the original structure, the stresses are well

below dynamic yield. Additionally, in the model designed for blast, we can see that no failures occur, as

the stresses are below dynamic yield stress, or 71,000 psi. This can be seen in Graphs 1 and 2 on the

following page.

Figure 29 (pages 46-47) shows several still pictures of the blast wave propagation through the structure.

As stated previously, the blast was detonated at 2.5ms into the model. Significant stresses aren’t

immediately felt by the members, though. It takes several more milliseconds for the blast wave to

induce stresses in the members. Additionally, it can be seen that the pressure distribution over the

members is not equal. The base of the column sees the pressure wave many milliseconds before the

top of the column. In the hand calculations, it was assumed that the pressure would be evenly

distributed along the beam. In reality, this would not be the case, and so the member may be overly

conservative in its design. This is true of other members in the bay as well, such as the beams, which

would see the blast load at the center of the span before either of the ends.

In addition to checking the stresses in the members, the pressure from the blast was checked. It was

found that the pressures from the blast were very close to what was calculated by hand. This data waspresented in the “Blast Load Determination” section previously. The maximum pressure was found to

be 147 psi, which occurred at 7 ms. The blast was detonated at 2.5 ms into the analysis. This drops off

quickly and remains between 22 psi and 15 psi between 14 ms and 20 ms. This is very close to the 117

peak pressure and 21 gas overpressure that was found through hand calculations. On the following

pages a progression of still images from the pressure analysis video obtained from LS-DYNA output can

be seen.


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Graph 1: Stress in Original Structure

Graph 2: Stress in Blast Designed Structure

19116

-5000

0

5000

10000

15000

20000

25000

0 5 10 15 20 25

V o n M i s e s S t r e s s ( p s i )

Time (ms)

Stress over Time

Columns

Beams b/t

ColumnsInfill Beams

Girders

Slab

-10000

0

10000

20000

30000

40000

50000

60000

70000

0 5 10 15 20 25

V o

n M i s e s S t r e s s ( p s i )

Time (ms)

Stress over Time

Columns

Beams b/t

Columns

Infill Beams

Girders

Slab


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Figure 29: Blast Pressure Progression


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Additional Comments and Conclusion

Lateral System Implications

The redesign of NBP-300 will likely have several implications to the lateral system. Originally, as stated

previously, the lateral system was composed of four eccentrically braced frames. Of the changes made

during blast design, one of the most important to note is that moment frames would be utilized.

Moment frames can provide more ductility than the present eccentrically braced frames in the building,

which would help distribute the large amount of energy and pressure produced by the blast.

With the increased weight of the structure, the seismic lateral loads would increase, and produce a

larger design load in both directions. Seismic load controlled the design in the east-west direction

originally, and would remain the controlling design load. In the north-south direction, though, wind

loads controlled the lateral design, so this would need to be checked to determine if that were still true.Seismic loading may control in the redesign due to the incredible amount of weight added to the

structure.

A further look into how ground shock affects the building’s foundations would be the next step to

determine additional strength requirements for the foundations. This is another area where additional

research and design could yield some interesting results, but as it was not in the scope of work for this

thesis, redesign of the foundations was not performed.

Effects on Foundations

As stated previously, the redesign adds tremendous weight to the structure. This means that the

current foundations may not be large enough or strong enough to support the additional weight.

Additionally, the lateral loads would increase, so overturning moments in the foundations would have to

be rechecked. Finally, due to ground shock from blast, the slab on grade may need additional strength.

A further look into how ground shock affects the building’s foundations would be then next step to

determine additional strength requirements for the foundations. This is another area where additional

research and design could yield some interesting results, but as it was not in the scope of work for this

thesis, redesign of the foundations was not performed.


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MAE Requirements

As an MAE student, certain masters’ courses were utilized in this thesis redesign. Overall, three masters’

courses were used throughout this thesis. A moment connection was designed using the information

taught in AE 534. Computer modeling was used to model and analyze the original and redesigned

structure, using material and information from AE 597A. Finally, AE 542 information on building façade

systems and glazing was utilized to complete a façade redesign.


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Breadth Topic I: Site Redesign

To redesign the site of NBP-300 several aspects of the site had to be taken into consideration. There

was originally parking located on-site, as well as truck access for garbage collection and delivery drop-

offs. Several pedestrian access points existed on multiple sides of the lot. Waste water management

also had to be considered during any modification to the site. Native plants and materials to the region

were researched. Any plants, trees, or shrubs that were to be planted had to be able to grow in the

climate and soil conditions present for the NBP-300 site. Figure 30 below shows a comparison between

the original site and the redesigned site.

Firstly, all on-site parking was eliminated. This negates the possibility of a car being used as an explosive

at a close range to the building. It will be assumed that the parking structure to the north-east of the

NBP-300 lot has the capacity to take on parking load eliminated from the site. Additionally, this lot will

also require a security system to prevent it from becoming a hazard.

The site was also expanded to meet the setback requirement of 400 ft. This means that at any point on

the building, the closest vehicular access way must be at least 400 ft away. Since there was only one

road running in front of NBP-300, the building could easily be moved further back on the site. The

driveway at the top of the site will remain, but will be for deliveries only and will have a security

checkpoint. The problem with moving the building back on the site is that this will add to the cost of the

project. An additional 80,856 square feet, or 1.86 acres would be required to meet the setback rule.

This is about a 50% increase from the original size. The cost benefit tradeoff makes this purchase worth

the money. If no setback is maintained, then the structure would be required to be designed for other,

possibly higher, blast loads.

An important change to the waste-water management had to be made. The distance to the main hook-

up was increased with the increase in setback. Fortunately, a clear path between the main run and the

building was able to be maintained, and so this wasn’t seen as a critical issue in the redesign.

The other issue that had to be dealt with is the grading of the site. Since NBP-300 has a partial

basement that is exposed at one side, it was preferable to keep this architecture undisturbed. After

looking at the site grading from the original CAD file, it was determined that the site slopes gradually

away from the back side of the building toward the east. Ultimately, it was not seen as an issue for the

architectural integrity or constructability of the building.

Site access for delivery trucks and maintenance vehicles was maintained, but moved. A gated accessroad to the north of the site leads to the loading dock at the northwest corner of the building. This

provides a security checkpoint for any vehicle entering the site, and meets the blast requirements for

site security. The road loops around to provide both entrance and exit points from site. There is a

spacious turn around area and parking area for any vehicle that may require such. The road maintains

at minimum a 24’width to provide adequate space for trucks.


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Next, a perimeter wall was designed to surround the site of NBP-300. This consisted of an 8’ high cast in

place concrete wall. The wall was sized for wind loads and overturning moments. Foundations were

sized up to a standard size rather than using the actual calculated sizes, as they were too small for the

wall size. A CAD drawing of the wall can be seen in Figure 31 on the following page. In front of the wall,decorative boulders will be used to protect the wall from vehicular impact.

More sidewalks and patio areas were added to the site to form a friendlier environment for the

occupants. There are three circular paths with bench seating connected by curved paths. This provides

a nice outdoor walking space for the occupants. Additionally, there is an outdoor patio with tables,

chairs, and umbrellas for those occupants who may wish to eat outside. Birch trees were chosen to line

the perimeter wall to obscure the view o

Documents

Spring Thesis_ Final Report_REPORT ONLY_Blast Proof