INTERIM GUIDELINES: Evaluation, Repair, Modification and Design

INTERIM GUIDELINES:Evaluation, Repair, Modification and

Design of Steel Moment FramesReport No. SAC-95-02

SAC Joint Venturea partnership of:

Structural Engineers Association of CaliforniaApplied Technology Council

California Universities for Research in Earthquake Engineering

Guidelines Development CommitteeRonald O. Hamburger, EQE International, Inc., Chair

Edward Beck, Law-Crandall, Inc.David Houghton, Myers, Nelson, Houghton, Inc.C. W. Pinkham, S. B. Barnes, Inc.Allan Porush, Dames & MooreThomas Sabol, Englekirk and Sabol, Inc.

C. Mark Saunders, Rutherford & Chekene, Inc.Barry Schindler, John A. Martin & AssociatesRobert Schwein, Schwein-Christensen LaboratoriesCharles Thiel Jr., Telesis Consultants

SAC Management CommitteeChairman - Arthur E. Ross

Structural Engineers Associationof California

Maryann PhippsArthur E. Ross

Applied Technology Council

John CoilChristopher Rojahn

California Universities forResearch in Earthquake

Engineering

Robin ShepherdCharles Thiel Jr.

SAC Technical Committee

Stephen A. MahinProgram Manager

James O. MalleyProject Director for Topical Research

Ronald O. HamburgerProject Director for Product Development

SAC Joint Venture555 University Avenue, Suite 126

Sacramento, California 95825916-427-3647

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INTERIM GUIDELINES:Evaluation, Repair, Modification and

Design of Steel Moment Frames

SAC Program to Reduce EarthquakeHazards in Steel Moment Resisting Frame

Structures

SAC Project Oversight CommitteeDr. William Hall, University of Illinois, Chair

Susan Dowty, International Conference of Building OfficialsRoger Ferch, Herrick CorporationJohn Gross, National Institute of Standards and TechnologyFred Herman, City of Palo AltoRichard Holguin, City of Los AngelesNestor Iwankiw, American Institute of Steel Construction

Roy Johnston, Brandow & JohnstonWilliam Mosseker, WHM ConsultantsJoseph Nicoletti, URS/BlumeRichard Ranous, California Office of Emergency ServicesM. P. Singh, National Science FoundationJohn Theiss, EQE International, Inc.

SAC Technical Advisory Board

Robert Bachman, Fluor-Daniel Corp.Vitelmo Bertero, University of California at BerkeleyJohn Fisher, Lehigh UniversitySubash Goel, University of MichiganThomas Heaton, United States Geologic SurveyThomas Henyey, Southern California Earthquake ConsortiumWilliam Holmes, Rutherford & Chekene, Inc.

William Honeck, Forell/Elsesser Engineers, Inc.Stanley Lindsey, Stanley V. Lindsey AssociatesHarry Martin, American Iron and Steel InstituteJohn Martin, Jr., John A Martin & AssociatesDuane Miller, Lincoln Electric CompanyCharles Thornton, Thornton-Tomasetti

Task Advisory Panel - Guidelines Development

Robert Bachman, Fluor-Daniel Corp.Vitelmo Bertero, Univ. of Calif. at BerkeleyDavid Bonneville, Degenkolb Engineers, Inc.Susan Dowty, International Conference of Building OfficialsDouglas Foutch, University of Illinois at UrbanaNancy Hamilton, Ove Arup & PartnersRichard Holguin, City of Los AngelesWilliam Holmes, Rutherford & Chekene, Inc.

John Hooper, RSP/EQEHenry Huang, County of Los AngelesHarry Martin, American Iron and Steel InstituteJohn Nissen, John A. Martin & AssociatesRobert Pyle, American Institute of Steel ConstructionJack Skiles, Omaha Public Power Corp.Charles Thornton, Thornton-TomasettiRaymond Tide, Wiss, Janney, Elstner

Interim Guidelines: Evaluation, Repair, Modification and Design of Steel Moment Frames

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Foreword and Disclaimer

The purpose of this document is to provide engineers and building officials with guidance on engineeringprocedures for evaluation, repair, modification and design of welded steel moment frame structures, to reduce therisks associated with earthquake-induced damage. The recommendations were developed by practicing engineersbased on professional judgment and experience and a preliminary program of laboratory, field and analyticalresearch. This preliminary research, known as the SAC Phase 1 program, commenced in November, 1994 andcontinued through the publication of these Interim Guidelines. Independent review and guidance was provided byan advisory panel comprised of experts from industry, practice and academia. Every reasonable effort has beenmade to assure the efficacy of the Interim Guidelines contained herein. However, users are cautioned thatresearch into the behavior of these structures is continuing. The results of this research may invalidate or suggestthe need for modification of recommendations contained herein. No warranty is offered with regard to therecommendations contained herein, either by the Federal Emergency Management Agency, the SAC JointVenture, the individual joint venture partners, their directors, members or employees. Theseorganizations and their employees do not assume any legal liability or responsibility for the accuracy,completeness, or usefulness of any of the information, products or processes included in this publication. The reader is cautioned to carefully review the material presented herein. Such information must be usedtogether with sound engineering judgment when applied to specific engineering projects. These InterimGuidelines have been developed by the SAC Joint Venture with funding provided by the Federal EmergencyManagement Agency, under contract number EMW-95-K-4672.

Acknowledgment

The SAC Joint Venture wishes to offer grateful acknowledgment to the Federal Emergency Management Agency(FEMA); FEMA’s project officer, Mr. Michael Mahoney; and technical advisor, Dr. Robert D. Hanson. Following the discovery of severe damage to steel moment-resisting frame buildings in the NorthridgeEarthquake, this agency recognized the significance of this issue to the engineering community as well as thepublic at large, and acted rapidly to provide the necessary funding to allow these Interim Guidelines to bedeveloped, published and distributed. Without the support of this agency, the important information and materialpresented herein could not have been made available.

SAC also wishes to recognize the American Institute of Steel Construction, the American Iron and Steel Institute,the American Welding Society, the California Office of Emergency Services, the Lincoln Electric Company, theStructural Shape Producers Council, and the many engineers, fabricators, inspectors and researchers whocontributed services, materials, data and invaluable advice and assistance in the production of this document.

The SAC Joint Venture555 University Avenue, Suite 126

Sacramento, CA 95825phone 916-427-3647; facsimile 916-568-0677


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Interim Guidelines: Evaluation, Repair, Modification andDesign of Steel Moment Frames

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OVERVIEW

The Northridge Earthquake of January 17, 1994, dramatically demonstrated that theprequalified, welded beam-to-column moment connection used for Special Moment ResistingFrames is much more susceptible to damage than was previously thought. The stability ofmoment frame structures in earthquakes is dependent on the capacity of the beam-columnconnection to remain intact and to resist tendencies to rotate, induced by the swaying of thebuilding. These connections were believed to be ductile and capable of withstanding repeatedcycles of large inelastic deformation. Although many affected connections were not damaged, awide spectrum of unexpected brittle connection damage did occur, ranging from minor crackingobservable only by detailed nondestructive testing (NDT) to completely severed columns. Themost commonly observed damage occurred at the welds of girder bottom flanges to columns. Complete brittle fractures of the girder flange to column connections occurred in some cases. While no casualties or collapses occurred as a result of these connection failures, and somewelded steel moment frame (WSMF) buildings were not damaged, the incidence of damage wassufficiently high in regions of strong motion to cause wide-spread concern by structural engineersand building officials.

No comprehensive tabulation is yet available to determine how many steel buildings weredamaged in the Northridge Earthquake. More than 100 damaged buildings have been identifiedso far, including hospitals and other health care facilities, government, civic and private offices,cultural facilities, residential structures, and commercial and industrial buildings. The effect ofthese observations has been a loss of confidence in the procedures used in the past to design andconstruct welded connections in steel moment frames, and a concern that structures incorporatingthese connections may not be adequately safe.

It must be understood that the structural engineering community was surprised by theperformance of these modern, code conforming structures. Prior to the discovery of this damage,many thought that WSMF structures were nearly invulnerable to earthquake damage. Theunexpected brittle fracturing and attendant loss of connection strength resulted in seriousdegradation of the overall lateral-load-resisting capability of some affected buildings. Further, theability of existing WSMF buildings to withstand earthquake-induced ground motion is nowunderstood to be significantly less than that previously assumed. Research conducted to date hasidentified some, but probably not all, of the factors leading to this observed unsatisfactorybehavior. At the same time, this research has indicated methods that can be used to improve theability of these critical connections to more reliably withstand multiple, large, inelastic cycles. These include alterations in the basic design approach as well as improved practices forspecification and control of materials and workmanship.

While the work is not yet complete, and future research is likely to provide both more reliableand more economical methods of improving the performance of these structures, the currentinvestigations have led to many design and retrofit measures that can be used today to providemore reliable and consistent performance of these buildings than occurred in the Northridge


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Earthquake. These are presented in these Interim Guidelines. They should not, however, beviewed as the only way of achieving these results, and the exercise of independent engineeringjudgment and alternative rational analytical approaches should be considered. It is anticipatedthat additional studies, planned by SAC and others, will lead to further improvements in ourunderstanding of the problems, ability to predict probable earthquake performance and methodsto design and construct more reliable structures.

There are many complex issues involved in the evaluation, repair, modification and design ofWSMF buildings for reliable earthquake performance. These include considerations ofmetallurgy, welding, fracture mechanics, systems behavior, and basic issues related to fabricationand erection practice. Much remains to be learned in each of these areas. Engineers not familiarwith the issues involved are cautioned to obtain qualified advice and third party review whencontemplating design decisions that represent significant departures from these InterimGuidelines.

The current judgment given in these Interim Guidelines is that the historic practices used forthe design and construction of WSMF connections do not provide adequate levels of buildingreliability and safety and should not continue to be used in the construction of new buildingsintended to resist earthquake ground shaking through inelastic behavior. The risk to public safetyassociated with the continued use of existing WSMF buildings is probably no greater than thatassociated with many other types of existing buildings with known seismic vulnerabilities, whichare not currently the subject of mandatory seismic rehabilitation programs. The earthquake risk ofWSMF buildings, in general, may be evaluated in accordance with the following generalprinciples:

1. The historic practices and designs used for WSMF connections are no longer appropriatefor design and construction of new steel buildings likely to experience large inelasticdemands from earthquakes. Until research is completed, and better information becomesavailable, the procedures contained in these Interim Guidelines for the design of newbuildings should be used in their place. The use of alternative systems, including boltedconstruction, braced construction, and moment-resisting frames incorporating partiallyrestrained (PR) joints could also be considered, but are not directly addressed by theseInterim Guidelines.

2. As a class, existing undamaged WSMF buildings appear to have a lower risk of collapsethan many other types of buildings with known seismic vulnerabilities, the performance ofwhich is currently implicitly accepted. Consequently, mandated or emergency programs toupgrade the performance of these buildings does not appear necessary to achieve levels oflife safety protection currently tolerated by society. However, the risk of collapse isdefinitely greater than previously thought. Individual owners should be made aware of theincreased level of seismic risk and encouraged to perform modifications to provide morereliable seismic performance, particularly in building housing many persons, or in criticaloccupancies.


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3. Following strong earthquake-induced ground shaking, WSMF buildings incorporating thevulnerable welded moment-resisting connections should be subjected to rigorousevaluations to determine the extent and implications of any damage sustained. TheseInterim Guidelines may be used to determine which buildings should be evaluated, and fordeveloping an appropriate program to perform such evaluations.

4. Structural repair and modification programs for damaged WSMF buildings shouldconsider the seismic risk inherent in the building including the local seismicity, sitegeologic conditions, the building’s individual construction characteristics, intendedoccupancy and the costs associated with alternative actions. The Interim Guidelinesprovided in this document for repair can restore a building’s pre-earthquake seismicresistance, but not significantly improve its original levels of safety or reduce the inherentseismic risk. The Interim Guidelines provided in this document for structural modification(upgrading) can be used both to improve building safety and reduce seismic risk. Exceptin those cases where regulation sets minimum acceptable standards for repair, the ultimateresponsibility for deciding whether a building should be modified for improvedperformance lies with the building owner. It is the structural engineer’s responsibility toprovide the owner with sufficient information upon which to base a decision. Thefollowing may be considered by engineers to provide such information:

a) When a WSMF has experienced damage to only a few of its moment-resistingconnections this damage should be repaired in an expeditious manner. Repair tothe original configuration, with proper materials and workmanship, will essentiallyrestore the structure’s original earthquake-resisting capacity. However, it will notresult in any significant improvement in the building’s future performance. Thefact that the building experienced only light damage should not be considered ademonstration that the building has a high degree of earthquake resistance and infuture earthquakes either more or less damage may be experienced, depending onthe particular characteristics of the event.

Connections which have been damaged can be economically modified at the sametime that repairs are made. However, in buildings where damage is limited,modification of the few damaged connections will not result in any significantimprovement in the future earthquake performance of the building. Modificationof connections throughout the structure, or provision of an alternative lateral forceresisting system should be considered as a method of substantially improvingprobable building performance; however, this will entail a significant cost premiumover the basic repair project.

b) When a WSMF has experienced damage to a significant percentage of its moment-resisting connections (on the order of 25% in any direction of resistance), inaddition to repair, consideration should be given to modifying the configuration ofthe individual damaged connections and possibly some or all of the undamagedconnections to provide improved performance in the future. Modification of only


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some connections, and not others, may cause an increase in vulnerability, due tounbalanced concentrations of stiffness and strength. Therefore, such partialmodifications should be made with due consideration of the effect on overallsystem behavior. Repair and/or modification should be completed expeditiously bystructural engineers who are experienced in the design of WSMF buildings andunderstand the features which caused the observed damage.

c) When a WSMF building has had many seriously damaged connections (on theorder of 50% in direction of resistance), owners should be informed that thisdamage may have highlighted basic deficiencies in the existing structural system, ora geologic feature which unusually amplifies site motion. In such cases the existingsystem should be both repaired and modified to provide an acceptably reliablestructural system. Modifications may consist either of local reinforcement ofindividual connections and/or alteration of the structure’s basic lateral-force-resisting system. Such modifications could include addition of braced frames,shear walls, energy dissipation devices, base isolation and similar measures.

These principles are for regular buildings that have good characteristics of design, materials,and construction workmanship. Buildings with clear and apparent seismic deficiencies posesubstantial life safety hazards regardless of the type of structural system employed, or materialtype. Such deficiencies include incomplete load paths, incompatible structural systems, irregularconfigurations such as soft or weak stories or torsional irregularity, and improper constructionpractices. Any such deficiencies found in a WSMF should be corrected.


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TABLE OF CONTENTSFOREWORD AND DISCLAIMER iiiACKNOWLEDGMENT iiiOVERVIEW v

1 INTRODUCTION1.1 Purpose 1-11.2 Scope 1-21.3 Background 1-31.4 The SAC Joint Venture 1-101.5 Sponsors 1-111.6 Summary of Phase 1 Research 1-111.7 Intent 1-141.8 Limitations 1-141.9 Use of the Guidelines 1-15

2 DEFINITIONS, ABBREVIATIONS & NOTATION

2.1 Definitions 2-12.1.1 Administrative 2-12.1.2 Technical 2-3

2.2 Abbreviations 2-92.3 Notations 2-11

3 CLASSIFICATION AND IMPLICATIONS OF DAMAGE3.1 Summary of Earthquake Damage 3-13.2 Damage Types 3-2

3.2.1 Girder Damage 3-33.2.2 Column Flange Damage 3-53.2.3 Weld Damage, Defects and Discontinuities 3-73.2.4 Shear Tab Damage 3-93.2.5 Panel Zone Damage 3-103.2.6 Other Damage 3-11

3.3 Safety Implications 3-123.4 Economic Implications 3-14

4 POST-EARTHQUAKE EVALUATION4.1 Scope 4-14.2 Preliminary Evaluation 4-2

4.2.1 Evaluation Process 4-34.2.1.1 Ground Motion 4-34.2.1.2 Additional Indicators 4-4

4.2.2 Evaluation Schedule 4-5


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4.2.3 Connection Inspections 4-64.2.3.1 Analytical Evaluation 4-74.2.3.2 Buildings with Enhanced Connections 4-7

4.2.4 Previous Evaluations and Inspections 4-84.3 Detailed Evaluation Procedure 4-10

4.3.1 Eight Step Inspection and Evaluation Procedure 4-114.3.2 Step 1 - Categorize Connections By Group 4-124.3.3 Step 2 - Select Samples of Connections for Inspection 4-13

4.3.3.1 Method A - Random Selection 4-144.3.3.2 Method B - Deterministic Selection 4-164.3.3.3 Method C - Analytical Selection 4-17

4.3.4 Step 3- Inspect the Selected Samples of Connections 4-184.3.4.1 Characterization of Damage 4-18

4.3.5 Step 4 - Inspect Connections Adjacent to Damaged Connections 4-214.3.6 Step 5 - Determine Average Damage Index for the Group 4-234.3.7 Step 6 - Determine the Probability that the Connections in a

Group at a Floor Level Sustained Excessive Damage 4-234.3.7.1 Some Connections In Group Not Inspected 4-234.3.7.2 All Connections in Group Inspected 4-25

4.3.8 Step 7 - Determine Recommended RecoveryStrategies for the Building 4-26

4.3.9 Step 8 - Evaluation Report 4-284.4 Alternative Group Selection for Torsional Response 4-304.5 Qualified Independent Engineering Review 4-32

4.5.1 Timing of Independent Review 4-334.5.2 Qualifications and Terms of Employment 4-334.5.3 Scope of Review 4-334.5.4 Reports 4-344.5.5 Responses and Corrective Actions 4-344.5.6 Distribution of Reports 4-344.5.7 Engineer of Record 4-344.5.8 Resolution of Differences 4-35

5 POST-EARTHQUAKE INSPECTION

5.1 Connection Types Requiring Inspection 5-15.1.1 Welded Steel Moment Frame (WSMF) Connections 5-15.1.2 Gravity Connections 5-35.1.3 Other Connection Types 5-3

5.2 Preparation 5-45.2.1 Preliminary Document Review and Evaluation 5-4

5.2.1.1 Document Collection and Review 5-45.2.1.2 Preliminary Building Walk-Through 5-45.2.1.3 Structural Analysis 5-45.2.1.4 Vertical Plumbness Check 5-5


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5.2.2 Connection Exposure 5-65.3 Inspection Program 5-7

5.3.1 Visual Inspection (VI) 5-75.3.1.1 Top Flange 5-85.3.1.2 Bottom Flange 5-95.3.1.3 Column and Continuity Plates 5-95.3.1.4 Beam Web Shear Connection 5-9

5.3.2 Nondestructive Testing (NDT) 5-95.3.3 Inspector Qualification 5-115.3.4 Post-Earthquake Field Inspection Report 5-125.3.5 Written Report 5-13

6 POST-EARTHQUAKE REPAIR AND MODIFICATION

6.1 Scope 6-16.2 Shoring 6-2

6.2.1 Investigation 6-26.2.2 Special Requirements 6-2

6.3 Repair Details 6-26.3.1 Approach 6-36.3.2 Weld Fractures - Type W Damage 6-36.3.3 Column Fractures - Type C1 - C5 and P1 - P6 6-66.3.4 Column Splice Fractures - Type C7 6-96.3.5 Girder Flange Fractures - Type G3-G5 6-106.3.6 Buckled Girder Flanges - Type G1 6-116.3.7 Buckled Column Flanges - Type C6 6-126.3.8 Gravity Connections 6-136.3.9 Reuse of Bolts 6-136.3.10 Welding Specification 6-14

6.4 Preparation 6-146.4.1 Welding Procedure Specifications 6-136.4.2 Welder Training 6-156.4.3 Welder Qualifications 6-156.4.4 Joint Mock-ups 6-156.4.5 Repair Sequence 6-156.4.6 Concurrent Work 6-166.4.7 Quality Control/Quality Assurance 6-16

6.5 Execution 6-166.5.1 Introduction 6-166.5.2 Girder Repair 6-206.5.3 Weld Repair (Types W1, W2, or W3) 6-216.5.4 Column Flange Repairs - Type C2 6-22

6.6 Structural Modification 6-226.6.1 Definition of Modification 6-226.6.2 Damaged vs. Undamaged Connections 6-24


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6.6.3 Criteria 6-256.6.4 Strength 6-276.6.5 Plastic Rotation Capacity 6-286.6.6 Connection Qualification and Design 6-30

6.6.6.1 Qualification Test Protocol 6-306.6.6.2 Acceptance Criteria 6-326.6.6.3 Calculations 6-32

6.6.6.3.1 Material Strength Properties 6-336.6.6.3.2 Determine Plastic Hinge Location 6-356.6.6.3.3 Determine Probable Plastic Moment at Hinges 6-356.6.6.3.4 Determine Beam Shear 6-366.6.6.3.5 Determine Strength Demands on Connection 6-376.6.6.3.6 Check Strong Column - Weak Beam Conditions 6-386.6.6.3.7 Check Column Panel Zone 6-38

6.6.7 Modification Details 6-396.6.7.1 Haunch at Bottom Flange 6-396.6.7.2 Top and Bottom Haunch 6-416.6.7.3 Cover Plate Sections 6-426.6.7.4 Upstanding Ribs 6-446.6.7.5 Side-Plate Connections 6-45

7 NEW CONSTRUCTION

7.1 Scope 7-17.2 General - Welded Steel Frame Design Criteria 7-3

7.2.1 Criteria 7-37.2.2 Strength 7-47.2.3 Configuration 7-47.2.4 Plastic Rotation Capacity 7-77.2.5 Redundancy 7-97.2.6 System Performance 7-107.2.7 Special Systems 7-10

7.3 Connection Design and Qualification Procedures - General 7-117.3.1 Connection Performance Intent 7-117.3.2 Qualification by Testing 7-117.3.3 Design by Calculation 7-11

7.4 Guidelines for Connection Qualification by Testing 7-137.4.1 Testing Protocol 7-137.4.2 Acceptance Criteria 7-14

7.5 Guidelines for Connection Design by Calculation 7-157.5.1 Material Strength Properties 7-157.5.2 Design Procedure 7-17

7.5.2.1 Determine Plastic Hinge Locations 7-177.5.2.2 Determine Probable Plastic Moment at Hinge 7-187.5.2.3 Determine Shear at Plastic Hinge 7-20


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7.5.2.4 Determine Strength Demands at Critical Sections 7-207.5.2.5 Check for Strong Column - Weak Beam Condition 7-217.5.2.6 Check Column Panel Zone 7-22

7.6 Metallurgy & Welding 7-227.7 Quality Control / Quality Assurance 7-237.8 Guidelines on Other Connection Design Issues 7-23

7.8.1 Design of Panel Zones 7-237.8.2 Design of Web Connections to Column Flanges 7-247.8.3 Design of Continuity Plates 7-247.8.4 Design of Weak Column and Weak Way Connections 7-25

7.9 Moment Frame Connections for Consideration in New Construction 7-267.9.1 Cover Plate Connections 7-277.9.2 Flange Rib Connections 7-297.9.3 Bottom Haunch Connections 7-307.9.4 Top and Bottom Haunch Connections 7-317.9.5 Side-Plate Connections 7-327.9.6 Reduced Beam Section Connections 7-357.9.7 Slip-Friction Energy Dissipating Connections 7-367.9.8 Column Tree Connections 7-377.9.9 Slotted Web Connections 7-38

7.10 Other Types of Welded Connection Structures 7-397.10.1 Eccentrically Braced Frames (EBF) 7-407.10.2 Dual Systems 7-407.10.3 Welded Base Plate Details 7-417.10.4 Vierendeel Truss Systems 7-417.10.5 Moment Frame Tubular Systems 7-427.10.6 Welded Connections of Collectors, Ties and Diaphragm Chords 7-427.10.7 Welded Column Splices 7-437.10.8 Built-up Moment Frame Members 7-43

8 METALLURGY & WELDING

8.1 Parent Materials 8-18.1.1 Steels 8-18.1.2 Chemistry 8-38.1.3 Tensile/Elongation Properties 8-48.1.4 Toughness Properties 8-68.1.5 Lamellar Discontinuities 8-9

8.2 Welding 8-108.2.1 Welding Process 8-108.2.2 Welding Procedures 8-108.2.3 Welding Filler Metals 8-118.2.4 Preheat and Interpass Temperatures 8-148.2.5 Postheat 8-168.2.6 Controlled Cooling 8-17


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8.2.7 Metallurgical Stress Risers 8-178.2.8 Welding Preparation & Fit-up 8-17

9 QUALITY CONTROL/QUALITY ASSURANCE

9.1 Quality Control 9-19.1.1 General 9-19.1.2 Inspector Qualification 9-19.1.3 Duties 9-19.1.4 Records 9-19.1.5 Engineer Obligations 9-29.1.6 Contractor Obligations 9-29.1.7 Extent of Testing 9-3

9.2 Quality Assurance & Special Inspection 9-49.2.1 General 9-49.2.2 Inspector Qualifications 9-49.2.3 Duties 9-49.2.4 Records 9-49.2.5 Engineer Obligations 9-59.2.6 Contractor Obligations 9-59.2.7 Extent of QA Testing 9-5

10 VISUAL INSPECTION

10.1 Personnel Qualification 10-110.2 Written Practice 10-110.3 Duties 10-2

11 NONDESTRUCTIVE TESTING

11.1 Personnel 11-111.1.1 Qualification 11-111.1.2 Written Practice 11-211.1.3 Certification 11-211.1.4 Recertification 11-2

11.2 Execution 11-211.2.1 General 11-211.2.2 Magnetic Particle Testing (MT) 11-311.2.3 Liquid Penetrant Testing (PT) 11-411.2.4 Radiographic Testing (RT) 11-411.2.5 Ultrasonic Testing (UT) 11-4

12 REFERENCES

Interim Guidelines:Evaluation, Repair, Modification andDesign of Steel Moment Frames Chapter 1 - Introduction

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1. INTRODUCTION

These Interim Guidelines apply to welded steel moment frame (WSMF) structures subject to largeinelastic demands from earthquakes. They provide recommended methods for: determining whichbuildings should be subjected to detailed post-earthquake evaluations; developing a program forpost-earthquake visual and non-destructive inspections of buildings suspected to have damage;evaluating the effect of discovered damage on residual building safety; identifying appropriatestrategies for continued occupancy, structural repair and/or modification of damaged buildings;and designing and constructing new buildings. These recommendations are based on an initial,Phase 1, program of research that included collection and analysis of data on buildings damagedby the Northridge Earthquake; detailed structural analyses of damaged and undamaged buildings;review of past literature on relevant research; and laboratory testing of large-scale connectionassemblies. They were developed by a group of researchers and practicing engineers, withassistance and consultation from experts in metallurgy, fracture mechanics, welding, design,structural steel production, fabrication erection and inspection.

A significant body of valuable information is presented in these Interim Guidelines, which can beused today to provide improved reliability in welded steel moment frame structures. However,much additional research remains to be performed. The parameters controlling the performanceof welded moment resisting connections are not yet fully understood, nor has consensus beenobtained on all recommendations contained herein. Engineers engaged in the design of WSMFstructures are advised to be watchful for new developments in the future.

Although portions of this document are written in code-like language, it is not a building code,nor is it intended to be used as such. Rather, it is intended to provide engineers and buildingofficials with information on what is known at the present time with regard to these structures,and to provide a series of recommendations that can be used on an interim basis to assist inpractice. The use of these Interim Guidelines is not intended to serve as a substitute for theapplication of informed engineering judgment, nor should they be used to prevent the applicationof such judgment in particular engineering applications.

1.1 Purpose

These Interim Guidelines have been prepared by the SAC Joint Venture to provide practicingengineers and building officials with:

• understanding of the types of damage buildings incorporating fully restrained (FR) weldedsteel moment frame (WSMF) connections may experience in strong earthquakes, and thepotential implications of such damage;

• a methodology for post-earthquake inspection of existing WSMF buildings, to determine ifsignificant structural damage has occurred;

Interim Guidelines: Evaluation, Repair, Modification and

Chapter 1 - Introduction Design of Steel Moment Frames

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• an approach for characterizing the relative severity of damage to a WSMF and todetermine appropriate occupancy and repair strategies;

• methods of repair for fractured, yielded and buckled elements in WSMF buildings andstructures;

• design approaches for modifications to existing WSMF buildings and structures with FRconnections to improve performance in future earthquakes; and

• design approaches for connections in new WSMF buildings and structures for improvedperformance in future strong earthquakes.

Earlier publications by the SAC Joint Venture on this topic include a series of three DesignAdvisories and the proceedings of an International Workshop (SAC-1994-1). The InternationalWorkshop, held in October, 1994 was attended by more than 100 invited researchers, practicingengineers, representatives of industry, and government agencies, and provided an initial focus to theinvestigations of fractures sustained by welded steel moment-resisting buildings in the NorthridgeEarthquake. Design Advisory No. 1 (SAC-1994-2) and Design Advisory No. 2 (SAC-1994-3)contained collections of papers and topical reports prepared by practicing engineers, building officials,industry groups and researchers, suggesting factors which contributed to the observed damage,methods of repairing damage and designing new structures to avoid such damage in the future. DesignAdvisory No. 3 (SAC-1995) categorized the information presented in the previous advisories into aseries of discrete engineering issues and presented the consensus opinions of a panel of practicingengineers, researchers and industry representatives with regard to appropriate response to these issues.Dissenting opinions and commentary were also provided as were specific recommendations fordirected research required to provide resolution to a number of these issues.

These Interim Guidelines provide specific engineering recommendations based on the results of aninitial limited program of research. This research included evaluation of the characteristics of groundmotion experienced throughout the Los Angeles area during the Northridge Earthquake, projection ofpotential ground motions resulting from future earthquakes in this region, analytical investigation ofboth damaged and undamaged structures affected by the Northridge Earthquake for their response to arange of ground motions, laboratory testing of representative beam-column connections in undamaged,damaged, repaired, and reinforced states, parametric studies on the effects of strain rate and toughnesson connection performance, surveys of engineers and building owners to collect data on the extent ofdamage sustained in the Northridge Earthquake, and statistical evaluation of the data collected andengineering analysis of all of the above.

1.2 Scope

These Interim Guidelines are applicable to steel moment-resisting frame structuresincorporating fully restrained connections in which the girder flanges are welded to the columnsand which are subject to significant inelastic demands from strong earthquake ground shaking. Recommendations are provided with regard to:


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• Designation of buildings to be inspected following an earthquake producing strongground motion;

• Scope of inspection for buildings so designated;

• Appropriate types of repairs for damaged buildings;

• Methods to modify buildings to reduce the probability of connection fracture damagein future earthquake events;

• Design of new Special Moment Resisting Frame (SMRF) buildings for seismicresistance;

• Design of new Ordinary Moment Resisting Frame (OMRF) buildings located inUniform Building Code (UBC) Seismic Zones 3 and 4 {National Earthquake HazardsReduction Program (NEHRP) Map Areas 6 and 7}; and

• Quality Assurance and Control in the repair, modification and construction of WSMFbuildings.

Commentary: The design recommendations contained in these InterimGuidelines are generally applicable to SMRF structures designed for earthquakeresistance and to those OMRF structures located within UBC Seismic Zones 3and 4 {NEHRP Map Areas 6 and 7}. The recommendations should be consideredfor the design of any welded steel moment frame structure that is desired to havea high degree of reliability for resisting earthquake induced forces. In particular,they should be considered for buildings occupied by a large number of people. Chapter 7 provides further guidelines on this applicability.

1.3 Background

Following the January 17, 1994 Northridge, California Earthquake, more than 100 steel buildingswith welded moment-resisting frames were found to have experienced beam-to-column connectionfractures. The damaged structures cover a wide range of heights ranging from one story to 26 stories;and a wide range of ages spanning from buildings as old as 30 years of age to structures just beingerected at the time of the earthquake. The damaged structures are spread over a large geographicalarea, including sites that experienced only moderate levels of ground shaking. Although relatively fewsuch buildings were located on sites that experienced the strongest ground shaking, damage to thesebuildings was quite severe. Discovery of these extensive connection fractures, often with littleassociated architectural damage to the buildings, has been alarming. The discovery has also causedsome concern that similar, but undiscovered damage may have occurred in other buildings affected bypast earthquakes. Indeed, there are isolated reports of such damage. In particular, a publicly ownedbuilding at Big Bear Lake is known to have been damaged by the Landers-Big Bear, Californiasequence of earthquakes, and at least one building, under construction in Oakland, California at thetime of the 1989 Loma Prieta Earthquake, was reported to have experienced such damage.



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WSMF construction is used commonly throughout the United States and the world, particularlyfor mid- and high-rise construction. Prior to the Northridge Earthquake, this type of construction wasconsidered one of the most seismic-resistant structural systems, due to the fact that severe damage tosuch structures had rarely been reported in past earthquakes and there was no record of earthquake-induced collapse of such buildings, constructed in accordance with contemporary US practice.However, the widespread severe structural damage which occurred to such structures in theNorthridge Earthquake calls for re-examination of this premise.

The basic intent of the earthquake resistive design provisions contained in the building codes is toprotect the public safety, however, there is also an intent to control damage. The developers of thebuilding code provisions have explicitly set forth three specific performance goals for buildingsdesigned and constructed to the code provisions (SEAOC - 1990). These are to provide buildings withthe capacity to

• resist minor earthquake ground motion without damage;

• resist moderate earthquake ground motion without structural damage but possibly somenonstructural damage; and

• resist major levels of earthquake ground motion, having an intensity equal to the strongesteither experienced or forecast for the building site, without collapse, but possibly with somestructural as well as nonstructural damage.

In general, WSMF buildings in the Northridge Earthquake met the basic intent of the buildingcodes, to protect life safety. However, many of these buildings experienced significant damage thatcould be viewed as failing to meet the intended performance goals with respect to damage control. Further, some members of the engineering profession (SEAOC - 1995b) and government agencies(Seismic Safety Commission - 1995) have stated that even these performance goals, are inadequate forsociety’s current needs.

WSMF buildings are designed to resist earthquake ground shaking, based on the assumption thatthey are capable of extensive yielding and plastic deformation, without loss of strength. The intendedplastic deformation consists of plastic rotations developing within the beams, at their connections to thecolumns, and is theoretically capable of resulting in benign dissipation of the earthquake energydelivered to the building. Damage is expected to consist of moderate yielding and localized buckling ofthe steel elements, not brittle fractures. Based on this presumed behavior, building codes require aminimum lateral design strength for WSMF structures that is approximately 1/8 that which would berequired for the structure to remain fully elastic. Supplemental provisions within the building code,intended to control the amount of interstory drift sustained by these flexible frame buildings, typicallyresult in structures which are substantially stronger than this minimum requirement and in zones ofmoderate seismicity, substantial overstrength may be present to accommodate wind and gravity loaddesign conditions. In zones of high seismicity, most such structures designed to minimum code criteriawill not start to exhibit plastic behavior until ground motions are experienced that are 1/3 to 1/2 theseverity anticipated as a design basis. This design approach has been developed based on historicalprecedent, the observation of steel building performance in past earthquakes, and limited research that


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has included laboratory testing of beam-column models, albeit with mixed results, and non-linearanalytical studies.

Observation of damage sustained by buildings in the Northridge Earthquake indicates that contraryto the intended behavior, in many cases brittle fractures initiated within the connections at very lowlevels of plastic demand, and in some cases, while the structures remained elastic. Typically, but notalways, fractures initiated at, or near, the complete joint penetration (CJP) weld between the beambottom flange and column flange (Figure 1-1). Once initiated, these fractures progressed along anumber of different paths, depending on the individual joint conditions. Figure 1-1 indicates just one ofthese potential fracture growth patterns. Investigators initially identified a number of factors whichmay have contributed to the initiation of fractures at the weld root including: notch effects created bythe backing bar which was commonly left in place following joint completion; sub-standard weldingthat included excessive porosity and slag inclusions as well as incomplete fusion; and potentially, pre-earthquake fractures resulting from initial shrinkage of the highly restrained weld during cool-down. Such problems could be minimized in future construction, with the application of appropriate weldingprocedures and more careful exercise of quality control during the construction process. However, it isnow known that these were not the only causes of the fractures which occurred.

Backing bar

Column flange

Beam flange

Fused zone

Fracture

Figure 1-1 - Common Zone of Fracture Initiation in Beam -Column Connection

Current production processes for structural steel shapes result in inconsistent strength anddeformation capacities for the material in the through-thickness direction. Non-metallic inclusions inthe material, together with anisotropic properties introduced by the rolling process can lead to lamellarweakness in the material. Further, the distribution of stress across the girder flange, at the connectionto the column is not uniform. Even in connections stiffened by continuity plates across the panel zone,significantly higher stresses tend to occur at the center of the flange, where the column web produces alocal stiffness concentration. Large secondary stresses are also induced into the girder flange tocolumn flange joint by kinking of the column flanges resulting from shear deformation of the columnpanel zone.

The dynamic loading experienced by the moment-resisting connections in earthquakes ischaracterized by high strain tension-compression cycling. Bridge engineers have long recognized thatthe dynamic loading associated with bridges necessitates different connection details in order to provideimproved fatigue resistance, as compared to traditional building design that is subject to “static”



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loading due to gravity and wind loads. While the nature of the dynamic loads resulting fromearthquakes is somewhat different than the high cycle dynamic loads for which fatigue-prone structuresare designed, similar detailing may be desirable for buildings subject to seismic loading.

In design and construction practice for welded steel bridges, mechanical and metallurgical notchesshould be avoided because they may be the initiators of fatigue cracking. As fatigue cracks grow underrepetitive loading, a critical crack size may be reached whereupon the material toughness (which is afunction of temperature) may be unable to resist the onset of brittle (unstable) crack growth. Thebeam-to-column connections in WSMF buildings are comparable to category C or D bridge details thathave a reduced allowable stress range as opposed to category B details for which special metallurgical,inspection and testing requirements are applied. The rapid rate of loading imposed by seismic events,and the complete inelastic range of tension-compression-tension loading applied to these connections ismuch more severe than typical bridge loading applications. The mechanical and metallurgical notchesor stress risers created by the beam-column weld joints are a logical point for fracture problems toinitiate. This, coupled with the tri-axial restraint provided by the beam web and the column flange, is arecipe for brittle fracture.

During the Northridge Earthquake, once fractures initiated in beam-column joints, they progressedin a number of different ways. In some cases, the fractures initiated but did not grow, and could not bedetected by visual observation. In other cases, the fractures progressed completely through thethickness of the weld, and if fireproofing was removed, the fractures were evident as a crack throughexposed faces of the weld, or the metal just behind the weld (Figure 1-2a). Other fracture patterns alsodeveloped. In some cases, the fracture developed into a through-thickness failure of the column flangematerial behind the CJP weld (Figure 1-2b). In these cases, a portion of the column flange remainedbonded to the beam flange, but pulled free from the remainder of the column. This fracture pattern hassometimes been termed a “divot” or “nugget” failure.

A number of fractures progressed completely through the column flange, along a near horizontalplane that aligns approximately with the beam lower flange (Figure 1-3a). In some cases, thesefractures extended into the column web and progressed across the panel zone Figure (1-3b). Investigators have reported some instances where columns fractured entirely across the section.

a. Fracture at Fused Zone b. Column Flange “Divot” Fracture

Figure 1-2 - Fractures of Beam to Column Joints


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a. Fractures through Column Flange b. Fracture Progresses into Column Web

Figure 1-3 - Column Fractures

Once these fractures have occurred, the beam - column connection has experienced a significantloss of flexural rigidity and capacity. Residual flexural strength and rigidity must be developed througha couple consisting of forces transmitted through the remaining top flange connection and the webbolts. Initial research suggests that residual stiffness is approximately 20% of that of the undamagedconnection and that residual strength varies from 10% to 40% of the undamaged capacity, whenloading results in tensile stress normal to the fracture plane. When loading produces compressionacross the fracture plane, much of the original strength and stiffness remain. However, in providingthis residual strength and stiffness, the beam shear connections can themselves be subject to failures,consisting of fracturing of the welds of the shear plate to the column, fracturing of supplemental weldsto the beam web or fracturing through the weak section of shear plate aligning with the bolt holes(Figure 1-4).

Figure 1-4 - Vertical Fracture through Beam Shear Plate Connection

Despite the obvious local strength impairment resulting from these fractures, many damagedbuildings did not display overt signs of structural damage, such as permanent drifts, or extreme damageto architectural elements. Until news of the discovery of connection fractures in some buildings began



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to spread through the engineering community, it was relatively common for engineers to performcursory post-earthquake evaluations of WSMF buildings and declare that they were undamaged. Inorder to reliably determine if a building has sustained connection damage it is necessary to removearchitectural finishes and fireproofing and perform nondestructive examination including visualinspection and ultrasonic testing. Even if no damage is found, this is a costly process. Repair ofdamaged connections is even more costly. A few WSMF buildings have sustained so much connectiondamage that it has been deemed more practical to demolish the structures rather than to repair them.

Immediately following the Northridge Earthquake, a series of tests of beam-column subassemblieswere performed at the University of Texas at Austin, under funding provided by the AISC as well asprivate sources. The test specimens used heavy W14 column sections and deep (W36) beam sectionscommonly employed in some California construction. Initial specimens were fabricated using thestandard prequalified connection specified by the Uniform Building Code (UBC). Section 2211.7.1.2of UBC-94 {NEHRP-91 Section 10.10.2.3} specified this prequalified connection as follows:

“2211.7.1.2 Connection strength. The girder top column connection may be considered to be adequateto develop the flexural strength of the girder if it conforms to the following:

1. the flanges have full penetration butt welds to the columns.

2. the girder web to column connection shall be capable of resisting the girder shear determined for thecombination of gravity loads and the seismic shear forces which result from compliance with Section2211.7.2.1. This connection strength need not exceed that required to develop gravity loads plus3(Rw/8) times the girder shear resulting from the prescribed seismic forces.

Where the flexural strength of the girder flanges is greater than 70 percent of the flexural strength ofthe entire section, (i.e. btf/(d-tf)Fy>0.7ZxFy) the web connection may be made by means of welding orhigh-strength bolting.

For girders not meeting the criteria in the paragraph above, the girder web-to-column connection shallbe made by means of welding the web directly or through shear tabs to the column. That welding shallhave a strength capable of developing at least 20 percent of the flexural strength of the girder web. Thegirder shear shall be resisted by means of additional welds or friction-type slip-critical high strength boltsor both.

and:

2211.7.2.1 Strength. The panel zone of the joint shall be capable of resisting the shear induced by beambending moments due to gravity loads plus 1.85 times the prescribed seismic forces, but the shearstrength need not exceed that required to develop 0.8ΣMs of the girders framing into the column flangesat the joint...”

In order to investigate the effects that backing bars and weld tabs had on connection performance,these were removed from the specimens prior to testing. Despite these precautions, the test specimensfailed at very low levels of plastic loading. Following these tests at the University of Texas at Austin,reviews of literature on historic tests of these connection types indicated a significant failure rate in past


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tests as well, although these had often been ascribed to poor quality in the specimen fabrication. It wasconcluded that the prequalified connection, specified by the building code, was fundamentally flawedand should not be used for new construction in the future.

In retrospect, this conclusion may have been premature. When the first test specimens for thatseries were fabricated, the welder failed to follow the intended welding procedures. Further, no specialprecautions were taken to assure that the materials incorporated in the work had specified toughness. Some engineers, with knowledge of fracture mechanics, have suggested that if materials with adequatetoughness are used, and welding procedures are carefully specified and followed, adequate reliabilitycan be obtained from the traditional connection details. Others believe that the conditions of high tri-axial restraint present in the beam flange to column flange joint (Blodgett - 1995) would prevent ductilebehavior of these joints regardless of the procedure used to make the welds. Further they point to theimportant influence of the relative yield and tensile strengths of beam and column materials, and othervariables, that can affect connection behavior. To date, there has not been sufficient researchconducted to resolve this issue.

In reaction to the University of Texas tests as well as the widespread damage discovered followingthe Northridge Earthquake, and the urging of the California Seismic Safety Commission, in September,1994 the International Conference of Building Officials (ICBO) adopted an emergency code change tothe 1994 edition of the Uniform Building Code (UBC-94) {1994 NEHRP Recommended ProvisionsSection 5.2}. This code change, jointly developed by the Structural Engineers Association ofCalifornia, AISI and ICBO staff, deleted the prequalified connection and substituted the following in itsplace:

“2211.7.1.2 Connection Strength. Connection configurations utilizing welds or high-strengthbolts shall demonstrate, by approved cyclic test results or calculation, the ability to sustaininelastic rotation and develop the strength criteria in Section 2211.7.1.1 considering the effect ofsteel overstrength and strain hardening.”

“2211.7.1.1 Required strength. The girder-to-column connection shall be adequate to develop thelesser of the following:

1. The strength of the girder in flexure.

2. The moment corresponding to development of the panel zone shear strength as determined fromformula 11-1.”

Unfortunately, neither the required “inelastic rotation”, or calculation and test procedures are welldefined by these code provisions. Design Advisory No. 3 (SAC-1995) included an InterimRecommendation (SEAOC-1995) that attempted to clarify the intent of this code change, and thepreferred methods of design in the interim period until additional research could be performed andreliable acceptance criteria for designs re-established. The State of California similarly published a jointInterpretation of Regulations (DSA-OSHPD - 1994) indicating the interpretation of the current coderequirements which would be enforced by the state for construction under its control. This appliedonly to the construction of schools and hospitals in the State of California. The intent of these Interim



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Guidelines is to supplement these previously published documents and to provide updatedrecommendations based on the results of the limited directed research performed to date.

1.4 The SAC Joint Venture

SAC is a joint venture of the Structural Engineers Association of California (SEAOC), theApplied Technology Council (ATC), and California Universities for Research in EarthquakeEngineering (CUREe), formed specifically to address both immediate and long-term needs relatedto solving the problem of the WSMF connection. SEAOC is a professional organizationcomprised of more than 3,000 practicing structural engineers in California. The volunteer effortsof SEAOC’s members on various technical committees have been instrumental in the developmentof the earthquake design provisions contained in the Uniform Building Code as well as theNational Earthquake Hazards Reduction Program (NEHRP) Recommended Provisions forSeismic Regulations for New Buildings. The Applied Technology Council is a non-profitorganization founded specifically to perform problem-focused research related to structuralengineering and to bridge the gap between civil engineering research and engineering practice. Ithas developed a number of publications of national significance including ATC 3-06, which servedas the basis for the NEHRP Recommended Provisions. CUREe’s eight institutional members are:the University of California at Berkeley, the California Institute of Technology, the University ofCalifornia at Davis, the University of California at Irvine, the University of California at LosAngeles, the University of California at San Diego, the University of Southern California, andStanford University. this collection of university earthquake research laboratory, library,computer and faculty resources is the most extensive in the United States. The SAC JointVenture allows these three organizations to combine their extensive and unique resources,augmented by subcontractor universities and organizations from around the nation, into anintegrated team of practitioners and researchers, uniquely qualified to solve problems inearthquake engineering.

The SAC Joint Venture developed a two phase program to solve the problem posed by thediscovery of fractured steel moment connections following the Northridge Earthquake. Phase 1of this program was intended to provide guidelines for the immediate post-Northridge problemsof identifying damage in affected buildings and repairing this damage. In addition, Phase 1included dissemination of the available design information to the professional community. Itincluded convocation of a series of workshops and symposiums to define the problem;development and publication of a series of Design Advisories (SAC-1994-1, SAC-1994-2, SAC-1995); limited statistical data collection, analytical evaluation of buildings and laboratory research;and the preparation of these Interim Guidelines. Phase 2 will consist of a longer term program ofresearch and investigation to more carefully define the conditions which lead to the prematureconnection fractures and to develop sound guidelines for seismic design and detailing of improvedor alternative WSMF connections for new buildings, as well as reliable retrofitting concepts forexisting undamaged WSMF structures.

The SAC Joint Venture’s unique capability to combine the efforts of researchers, industryrepresentatives, code writers and practicing structural engineers is being applied to all major tasks.


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In addition, a Technical Oversight Committee and Technical Advisory Board with nationwidemembership from the engineering, research and steel construction communities has been established tooversee the input of information, quality of technical investigations, and development ofrecommendations, and to assist in disseminating the information obtained.

1.5 Sponsors

Funding for the Phase 1 SAC Steel Program was provided by the California Office of EmergencyServices and the Federal Emergency Management Agency. Special efforts have been made to maintaina liaison with the engineering profession, researchers, the steel industry, fabricators, code writingorganizations and model code groups, building officials, insurance and risk-management groups andfederal and state agencies active in earthquake hazard mitigation efforts. SAC wishes to acknowledgethe support and participation of each of the above groups as well as the American Iron and SteelInstitute, the American Institute of Steel Construction, the Structural Shape Producers Council, theAmerican Welding Society and the Lincoln Electric Company for the contribution of technical adviceand assistance as well as material directly used in the research program. Acknowledgment is also madeof the many engineers, fabricators, inspectors and researchers who contributed services and data foruse in the development of these Guidelines.

1.6 Summary of Phase 1 Research

These Interim Guidelines are based on the material presented in Design Advisory No. 3 (SAC-1995), professional judgment and experience, a review of past relevant research, concurrent researchbeing performed under grants provided by the National Science Foundation and supplementalinformation obtained in the SAC Phase 1 research program. This research included:

• Collection of data on buildings damaged by the Northridge Earthquake. This consisted ofthe collection of detailed information on the configuration and detailing of WSMF buildingsdamaged by the Northridge Earthquake, together with data on the distribution, type andseverity of damage within each structure. This work was conducted as an extension of anearlier survey, performed under funding from the National Institute of Standards andTechnology (Youssef, et. al. - 1994). Data on a total of 89 buildings is available from thesecombined studies (Bonowitz & Youssef - 1995)

• A telephone survey was conducted on a random sample of 200 steel framed buildingslocated within the zone which experienced estimated ground motion with a peak horizontalacceleration of 0.2g or greater during the Northridge Earthquake. The intent of this surveywas to determine the geographic distribution of inspected, damaged and repaired structuresin order to correlate damage with ground motion parameters and other factors. (MichaelDurkin & Associates - 1995)

• A series of interviews were conducted with engineers, inspectors, building officials andothers engaged in the investigation and repair of a number of damaged WSMF buildings. The purpose of these interviews was to collect data on pertinent interpretations or trendsnoted by engineers and others engaged in this work. (Gates & Morden - 1995)



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• Maps of ground motion parameters(peak ground acceleration and pseudo spectral velocityat various periods) were developed for the San Fernando Valley and surrounding areasaffected by strong ground motion in the Northridge Earthquake, based on fault rupture andground motion propagation modeling techniques. Time histories of ground motion weredeveloped for various discrete sites, using these same modeling techniques. Theseestimated ground motions were developed for use in comparing geographic distributions ofdamage with ground motion parameters, and as a basis for performing structural analysesof selected buildings. (Sommerville- 1995)

• A fracture model element was developed for use with the DRAIN-2D, non-linear analysissoftware, to permit analytical simulation of the effect of beam-column connection fractureson overall structural behavior. (Campbell - 1995)

• A series of linear and non-linear structural analyses were performed on eight WSMFbuildings which were damaged by the Northridge Earthquake and on two WSMF buildingsadjacent to two of these structures, which were not damaged. The purpose of theseanalyses was to explore the ability of analytical methods to predict the presence of damagewithin buildings as well as to predict specific locations within buildings where damage islikely to have occurred. In addition, these analyses were intended to indicate thresholddemand levels at which damage is likely to have occurred, to provide information on thetotal demands developed in structures during response to various earthquake groundmotions, and to explore the potential for earthquake induced collapse. (Krawinkler, et. al.1995), (Engelhardt, et. al. - 1995a), (Hart, et. al. - 1995), (Kariotis & Eimani - 1995),(Anderson & Fillippou - 1995), (Naeim, et. al. - 1995), (Uang, et. al. - 1995), (Paret &Sasaki - 1995)

• A series of parametric analytical investigations were performed to assess the influence ofvarious ground motions and structural characteristics on seismic response of WSMFbuildings. These included investigations involving hypothesized fractures of beam-columnconnections for various real and idealized frame structures subject to various intenseground motion records. The consequences of these ground motions were assessed as wasthe sensitivity of response to vertical ground motion and to various analytical modelingassumptions. (Iwan - 1995), (Hall - 1995), (Hart et. al. - 1995b), (Englehardt, et. al.1995b), (Krawinkler, et. al. - 1995)

• Four damaged beam-column connections were removed from a WSMF building whichwas affected by the Northridge Earthquake and subsequently demolished. Thesespecimens were moved to a laboratory and subjected to testing to determine their residualstrength and stiffness, for use in making assessments as to the consequences of fracturedamage to overall building stability. Following this testing, the specimens were repairedand re-tested, to judge the effectiveness of the repair techniques. In addition, detailedbuilding analyses were performed. (Anderson - 1995)

• A total of 12 large scale beam-column assemblages were fabricated using typical pre-Northridge detailing practice and following correct welding procedures. These werecycled inelastically, using a testing protocol similar to that indicated in ATC-24 (Applied


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Technology Council - 1988) and experienced failure at low levels of plastic demand. Following initial testing and failure, the specimens were repaired using specificationsfollowed by engineers in the Los Angeles area, or repaired and reinforced using detailsproposed by Los Angeles area engineers. The purpose of these tests was to explorewhether initial structural capacity could be re-established in damaged structures bycommon repair techniques, and to determine the efficacy of proposed structuralreinforcement techniques. (Popov et. al. - 1995), (Bertero and Whitaker- 1995), (Uang -1995b), (Engelhardt - 1995c)

• Four large scale beam-column subassemblies were fabricated using selected detailsrecommended in these guidelines for new construction and subjected to cyclic testing tofailure.

• A series of acoustic emission recordings were made on the large scale structuralassemblages tested in the laboratory to assist in interpretation of the fracture sequence andto explore the ability of acoustic instrumentation techniques to identify damage in WSMFbuildings affected by strong ground motion. (Thewalt - 1995), (Engelhardt, et. al. - 1995d)

• A series of ambient vibration tests were performed on damaged buildings in order todetermine the ability of low level vibration testing to be used as a method of detectingdamage in WSMF buildings affected by strong ground motion, and to calibrate analyticalmodels. (Beck - 1995)

• Specimens from damaged connections in buildings affected by the Northridge Earthquakewere removed from the buildings and subjected to metallurgic and fractographic analysesto determine the fracture mechanisms and effect of metallurgy on fracture behavior. (ATLSS - 1995a)

• A series of moderate-scale “T” specimens were fabricated to simulate the connection of abeam bottom flange to a column flange in a major axis WSMF connection. These testswere performed to explore the ability to economically use moderate scale models toexplore the behavior of large scale beam-column assemblages and also to performparametric experimental studies on the effects of strain-rate on specimen behavior and theeffects of weld metal notch-toughness and weld procedure on connection behavior.(ATLSS - 1995b)

Additional information was collected from various other sources, including researchperformed under funding provided by the American Institute of Steel Construction, the NationalScience Foundation, and the National Institute of Standards and Technology, as well as testingperformed as part of privately sponsored research (Allen, et. al. -1995, Jokerst - 1995) andlessons learned in the inspection, evaluation and repair of buildings which has taken place to date.



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1.7 Intent

These Interim Guidelines are primarily intended for two different groups of potential users:

a) Engineers engaged in evaluation, repair, and upgrade of existing WSMF buildings and inthe design of new WSMF buildings incorporating either Special Moment-Resisting Framesor Ordinary Moment-Resisting Frames utilizing welded beam-column connections. Therecommendations for new construction are applicable to all WSMF construction expectedto resist earthquake demands through plastic behavior.

b) Regulators and building departments responsible for control of the evaluation, repair, andoccupancy of WSMF buildings that have been subjected to strong ground motion and forregulation of the design, construction, and inspection of new WSMF buildings.

The fundamental goal of the information presented in these Interim Guidelines is to help identifyand reduce the risks associated with earthquake-induced fractures in WSMF buildings throughprovision of timely information on how to inspect existing buildings for damage, repair damage iffound, upgrade existing buildings and design new buildings. The information presented here primarilyaddresses the issue of beam-to-column connection integrity under the severe plastic demands that canbe produced by building response to strong ground motion. Users are referred to the applicableprovisions of the locally prevailing building code for information with regard to other aspects ofbuilding construction and earthquake damage control.

1.8 Limitations

The information presented in these Interim Guidelines is based on limited research conducted sincethe Northridge Earthquake, review of past research and the considerable experience and judgment ofthe professionals engaged by SAC to prepare and review this document. Additional research on suchtopics as the effect of floor slabs on frame behavior, the effect of weld metal and base metal toughness,the efficacy of various beam-column connection details and the validity of current standard testingprotocols for prediction of earthquake performance of structures are planned as part of the Phase 2program and will likely provide important information not available at the time these Guidelines wereformulated. Therefore, some recommendations cited herein may change as a result of forthcomingresearch results.

Although the information presented is limited almost exclusively to technical engineering issues, itis well recognized that acceptable solutions to the steel WSMF problems must eventually address thenon-technical concerns of building officials, owners, tenants, contractors, lenders, insurers, andlegislators. It is hoped that by limiting the scope of this document to technical matters, this materialcan provide an objective basis for further discussion and debate.

The information presented here is based on consideration of the typical building and WSMF frameconfigurations found in buildings today. Non-building structures (e.g. bridges, towers, or openframeworks) are not specifically addressed; however, to the extent that construction of these structures


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is similar to that for buildings, the information presented may be applicable. Beams and columns areassumed to be constructed of hot-rolled or built-up wide flange sections with beams framing into thecolumn flange, although some recommendations should also apply to box columns and beams framingto column webs.

The recommendations presented herein represent the group consensus of the committee ofGuideline Writers employed by SAC following independent review by a technical advisory panel,Project Oversight Committee and Technical Advisory Board. They may not reflect the individualopinions of any single participant. They do not necessarily represent the opinions of the SACJoint Venture, the Joint Venture partners, or the sponsoring agencies. Users are cautioned thatavailable information on the nature of the WSMF problem is in a rapid stage of development andany information presented herein must be used with caution and sound engineering judgment.

1.9 Use of the Guidelines

It is anticipated that the users of these Interim Guidelines will generally desire information inone or more of the following specific areas:

1. a general understanding of the performance of WSMF buildings in the NorthridgeEarthquake and the probable performance of such buildings in future earthquakes;

2. inspection, evaluation and repair of buildings which have been affected by theNorthridge Earthquake or other earthquakes;

3. seismic upgrade of existing WSMF buildings to provide more reliable performance infuture earthquakes; and

4. design of new WSMF buildings to provide more reliable performance in futureearthquakes.

In order to provide information useful to all such users, this document has been made quitebroad. Table 1-1 provides a quick reference to the contents of this document.



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Table 1-1 - Quick Reference Guide

User Need Section Contents

General Information Chapter 1 Introductory material

Chapter 2 Abbreviations, Notation & Terminology

Chapter 3 Damage Classification, Safety Issues, EconomicLoss Data

Post-Earthquake Inspection, Chapters 1-3 Background Information

Evaluation, and Repair Chapters 4 and 5 Inspection

Chapter 6 Repair

Chapter 8 Metallurgy and Welding

Chapter 9, 10, 11 Inspection & Quality Control

New Building Design Chapters 1-3 Background Information

Chapter 7 Design Criteria

Chapter 8 Metallurgy & Welding

Chapter 9, 10, 11 Inspection & Quality Control

Interim Guidelines:Evaluation, Repair, Modification andDesign of Steel Moment Frames Chapter 2 - Definitions, Abbreviations & Notations

2-1

2. DEFINITIONS, ABBREVIATIONS & NOTATIONS

This Chapter provides the definition of terms used throughout these Interim Guidelines. Inaddition, abbreviations and symbols, used in other sections of the Interim Guidelines are listedhere, together with their typical usage.

2.1 Definitions

As used in this document, the terms defined below shall be interpreted to have the meaningindicated, unless specifically indicated elsewhere in this document to have other meaning in a specificcontext.

2.1.1 Administrative

The definitions of this section apply to the titles of persons involved in the design, construction,regulation, or use of buildings and to the standards, codes and ordinances by which such use isregulated.

Building Code The locally enforced set of regulations governing the design, construction, alteration,occupancy and repair of building structures.

Commentary: Although some municipalities and government agencies develop andmaintain independent building codes, most building construction in the United Statesis regulated under locally adopted editions of one of three model building codes: theUniform Building Code (UBC), the National Building Code (NBC) and the StandardBuilding Code (SBC). The UBC has been used as a model in this advisory becausemost buildings damaged by the Northridge Earthquake were designed under earliereditions of that code, and because the seismic design regulations contained in theother two codes, were until 1993, based on those contained in the UBC. In 1993,both the NBC and SBC adopted seismic design regulations based on the NEHRPRecommended Provisions for the Development of Seismic Regulations for NewBuildings (Building Seismic Safety Council - 1991). Where references to theUBC provisions are contained in these Interim Guidelines, they are generally tothe 1994 edition of that document, unless another edition is specificallyidentified. Where these Interim Guidelines make reference to specific provisionsin the UBC, parallel provisions in the NEHRP Recommended Provisions aregenerally identified in {parentheses}, where parallel provisions exist. Note thatthe formulae and requirements contained in these parallel provisions are notalways identical, and caution should be exercised when referencing the NEHRPRecommended Provisions from these Interim Guidelines.

Building Official That officer or authorized representative who has been appointed with legal authorityto regulate the construction, alteration, occupancy and use of building structures withina recognized state, county, or municipality.


Chapter 2 - Definitions, Abbreviations & Notations Design of Steel Moment Frames

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Building Owner That person, corporation or agency holding legal title to the property being constructed,inspected, or repaired, or persons designated with authority to act on their behalf withregard to the building.

Contract Documents The drawings, specifications and contractual terms under which the responsibilities ofthe various parties in a project to construct or modify a building are defined.

Contractor That corporation, partnership, or person retained by the Building Owner to manageand/or perform construction work on a building.

Engineer of Record The structural engineer in responsible charge of the preparation of drawings andspecifications for the inspection, repair, modification or construction of a structure.

Erector A contractor performing the erection, repair and/or modification of structural steelframes.

Evaluation The process, including preliminary screening, on-site inspection, and structuralanalysis, of determining if a building has been structurally damaged, the effect ofdamage on the building’s integrity, and development of strategies for the occupancy,structural repair and/or modification of the building.

Fabricator A contractor performing fabrication of structural steel elements to be incorporated in astructural steel frame.

Inspection On-site investigation of the condition of a structure (or components of a structure)through direct visual observation, aided as necessary by special non-destructive testingtechniques.

Owner’s Inspector A welding inspector retained by the Building Owner to perform quality assuranceinspections of weldments. The AWS D1.1 Code defines this individual as the“Verification Inspector.”

Peer Review An independent technical review of project construction documents as well assupporting data, calculations and assumptions, conducted by structural engineers andintended to provide the Owner and Engineer of Record with an opinion as to the extentthat the design complies with applicable standards of care and is likely to achieve itsintended objectives.

Special Inspector An Inspector employed by the Building Owner under the requirements of Section 1701of the Building Code. When such person performs special inspections related toweldments, he/she shall possess the qualifications noted for a Welding Inspector.

Structural Engineer A person holding professional engineering registration with the state havingjurisdiction, for the practice of structural engineering. The person should haveparticular training, knowledge and expertise in the structural design of buildings andstructures. In some states such a person may hold registration as a Civil Engineer orProfessional Engineer.


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Welding Code American Welding Society publication ANSI/AWS D1.1-94, Structural WeldingCode - Steel, 1994 Edition.

Welding Engineer A person with particular training, knowledge and expertise in metallurgy, the joining ofmetal elements to each other by the process of welding, and non-destructive testingtechniques.

Welding Inspector A person meeting the requirements of AWS D1.1, Section 6.1.3.1 (and certified byICBO where applicable) to perform inspections of structural steel weldments. In AWSD1.1, this person is known as “Inspector.”

Welder A person qualified to perform welding in accordance with the provisions of AWSD1.1.

2.1.2 Technical

The definitions of this section indicate the terms by which specific structural components andelements are indicated in this document.

Assembly The substructure of a steel frame that occurs at a floor level and consists of a singlecolumn and one or more floor girders and/or beams that attach directly to it.

Backing A material or device placed against the back side of the joint, or at both sides of a weldin electroslag welding, to support and retain molten weld metal. The material may bepartially fused or remain unfused during welding, and may be either metal or nonmetal.

Backup Bar A non-preferred term, in common use, for a steel bar used as backing in a completejoint penetration weld. More appropriate terminology is “steel backing.”

Chord A direct tension or compression element placed at diaphragm edges to resist flexuraldemands on the diaphragm.

Collector A structural element used to accumulate shear forces from a diaphragm and distributethem to vertical elements such as frames or walls located along a common line. Alsosee Strut and Tie.

Connection The attachment of one structural element, for example a beam, to another, for examplea column. As typically used in this document, connection means the attachment of abeam to a column for moment resistance. Important components of this connectioninclude the beam itself, the beam shear tab, the column and its associated panel zone,continuity and doubler plates, and any additional plates used to join these elementstogether. Other types of connections include bracing connections, gravity connections,base plate connections and column splice connections.

Damage Degradation in the strength or stiffness of a structural element or alteration of theconfiguration of the structure or its elements resulting from structural loading, such asinduced by an earthquake.



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Damage Index A numerical index used to quantify the amount of degradation a moment resistingconnection (or a group of moment resisting connections) has experienced. A value of 0indicates no damage and a value of 10, total damage.

Design Basis Earthquake Earthquake ground motion with a probability of exceedance at a site of 10% in 50years. Such ground motions has an average return period of 475 years.

Diaphragm A horizontal (or nearly horizontal) element of the lateral force resisting system used todistribute lateral loads to the vertical elements of the lateral force resisting system.

Drift The total lateral deformation of a structure over its height.

Drift Index Dimensionless quantity indicating the ratio of a structure’s lateral deformation to itsheight.

Dual System A structural system in which lateral load resistance is provided by a moment resistingframe in parallel with one or more braced frames and/or shear walls, and meeting thecriteria of UBC-94 Section 1627.6.5.

Ductility The ability of a material, component, element or structure to deform inelasticallybeyond its yield strength without significant loss in load carrying ability.

Electrode A component of the electrical circuit that terminates at the arc, molten conductive slag,or base metal.

End Dam A small plate located at the edge of a beam flange to column flange joint, orientedperpendicular to the joint and intended to serve as a boundary for weld deposition.

Commentary: End dams are a mis-application of the requirement for weld tabs thatwas adopted by some erectors in Southern California. End dams as such are notmentioned in the AWS D1.1 code and they do not constitute weld tabs as requiredand defined in the code.

Expected Yield Stress The average stress at which material conforming to an ASTM specification will exhibityield behavior, as determined by statistical evaluation of production samples.

Flux A material used to hinder or prevent the formation of oxides and other undesirablesubstances in molten metal and on solid metal surfaces, and to dissolve or otherwisefacilitate the removal of such substances.

Flux-Cored Arc An arc welding process that produces coalescence of metals by heating them with anWelding arc between a continuous filler metal electrode and the work. Shielding is provided by

a flux contained within the tubular electrode. Additional shielding may or may not beobtained from an externally supplied gas or gas mixture.

Fully Restrained A beam to column connection with sufficient rigidity to hold the originalConnection angles between the intersecting members virtually unchanged at loads approaching the

strength of the weakest member.


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Gas Shielded FCAW A flux-cored arc welding process variation in which additional shielding is obtainedfrom an externally supplied gas or gas mixture.

Group A set consisting of those moment resisting connections in a building primarily intendedto resist lateral forces in a given direction of building response, and selected as havingsimilar seismic response characteristics, and therefore, similar probability of beingdamaged in an earthquake

Gravity Connection A connection designed to transmit gravity loads from one structural element to another,but not intended to participate in the lateral force resisting system for the structure.

Heat Affected Zone The portion of the base metal whose mechanical properties or microstructure have beenaltered by the heat of welding, brazing, soldering, or thermal cutting.

Heat Treatment A controlled heating and cooling of a metal, usually involving re-crystallization.

Incipient Root Crack A small planar discontinuity or cracking at the root of a weld.

Interpass Temperature In a multipass weld, the temperature of the weld area between weld passes.

Interstory Drift The lateral deformation of a structure within a given story.

Interstory Drift Index The drift index for a particular story of a structure.

Joint The juncture of one piece of base metal (for example a beam flange) to another (forexample a column flange).

Lamellar Discontinuities Defects in rolled structural shapes or plate, typically consisting of non-metallic sulfideand oxide inclusions which have been flattened by the rolling process and alignedparallel to the direction of rolling.

Lamellar Tear A subsurface terrace and step-like crack in the base metal with a basic orientationparallel to the wrought surface caused by tensile stresses in the through-thicknessdirection of the base metal weakened by the presence of small dispersed, planarshaped, nonmetallic inclusions parallel to the metal surface.

Lateral Force Those elements of a structure which are intended to provide lateral strengthResisting System and stiffness for the resistance of lateral forces due to wind or earthquake.

Liquid Dye A method of NDT in which a highly fluid, red dye penetrant is sprayed on thePenetrant Testing surface of a joint to detect open surface defects. (PT)

Magnetic Particle A method of NDT which uses a flux field and iron powder to detect surfaceTesting and sub-surface discontinuities. (MT)

Magnitude A scale indicating the energy released by an earthquake.

Maximum Capable The most severe ground motion likely to be experienced at a site, given the



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Earthquake known seismologic and geotectonic environment. This may be determined bydeterministic methods in regions with well defined seismic sources, or by probabilisticmethods. If probabilistic methods are used, it may be taken as that level of groundmotion with a 10% probability of exceedance in 100 years. Such ground motion hasan average return period of approximately 1,000 years.

Metallurgical Stress A significant deviation in the mechanical properties (usually hardness andRiser micro-structure) between two adjacent regions in a weldment. These may result from

arc strikes, improperly made tack welds, and improperly prepared thermally cutsurfaces.

Minimum Specified The lower bound of acceptable yield strength permitted by ASTMYield Strength specifications, as measured by simple tensile test in accordance to ASTM

requirements.

Modification A structural alteration intended to improve the strength, stiffness, or energy dissipationcapacity of a structure and/or its elements.

Moment Frame A continuous plane of framing in which the beams are joined to the columns withmoment resisting connections.

Moment Magnitude A scale indicating the energy released by an earthquake. Moment magnitude can becalculated based on the surface area of fault rupture amount of slip across the surface,and the stress drop during the event. For moderate magnitude events (<7) momentmagnitude and Richter or local magnitude are approximately the same. Above thatlevel, moment magnitude is a more accurate representation.

Notch Toughness The ability of a material to absorb energy (usually when loaded dynamically) in thepresence of a flaw.

Ordinary Moment A moment-resisting frame not meeting the requirements of UBC-94 Section 2211.7Resisting Frame

Panel Zone In a moment-resisting beam-column connection, that portion of the column web (orwebs) effective in developing the flexural stresses from the girder(s) through shearbehavior.

Partially Restrained A connection between beams and columns that does not possess sufficientConnection rigidity to hold virtually unchanged the original angles between the members at load

levels approaching the strength of the weaker member.

Peening The mechanical working of metals using impact blows.

Plastic Hinge In a flexural element, that region along a beam’s span at which flexural yieldingoccurs.

Plastic Moment The moment that causes a plastic hinge to form in a flexural member.


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Plastic Rotation The angular deformation which occurs in a plastic hinge, once yielding has initiated.Expressed in radians.

Potentially Hazardous A building declared by the building official to be considered hazardous but not yetevaluated by a structural engineer in accordance with these Interim Guidelines.

Postheating The application of heat to an assembly after welding, brazing, soldering, thermalspraying, or thermal cutting.

Preheat The heat applied to the base metal or substrate to attain and maintain preheattemperature.

Preheat Temperature A specified temperature that the base metal must attain in the welding, brazing,soldering, thermal-strain, or cutting area immediately before these operations areperformed.

Quality Assurance The auditing of the Contractor/Fabricator/Erector’s quality control system andprocedures, usually performed by the Owner’s Inspector or Special Inspector.

Quality Control The Contractor/Fabricator/Erector’s quality program.

Radiographic Testing An NDT process in which X-rays or gamma rays are passed through a weldment toexpose a film, which when developed can indicate the presence of discontinuities anddefects. (RT)

Repair Construction work intended to restore a damaged structure or structural element toapproximately the same configuration, stiffness, and strength that existed prior to theonset of damage.

Rigid Connection See “Fully Restrained Connection”

Runoff Tab A non-preferred usage for “weld tab”.

Self Shielded FCAW A flux-cored arc welding process variation in where shielding is exclusively providedby a flux contained within the tubular electrode.

Semi-Rigid Connection Same as “Partially Restrained Connection”

Shielded Metal Arc An arc welding process that produces coalescence of metals by heating themWelding with an arc supplied between a covered metal electrode and the work. Shielding is

obtained from decomposition of the electrode coating.

Special Moment- A welded moment-resisting frame meeting the requirements of UBC-94 Section 2211.7Resisting Frame

Steel Backing Backing comprised of steel.



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Strength The capacity of a section to resist applied axial loads, shears and/or moments, asindicated in UBC-94 section 2211.4.2

Stress Relief Uniform heating of a structure or a portion thereof to a sufficient temperature to relievethe major portion of the residual stresses, followed by uniform cooling.

Stop Drill Drilling a hole at the end of a crack to stop it from running.

Strut A compressive element, provided to control differential displacements between twoelements of a structural system.

Through Thickness For elements of hot rolled steel shapes and plates, a term referring to stresses or strainsimposed on the element perpendicular to a plane aligned with the direction of rolling.

Tie A tensile element, typically placed in a diaphragm, to provide continuity, but alsoprovided at foundation level to control differential lateral displacements of individualfoundations.

Toughness The ability of a smooth member (unnotched) to absorb energy, usually when loadedslowly.

Ultimate Tensile The maximum load divided by the original cross-sectional area of the specimen.Strength

Ultrasonic Testing An NDT process in which high frequency sound waves are reflected through a materialand recorded by an instrument to indicate the presence of discontinuities (UT).

Welding Specification A specification which sets the general requirements for welding work performed on aproject, including the responsibilities of individuals and the processes which may beused. This specification is part of the contract documents.

Welding Procedure A rigorous written specification of all important welding parameters for aSpecification given welded connection including welding process, material thickness and fit-up of

parts, welding position, electrode type and stick out, voltage, amperage, polarity,preheat and interpass temperatures, etc.

Welded Steel Moment- A plane (or nearly so) frame structure deriving lateral load stability from rigidResisting Frame, interconnection of the beams and columns (WSMF). Rigid connections may consistWelded Steel Moment either of fully welded connections or connectionsFrame which are partially welded and partially bolted. This includes both ordinary moment-

resisting frames (OMRFs) and special moment-resisting frames (SMRFs) as defined inthe Uniform Building Code.

Weld Tab Additional material, upon which a weld may be initiated or terminated.

Yield Stress The average tensile stress during yielding in the plastic range, and/or the stressdetermined in a tension test when the strain reaches 0.005 in. per in.


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Yield Strength The uniaxial tensile stress at which a material exhibits a specified limiting deviationfrom the proportionality of stress to strain. Deviation expressed in terms of strain.

2.2 Abbreviations

When used in this document, abbreviations shall refer to the following terms:

AASHTO American Association of State Highway and Transportation Officials

AASHTO FCP A document published by AASHTO for the fabrication of fracture-critical,non-redundant steel bridges, often called the Fracture Control Plan.

AISC American Institute of Steel Construction

ANSI/AWS D1.1 Structural Welding Code - Steel, published by the American Welding Society

ASNT American Society for Nondestructive Testing

ASTM American Society for Testing and Materials

ATC Applied Technology Council

ATLSS NSF Center on Advanced Technology for Large Structural Systems at LehighUniversity

AWS American Welding Society

CJP Complete Joint Penetration

CUREe California Universities for Research in Earthquake Engineering

EBF Eccentric Braced Frame

EGW Electro Gas Welding

ESW Electro Slag Welding

FEMA Federal Emergency Management Agency

FCAW Flux Cored Arc Welding

FCAW-g Flux Cored Arc Welding - Gas Shielded

FCAW-ss Flux Cored Arc Welding - Self Shielded

GMAW Gas Metal Arc Welding



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HAZ Heat Affected Zone

ICBO International Conference of Building Officials

LAST Lowest Anticipated Service Temperature

MT Magnetic Particle Testing

NEHRP National Earthquake Hazards Reduction Program

NDE Nondestructive Examination.

NDT Nondestructive Testing

NIST National Institute of Standards and Technology

NSF National Science Foundation

OMRF Ordinary Moment-Resisting Frame

PGA Peak Ground Acceleration, horizontal unless otherwise specified.

PT Liquid Dye Penetrant Testing

RT Radiographic Testing

SAC A joint venture of SEAOC, ATC and CUREe

SAW Submerged Arc Welding

SEAOC Structural Engineers Association of California

SMAW Shielded Metal Arc Welding

SMRF Special Moment-Resisting Frame

SSPC Structural Shape Producers Council

UBC Uniform Building Code

UT Ultrasonic testing

VI Visual Inspection

WPS Welding Procedure Specification

WSMF Welded Steel Moment Frame


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W Designation for a wide flange structural shape

WT Designation for a structural “T” section cut from a wide flange cross section

2.3 Notations

α In the design of modifications to existing WSMF structures, a coefficient thataccounts for strain hardening and modeling uncertainty.

β In the design of new WSMF structures, a coefficient to adjust specified yieldstress to an expected mean value for the material grade, and to account forstrain hardening and modeling uncertainty.

σ The standard deviation for the defect indices in a group of inspectedconnections.

b A factor which represents the number of multiples of the standard deviation fora normal distribution above the mean that would be required to exceed adamage index D of 33%.

bf Width of a beam flange - inches

d Damage repair cost for a building, expressed as a % of building replacementcost.

dj A damage index, assigned to connection “j”, and used to determine the overalldamage index “D “within a structure as well as to determine if repair and ormodification is warranted.

davg The mean value of the damage index for a group, considering all connectionsinspected in the group

fa The axial stress in a column

k The total number of connections in a group of connections, at a typical floor inthe group

ki The total number of connections in a group at floor “i”

mi The number of inspected connections in a group at floor “i” includingadditional connections inspected due to their proximity to damagedconnections

p The number of floors in a building

n The number of connections in a group, inspected as part of an initial sample



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s Standard deviation for material strength parameters, based on industry study

tf Thickness of the flange of a beam section - inches.

Aa Effective peak ground acceleration coefficient, contained in building codesbased on the NEHRP Provisions.

D The average damage index for a group of connections at a typical floor.

Di A damage index representing the proportional damage to the connections inthe lateral force resisting system in one group of connections (or in onehorizontal direction) at a floor, based on combining the damage to inspectedconnections at that floor and the average observed damage at other floors, toall uninspected connections in the group at the floor.

Dmax The maximum damage index for a group of connections at any floor.

Fy Specified minimum uniaxial tensile yield stress.

Fya Actual yield stress of component in an existing building

Fye Expected yield stress of component in a new building

Fym Mean yield strength for specified material, based on industry published data

L Width of a frame bay - ft (meters)

L’ Distance between plastic hinges along a beam - ft (meters)

M The magnitude of an earthquake.

Mc The moment demand at the center of the column when a beam mechanism isformed

Me The design moment for a connection.

Mp That bending moment which causes a plastic hinge to form in a flexural elementof a frame, at minimum specified yield stress.

Mpr The bending moment expected to cause a plastic hinge to form in a flexuralelement of a frame, considering the expected yield stress.

Mw The moment magnitude of an earthquake.

P The probability that damage to connections on at least one floor hasresulted in a damage index Di of 1/3 or more.


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Pf The probability that the damage to connections at any floor have resulted ina damage index Di of 1/3 or more.

S The standard deviation for the damage indices of a the set of inspectedconnections belonging to a group, at one floor level.

T Fundamental period of vibration - seconds

TI Fundamental period of vibration of a base isolated structural system -seconds

Z Seismic zone coefficient defined in the UBC, and representative of theeffective peak ground acceleration produced by a design earthquake.

Zb Plastic section modulus of the beam

Zc Plastic section modulus of the column



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Interim Guidelines:Evaluation, Repair, Modification andDesign of Steel Moment Frames Chapter 3 - Damage Classification

3-1

3. CLASSIFICATION AND IMPLICATIONS OF DAMAGE

A broad range of damage and defects were found in steel moment resisting connections, followingthe Northridge Earthquake. Communication between engineers and technicians was oftenconfused, due to the use of different terminology for reporting these conditions. To avoid suchconfusion, a uniform system for damage classification is presented in this Chapter, and referencedthroughout these Interim Guidelines. The implication of each damage type is also discussed.

Some reported damage in WSMF buildings; i.e. local buckling and yielding, was consistent withthe expected behavior of these structures. However, the widespread brittle fractures whichoccurred were inconsistent with previous expectations. This calls into question the ability ofexisting WSMF structures to provide adequate protection of life safety in major earthquakes.

There has never been an earthquake-induced collapse of a WSMF building in the United States.However, the severe damage experienced by some WSMF buildings in the Northridge Earthquakesuggests that collapse is credible given the right combination of building characteristics andground motion. Based on historic evidence, it seems unlikely that earthquakes with magnitudesless than approximately 7 would produce such ground motion, except in the very near field.However, larger events could produce such ground motions over large regions. Therefore, newmoment frame buildings should not continue to be designed and constructed using traditionalmethods.

The risk associated with existing WSMF structures should be assessed in comparison with otherconstruction types. Many other types of buildings have occasionally experienced collapse in pastmoderate earthquakes. Therefore, it seems unwarranted to mandate upgrades of existing WSMFstructures prior to addressing these other building types. However, some building owners maywish to perform such upgrades in order to reduce the risks associated with individual buildings.

The repair costs associated with some WSMF structures, following the Northridge Earthquake,were substantially higher than would previously have been projected. Statistical analysis of datacollected on damaged buildings has been used to project, with varying levels of confidence, thelikely repair costs for such structures, when subjected to ground motion of different severities. Asummary of these statistics is presented to permit estimation of the probable repair costs forWSMF buildings that have experienced different levels of ground motion.

3.1 Summary of Earthquake Damage

Following the Northridge Earthquake, structural damage observed in Los Angeles areaWSMF buildings included yielding, buckling and excessive fracturing of the steel framingelements (beams and columns) and their connections, as well as permanent lateral drift in somestructures. Damaged elements included girders, columns, column panel zones (including girderflange continuity plates and column web doubler plates), the welds of the beam to column flangesand the shear tabs which connect the girder webs to column flanges. There has been speculation


Chapter 3 - Damage Classification Design of Steel Moment Frames

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that column splices and base plates would also be subject to fracture damage, however noinstance of such damage has been reported in WSMF buildings damaged by the NorthridgeEarthquake. There have been reports of such damage in buildings affected by the 1995 Kobe(Great Hanshin), Japan earthquake. Figure 3-1 illustrates the location of these elements.

Frame Elevation

Column splice

Girder

Base Plate

Column

Doubler Plate

Continuity Plate

Weld

Panel Zone

Shear Tab

Figure 3-1 - Elements of Welded Steel Moment Frame

3.2 Damage Types

Damage to framing elements of WSMFs may be categorized as belonging to the weld (W),girder (G), column (C), panel zone (P) or shear tab (S) categories. This section defines a uniformsystem for classification and reporting of damage to elements of WSMF structures, that is utilizedthroughout these Interim Guidelines. The damage types indicated below are not mutuallyexclusive. A given girder-column connection may experience several types of damagesimultaneously. In addition to the individual element damage types, a damaged WSMF may alsoexhibit global effects, such as permanent interstory drifts.

Following a post-earthquake inspection, classification of the damage found, as to its type anddegree of severity is the first step in performing an assessment of the condition and safety of adamaged WSMF structure. In Section 4 these classifications are used for the assignment ofdamage indices. These damage indices are statistically combined and extrapolated to provide anindication of the severity of damage to a structure’s lateral force resisting system and are used as abasis for selecting building repair strategies. Section 6 addresses specific techniques and designcriteria recommended for the repair and modification of the different types of damage.


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Commentary: The damage types contained in this Chapter are based on a systemfirst defined in a statistical study of damage reported in NISTR-5625 (Youssef et.al.- 1994). The original classes contained in that study have been expandedsomewhat to include some conditions not previously identified. Damage classeshave not been standardized within the profession, and many individual engineersand inspection agencies engaged in the inspection and repair of structuresdamaged by the Northridge Earthquake have used other terminologies. It isrecommended that the definitions given below be adopted as the uniform standardfor reporting and classifying damage in the future. This will provide a commonbasis for communication as well as enhance the ability to develop anunderstanding of the performance of WSMF structures in earthquakes.

3.2.1 Girder Damage

Girder damage may consist of yielding, buckling or fracturing of the flanges of girders at ornear the girder-column connection. Eight separate types are defined in Table 3-1. Figure 3-2illustrates these various types of damage. See section 3.2.3 and 3.2.4 for damage to adjacentwelds and shear tabs, respectively.

G1

G2G3

G4

G7

G6

G8

Note: condition G5 consists of types G3 and/or G4 damage occurring at both the top and bottom flanges.

Figure 3-2 - Types of Girder Damage

Table 3-1 - Types of Girder Damage

Type DescriptionG1 Buckled flange (top or bottom)G2 Yielded flange (top or bottom)G3 Flange fracture in HAZ (top or bottom)G4 Flange fracture outside HAZ (top or bottom)G5 Flange fracture top and bottomG6 Yielding or buckling of webG7 Fracture of webG8 Lateral torsional buckling of section



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Commentary: Minor yielding of girder flanges (type G2) is the least significanttype of girder damage. It is often difficult to detect and may be exhibited only bylocal flaking of mill scale and the formation of characteristic visible lines in thematerial, running across the flange. Removal of finishes, by scraping, may oftenobscure the detection of this type of damage. Girder flange yielding, withoutlocal buckling or fracture, results in negligible degradation of structural strengthand typically need not be repaired.

Girder flange buckling (type G1) can result in a significant loss of girderplastic strength. For compact sections, this strength loss occurs gradually, andincreases with the number of inelastic cycles and the extent of the inelasticexcursion. Following the initial onset of buckling, additional buckling will oftenoccur at lower load levels and result in further reductions in strength, comparedto previous cycles. The localized secondary stresses which occur in the girderflanges due to the buckling can result in initiation of flange fracture damage(G4). Once this type of damage occurs, the girder flange may rapidly loose alltensile capacity under continued or reversed loading, however, it may retain somecapacity in compression. Visually evident girder flange buckling should berepaired.

With the conventional structural steels used in WSMF buildings, girder flangecracking within the HAZ (type G3) is most likely to occur at connections in whichimproper welding procedures were followed, resulting in local embrittlement ofthe HAZ. Like the visually similar type G4 damage, it results in a complete lossof flange tensile capacity, and consequently, significant reduction in thecontribution to frame lateral strength and stiffness from the connection. Little G4or G5 damage was actually seen in buildings following the NorthridgeEarthquake. In some cases, this damage was found to extend from the weldaccess hole in the web of the girder, a metallurgically complex area, into theflange. As shown in Figure 3-2, this damage occurs at a location of local flangebuckling, which is where it has been observed in some testing of large-scaleassemblies, after many cycles of load.

In the Northridge Earthquake girder damage has most commonly beendetected at the bottom flanges, although some instances of top flange failure havealso been reported. There are several apparent reasons for this. First thecomposite action induced by the presence of a floor slab at the girder top flange,tends to shift the neutral axis of the beam towards the top flange. This results inlarger tensile deformation demands on the bottom flange than on the top. Inaddition, the presence of the slab tends to greatly reduce the chance of localbuckling of the top flange. The bottom flange, however, being less restrained canexperience buckling relatively easily. Preliminary large-scale testing conductedby SAC included specimens without a slab present. Flange fractures in thesespecimens tended to occur randomly, sometimes at the top and sometimes at the


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bottom flange, somewhat confirming that the slab may have significant influenceon connection behavior.

There are a number of other factors that could lead to the greater incidenceof bottom flange fractures observed in the field. The location of the weld backingis one of the most important of these. At the bottom flange joint, the backing islocated at the extreme tension fiber, while at the top flange, it is located at a pointof lesser stress and strain demand, both due to the fact that it is located on theinside face of the flange and because the floor slab tends to alter the sectionproperties. Therefore, any notch effects created by the backing are more severeat the bottom flange. Another important factor is that welders can typically makethe CJP weld at the girder top flange without obstruction, while the bottom flangeweld must be made with the restriction induced by the girder web. Also thewelder typically has better and more comfortable access to the top flange joint.Thus, top flange welds tend to be of higher quality, and have fewer stress risers,which can initiate fracture. Finally, studies have shown that UT inspection of thetop flange weld is more easily achieved than at the bottom flange, contributing tothe better quality likely to occur in top flange welds.

3.2.2 Column Flange Damage

Seven types of column flange damage are defined in Table 3-2 and illustrated in Figure 3-3.Column damage typically results in degradation of a structure’s gravity load carrying strength aswell as lateral load resistance.. For related damage to column panel zones, refer to Section 3.2.5.

C1

C2

C3

C4

C5

C6

C7

Figure 3-3 - Types of Column Damage



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Table 3-2 - Types of Column DamageType DescriptionC1 Incipient flange crackC2 Flange tear-out or divotC3 Full or partial flange crack outside HAZC4 Full or partial flange crack in HAZC5 Lamellar flange tearingC6 Buckled flangeC7 Column Splice Failure

Commentary: Column flange damage includes types C1 through C7. Type C1damage consists of a small crack within the column flange thickness, typically atthe location of adjoining girder flange. C1 damage does not go through thethickness of the column flange and can be detected only by NDT, such as UT.Type C2 damage is an extension of type C1, in which a curved failure surfaceextends from an initiation point, usually at the root of the girder to column flangeweld, and extends longitudinally into the column flange. In some cases thisfailure surface may emerge on the same face of the column flange where itinitiated. When this occurs, a characteristic “nugget” or “divot” can bewithdrawn from the flange. Types C3 and C4 fractures extend through thethickness of the column flange and may extend into the panel zone. Type C5damage is characterized by a stepped shape failure surface within the thickness ofthe column flange and aligned parallel to it. This damage is often detectable onlywith the use of NDT.

Type C1 damage does not result in an immediate large strength loss to thecolumn; however, such small fractures can easily progress into more serioustypes of damage if subjected to additional large tensile loading by aftershocks orfuture earthquakes. Type C2 damage results in both a loss of effective attachmentof the girder flange to the column for tensile demands and a significant reductionin available column flange area for resistance of axial and flexural demands.Type C3 and C4 damage result in a loss of column flange tensile capacity andunder additional loading can progress into other types of damage.

Type C5 damage may occur as a result of non-metallic inclusions within thecolumn flange. The potential for this type of fracture under conditions of highrestraint and large through-thickness tensile demands has been known for anumber of years and has sometimes been identified as a contributing mechanismfor type C2 column flange through-thickness failures. Many engineers haveadopted a practice of specifying mandatory NDT investigation of column sectionsin the vicinity of girder-column connections, in accordance with ASTM A898,both before and after welding to detect type C5 discontinuities.


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As a result of the potential safety consequences of complete column failure,all column damage should be considered as significant, and repairedaccordingly.

3.2.3 Weld Damage, Defects and Discontinuities

Six types of weld discontinuities, defects and damage are defined in Table 3-3 and illustratedin Figure 3-4. All apply to the CJP welds between the girder flanges and the column flanges.This category of damage was the most commonly reported type following the NorthridgeEarthquake.

Table 3-3 - Types of Weld Damage, Defects and Discontinuities

Type DescriptionW1 Weld root indications

W1a Incipient indications - depth <3/16” or tf/4;width < bf/4

W1b Root indications larger than that for W1aW2 Crack through weld metal thicknessW3 Fracture at column interfaceW4 Fracture at girder flange interfaceW5 UT detectable indication - non-rejectable

W1, W5W2

W3W4

Note: See Figure 3-2 for related column damage and Figure 3-3 for girder damageFigure 3-4 - Types of Weld Damage

Commentary: Type W1 damage, discontinuities and defects and type W5discontinuities are detectable only by NDT, unless the backing bar is removed,allowing direct detection by visual inspection or magnetic particle testing. TypeW5 consists of small discontinuities and may or may not actually be earthquakedamage. AWS D1.1 permits small discontinuities in welds. Larger discontinuitiesare termed defects, and are rejectable per criteria given in the Welding Code. It



3-8

is likely therefore that some weld indications detected by NDT in a post-earthquake inspection may be discontinuities which pre-existed the earthquakeand do not constitute a rejectable condition, per the AWS standards. Repair ofthese discontinuities, designated as type W5 is not generally recommended. Sometype W1 indications are small planar defects, which are rejectable per the AWSD1.1 criteria, but are not large enough to be classified as one of the types W2through W4. Type W1 is the single most commonly reported non-conformingcondition reported in the post-Northridge statistical data survey, and in somestructures, represents more than 80 per cent of the total damage reported. TheW1 classification is split into two types, W1a and W1b, based on their severity.Type W1a “incipient” root indications are defined as being nominal in extent,less than 3/16” deep or 1/4 of the flange thickness, whichever is less, and havinga length less than 1/4 of the flange width. Some engineers believe that type W1aindications are not earthquake damage at all, but rather, previously undetecteddefects from the original construction process. A W1b indication is one thatexceeds these limits but is not clearly characterized by one of the other types. Itis more likely that W1b indications are a result of the earthquake than theconstruction process.

As previously stated, some engineers believe that both type W1a and sometype W1b conditions are not earthquake related damage at all, but instead, arerejectable conditions not detected by the quality control and assurance programsin effect during the original construction. However, in recent large-scale sub-assembly testing of the inelastic rotation capacity of girder-column connectionsconducted in SAC Phase 1 at the University of Texas at Austin and theEarthquake Engineering Research Center of the University of California atBerkeley, it was reported that significantly more indications were detectable inunfailed CJP welds following the testing than were detectable prior to the test.This tends to indicate that type W1 damage may be related to stresses induced inthe structures by their response to the earthquake ground motions. Regardless ofwhether or not type W1 conditions are directly attributable to earthquakeresponse, it is clear that these conditions result in a reduced capacity for the CJPwelds and can act as stress risers, or notches, to initiate fracture in the event offuture strong demands.

Type W2 fractures extend completely through the thickness of the weld metaland can be detected by either MT or VI techniques. Type W3 and W4 fracturesoccur at the zone of fusion between the weld filler metal and base material of thegirder and column flanges, respectively. All three types of damage result in aloss of tensile capacity of the girder flange to column flange joint and should berepaired.

As with girder damage, damage to welds has most commonly been reported atthe bottom girder to column connection, with fewer instances of reported damage


3-9

at the top flange. Available data indicates that approximately 25 per cent of thetotal damage in this category occurs at the top flange, and most often, top flangedamage occurs in connections which also have bottom flange damage. For thesame reasons previously described for girder damage, less weld damage may beexpected at the top flange. However, it is likely that there is a significant amountof damage to welds at the top girder flange which have never been discovered dueto the difficulty of accessing this joint. Later sections of these Interim Guidelinesprovide recommendations for situations when such inspection should beperformed.

3.2.4 Shear Tab Damage

Eight types of damage to girder web to column flange shear tabs are defined in Table 3-4 andillustrated in Figure 3-5. Severe damage to shear tabs is often an indication that other damage hasoccurred to the connection including column, girder, panel zone, or weld damage.

Table 3-4 - Types of Shear Tab Damage

Type DescriptionS1 Partial crack at weld to column

S1a girder flanges soundS1b girder flange cracked

S2 Fracture of supplemental weldS2a girder flanges soundS2b girder flange cracked

S3 Fracture through tab at bolts or severe distortionS4 Yielding or buckling of tabS5 Loose, damaged or missing boltsS6 Full length fracture of weld to column

S1

S2S3

S4

S6 S5

Figure 3-5 - Types of Shear Tab Damage



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Commentary: Shear tab damage should always be considered significant, asfailure of a shear tab connection can lead to loss of gravity load carryingcapacity for the girder, and potentially partial collapse of the supported floor.Severe shear tab damage typically does not occur unless other significant damagehas occurred at the connection. If the girder flange joints and adjacent basemetal are sound, than they prevent significant differential rotations fromoccurring between the column and girder. This protects the shear tab fromdamage, unless excessively large shear demands are experienced. If excessiveshear demands do occur, than failure of the shear tab is likely to trigger distressin the welded joints of the girder flanges.

3.2.5 Panel Zone Damage

Nine types of damage to the column web panel zone and adjacent elements are defined inTable 3-5 and illustrated in Figure 3-6. This class of damage can be among the most difficult todetect since elements of the panel zone may be obscured by beams framing into the weak axis ofthe column. In addition, the difficult access to the column panel zone and the difficulty ofremoving sections of the column for repair, without jeopardizing gravity load support, make thisdamage among the most costly to repair.

Table 3-5 - Types of Panel Zone Damage

Type Description

P1 Fracture, buckle or yield of continuity plateP2 Fracture in continuity plate weldsP3 Yielding or ductile deformation of webP4 Fracture of doubler plate weldsP5 Partial depth fracture in doubler plateP6 Partial depth fracture in webP7 Full or near full depth fracture in web or doublerP8 Web bucklingP9 Severed column

P1

P2

P4

P7 P3 P5, P6

P8P9

Figure 3-6 - Types of Panel Zone Damage


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Commentary: Fractures in the welds of continuity plates to columns (type P2), ordamage consisting of fracturing, yielding, or buckling of the continuity platesthemselves (type P1) may be of relatively little consequence to the structure, solong as the fracture does not extend into the column material itself. Fracture ofdoubler plate welds (type P4) is more significant in that this results in a loss ofeffectiveness of the doubler plate and the fractures may propagate into thecolumn material.

Although shear yielding of the panel zone (type P3) is not by itselfundesirable, under large deformations such shear yielding can result in kinking ofthe column flanges and can induce large secondary stresses into the girder flangeto column flange connection. In recent SAC Phase 1 testing at the University ofCalifornia at Berkeley, excessive deformation of the column panel zone wasidentified as a contributing cause to the initiation of type W2 fractures at the topgirder flange. It is reasonable to expect that such damage could also be initiatedin real buildings, under certain circumstances.

Fractures extending into the column web panel zone (types P5, P6 and P7)have the potential under additional loading to grow and become type P9resulting in a complete disconnection of the upper half of a column from thelower half, and are therefore potentially as severe as column splice failures.When such damage has occurred, the column has lost all tensile capacity and itsability to transfer shear is severely limited. Such damage results in a total loss ofreliable seismic capacity. It appears that such damage is most likely to occur inconnections that are subject to column tensile loads, and/or in which beam yieldstrength exceeds the yield strength of the column material.

Panel zone web buckling (type P8) may result in rapid loss of shear stiffnessof the panel zone with potential negative effects as described above. Suchbuckling is unlikely to occur in connections which are stiffened by the presence ofa vertical shear tab for support of a beam framing into the column’s minor axis.

3.2.6 Other Damage

In addition to the types of damage discussed in the previous sections, other types of structuraldamage may also be found in WSMF buildings. Other framing elements which may experiencedamage include column base plates, beams, columns, and their connections that were not intendedin the original design to participate in lateral force resistance, and floor and roof diaphragms. Inaddition, large permanent interstory drifts may develop in the structures. Based on observationsof structures affected by the Northridge Earthquake, such damage is unlikely unless extensivedamage has also occurred to the lateral force resisting system. When such damage is discoveredin a building, it should be reported and repaired, as suggested by later sections of these InterimGuidelines.



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3.3 Safety Implications

The implications of the damage described above with regard to building safety are discussed inthis section. There is insufficient knowledge at this time to permit determination of the degree ofrisk with any real confidence. However, based on the historic performance of modern WSMFbuildings, typical of those constructed in the United States, it appears that the risk of collapse inmoderate magnitude earthquakes, ranging up to perhaps M7, is low for buildings which have beenproperly designed and constructed according to prevailing standards. A possible exception to thismay be buildings located in the near field (< 10 km from the surface projection of the faultrupture) of such earthquakes (Heaton, et. al. - 1995), however, this is not uniquely a problemassociated with steel buildings. Our current building codes in general, may not be adequate toprovide for reliable performance of buildings within the near field of large earthquakes. As is alsothe case with all other types of construction, buildings with incomplete lateral force resistingsystems, severe configuration irregularities, inadequate strength or stiffness, poor constructionquality, or deteriorated condition are at higher risk than buildings not possessing thesecharacteristics.

No modern WSMF buildings have been sited within the areas of very strong ground motionfrom earthquakes larger than M7, or for that matter, within the very near field for eventsexceeding M6.5. This style of construction has been in wide use only in the past few decades.Consequently, it is not possible to state what level of risk may exist with regard to buildingresponse to such events. This same lack of performance data for large magnitude, long durationevents exists for virtually all forms of contemporary construction. Consequently, there isconsiderable uncertainty in assigning levels of risk to any building designed to minimum coderequirements for these larger events.

Commentary: Research conducted to date has not been conclusive with regard tothe risk of collapse of WSMF buildings. Some testing of damaged connectionsfrom a building in Santa Clarita, California have been conducted at theUniversity of Southern California (Anderson - 1995). In these tests, connectionassemblies which had experienced type P6 damage were subjected to repeatedcycles of flexural loading, while the column was maintained under axialcompression. Under these conditions, the specimens were capable of resisting asmuch as 40 per cent of the nominal plastic strength of the girder for severalcycles of slowly applied loading, at plastic deformation levels as large as 0.025radians. However, damage did progress in the specimen, as this testing wasperformed. It is not known how these assemblies would have performed if thecolumns were permitted to experience tensile loading. Data from other testssuggests that the residual strength of connections which have experienced typesG1, G4, W2, W3, and W4 damage is on the order of 15 per cent of theundamaged strength. Some analytical research (Hall - 1995) in which nonlineartime history analyses simulating the effects of connection degradation due tofractures were included, indicates that typical ground motions resulting in thenear field of large earthquakes can cause sufficient drift in these structures to


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induce instability and collapse. Other researchers (Astaneh - 1995) suggest thatdamaged structures, even if unrepaired, have the ability to survive additionalground motion similar to that of the Northridge Earthquake.

Even though there were no collapses of WSMF buildings in the 1994Northridge Earthquake, it should not be assumed that no risk of such collapseexists. Indeed, a number of WSMF buildings did experience collapse in the 1995Kobe Earthquake. The detailing of these collapsed Japanese buildings wassomewhat different than that found in typical US practice, however, much of thefracture damage that occurred was similar to that discovered following theNorthridge event.

Because of a lack of data and experience with the effects of larger, longerduration earthquakes, there is considerable uncertainty about the performance ofall types of buildings in large magnitude seismic events. It is believed thatseismic risks in such large events are highly dependent on the individual groundmotion at a specific site and the characteristics of the individual buildings.Therefore, generalizations with regard to the probable performance of individualtypes of construction may not be particularly meaningful.

The risks to occupants of WSMF buildings is regarded as less, in most cases,than to occupants of the types of buildings listed below. However, because of theuncertainties involved, the degree of risk in large events cannot be definitivelyquantified, nor can it categorically be stated that properly constructed WSMFbuildings sited in the near field of large events are either more or less at risk thanmany other code designed building systems which do not appear on the followinglist:

• Concentric braced steel frames with bracing connections that are weaker than thebraces

• Knee braced steel frames

• Unreinforced masonry bearing wall buildings

• Non-ductile reinforced concrete moment frames (infilled or otherwise)

• Reinforced concrete moment frames with gravity load bearing elements that werenot designed to participate in the lateral force resisting system and that do nothave capacity to withstand earthquake-induced deformations

• Tilt-up and reinforced masonry buildings with inadequate anchorage of theirheavy walls to their horizontal wood diaphragms

• Precast concrete structures without adequate interconnection of their structuralelements.



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In addition, WSMF structures would appear to have lower inherent seismicrisk than structures of any construction type that:

• do not having complete, definable load paths

• have significant weak and/or soft stories

• have major torsional irregularity and insufficient stiffness and strength to resistthe resulting seismic demands

• minimal redundancy and concentrations of lateral stiffness

These are general statements that represent a global view of systemperformance. As with all seismic performance generalizations, there are manysteel moment frame buildings that are more vulnerable to damage than someindividual buildings of the general categories listed, just as there are many thatwill perform better.

3.4 Economic Implications

This section provides data which may be used to estimate probable repair costs for WSMFbuildings conforming to typical pre-Northridge Earthquake design and construction practices, inthe event that they are affected by future strong earthquake ground motion. This information maybe considered, together with other data, when making investment decisions relative to suchbuildings, or when conducting cost-benefit studies to determine if structural upgrade of existingbuildings is economically justified.

Economic losses resulting from earthquake induced building damage include directs costsresulting from inspection to determine the extent of damage, engineering design fees, actual costsrelated to the structural repairs, demolition and replacement costs for architectural finishes andutilities (that must be removed to allow access for inspection and repair), and repair of damagednon-structural components, as well as indirect costs resulting from loss of use, lost income fromrents that are not collected on spaces vacated during the repair period, and project financingcosts. The loss estimation data provided in this section only includes consideration of the directdamage repair costs. It does not include consideration of indirect costs related to lost rents,interruption of business and similar issues. These indirect costs often result in a greater economicimpact than do the actual costs of repair, but are difficult to estimate on a general basis.Allowance for such indirect costs should be made in any economic analysis conducted forindividual buildings.

The loss estimation data presented in this section is compatible with that presented in ATC-13(Applied Technology Council -1985), a document frequently used as the basis for loss estimationstudies. In that document, vulnerability functions are presented for broad classes of buildings,based on the expert opinion of groups of individuals familiar with the performance of thosestructures. The vulnerability functions relate the expected repair costs, expressed as a percentageof building replacement value, to a ground motion parameter (Modified Mercalli Intensity), and alevel of confidence.


3-15

Table 3-6 presents a proposed vulnerability function for WSMF buildings typical of Californiaconstruction prior to the Northridge Earthquake. Each column of the table provides an estimateof the percentage of the total population of these buildings within a region affected by groundmotion of defined intensity, expected to have repair costs “d”, expressed as a percentage ofbuilding replacement value, within the indicated ranges. Figure 3-7 provides a plot of this data ina format which may be more useful for application to loss estimation estimates. The statisticscontained in the table were calculated using a loss estimation model developed by Thiel andZsutty (Thiel and Zsutty - 1987), and data obtained on the performance of 89 buildings affectedby the Northridge Earthquake (Bonowitz and Youssef - 1995).

Table 3-6 - Estimated Distribution of WSMF Buildings1

by Severity of Damage in Regions of Varying Ground Motion Intensity

Modified Mercalli IntensityDamage d2 VII VIII IX X

d<5% 71% 57% 40% 30%5%<d<25% 21% 29% 34% 35%25%<d<50% 7% 12% 20% 26%50%<d<75% 1% 2% 5% 8%75%<d<100% 0% 0% 1% 1%

1. WSMF buildings conforming to pre-Northridge Earthquake design andconstruction practice for regions of high seismicity (UBC seismic zones 3and 4) {NEHRP Map Areas 6 and 7}.

2. “d’ is the direct damage repair cost, expressed as a percentage of buildingreplacement cost

0 10 20 30 40 50 60 70 80 90 100

0

10

20

3

0

40

50

6

0

70

80

9

0

100

% of Buildings In Population Reaching or Exceeding Repair Cost Percentage

Rep

air C

ost a

s a

% o

f Rep

lace

men

t Cos

t

Legend

MMI VIIMMI VIIIMMI IXMMI X

Figure 3-7 - Vulnerability Estimates for WSMF BuildingsConforming to Typical California Practice Prior to the Northridge Earthquake



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These loss estimation statistics should be used with caution, when applied to individualbuildings. The unique characteristics of any individual building, including the strength andstiffness of its lateral force resisting system, its inherent redundancy, its condition, and the qualityof its construction, will affect the relative vulnerability of the building. The statistics presentedmay be considered as representative of average buildings, in general conformance with theapplicable building code provisions. Buildings that have substantial deficiencies relative to thoseprovisions would be expected to be significantly more vulnerable. Similarly, buildings that havesuperior earthquake resisting characteristics, relative to the requirements of the building code,would be expected to be less vulnerable.

The statistics contained in Table 3-6 were established based on case studies conducted bySAC of the damage experienced by selected buildings affected by the Northridge Earthquake. Itappears that typical repair costs for structural damage to connections can range from about$7,000 per connection to approximately $20,000. These costs are dominated not by the structuralwork, but rather by costs related to mobilizing into discrete areas of the building, performing localdemolition of finishes and utilities as required to gain access and to create a safe workingenvironment, and reconstruction of these finishes and utilities upon completion of the structuralwork. The cost of the structural work itself tends to vary from about $2,000 for the simplestrepairs of damage (type W1 and W2) to perhaps $5,000 or more for repairs of the most complextypes. These cost estimates do not include allowances for hazardous materials abatement, whichwill be required if either asbestos containing materials or lead based paint are present in theoriginal construction. Such materials are likely to be present in buildings constructed prior toabout 1980. The above costs relate only to the restoration of connections. They do not includecosts related to re-establishing vertical plumbness of the building, which may be impractical toaccomplish, or costs related to repair of architectural, mechanical, and electrical componentswhich are directly damaged by the building’s response to the ground motion. These statisticsassume that the building is repaired, rather than demolished and reconstructed. It should be notedthat at least one building, in Santa Clarita, was demolished and reconstructed rather than repaired.A number of factors may have contributed to the owner’s decision to take such action, however,it is clear that the cost associated with this decision was much greater than would be indicated bythe statistics presented in this Section.

Commentary: The damageabilities indicated in Table 3-6 and Figure 3-7 wereestimated based on statistics available on a data set of 89 buildings (Bonowitzand Youssef - 1995). From this data set, it was possible to establish theprobability of a building incurring damage to a given percentage of its totalconnections. This data set also allowed estimation of the number of connectionsper square foot of floor space provided by a building. From these statistics, anestimated average repair cost per connection of $12,500 was applied against theprobable number of damaged connections per square foot of floor space.Building value was taken as $125/square foot of floor space. This computationpermitted calculation of the expected loss percentage to a typical building. Thisdata was then entered into a loss estimation model developed by Thiel and Zsutty(Thiel and Zsutty - 1987). The model was developed to replicate damage statistics


3-17

observed in historic earthquakes and extended to current construction typesusing, in part, the expert opinion results of ATC-13.

Ground motion is characterized in Table 3-6 and Figure 3-7 using ModifiedMercalli Intensity (MMI). Although MMI has been the most common groundmotion parameter used for loss estimation studies in the past, it is subjective andinterpretation can be varied. MMI can only be assigned after an earthquake hasoccurred and is based on observation of damage and other effects that haveactually occurred. It is dependent, to a very great extent, on the types ofconstruction which are present in the affected region. The distributions ofdamage indicated in Table 3-6 and Figure 3-7 are considered appropriate forCalifornia, and other regions with similar seismic design and constructionpractices. However, these data may not be appropriate for other regions.

It should be noted that when the repair cost for a building approaches 60 percent or more of its replacement value (d>60%) many owners will determine,based on a number of factors, that complete building replacement, rather thanrepair is warranted. Therefore, it is probable that the actual costs for repair ofsome buildings will be 100per cent of the replacement value. This possibility hasnot been reflected in the development of the damage repair cost distributionspresented in Table 3-6.

It should also be noted that the statistics used to develop the abovevulnerability estimates were taken from an incomplete data set of buildings. Thedata set may or may not have been representative of the distribution of damage inthe total set of buildings affected by the Northridge Earthquake. If the data set isbiased, this is likely to be a bias towards buildings that are more heavilydamaged, since the data was collected soon after the earthquake, when only thosebuildings most likely to have been damaged had been inspected. A review of theapplicability of the statistics used for generating the vulnerability estimatesshould be conducted, when more complete data on the distribution of damagebecomes available.



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Interim Guidelines:Evaulation, Repair, Modification andDesign of Steel Moment Frames Chapter 4 - Post-Earthquake Evaluation

4-1

4. POST-EARTHQUAKE EVALUATION

Post-earthquake evaluation is the multi-step process conducted following an earthquake to:determine the extent and severity of damage experienced by a building; assess the implications ofdamage with regard to building safety; and determine appropriate occupancy, structural repair andmodification strategies.

Detailed post-earthquake evaluations of buildings are costly. An initial screening (preliminaryevaluation) process is recommended to identify those buildings most likely to have been damaged.Screening criteria include ground shaking severity estimates, proximity to other structures knownto be damaged and significant observable damage to the building itself. Buildings identified byscreening as likely to have been damaged should be subjected to detailed evaluation.

Analyses of damaged buildings show that although damage occurred at slightly higher frequencyin locations predicted to have high strength and deformation demands, damaged connections tendto be widely distributed throughout the building frames, often at locations analyses would notpredict. This approximates a random distribution. To detect all such damage, it would benecessary to subject each connection to detailed inspections. In order to reduce inspection costs,but still reliably detect damage, these Interim Guidelines recommend inspection of representativesamples of connections and the use of statistical techniques to project damage observed in thesamples to that likely experienced by the entire building.

In order to obtain valid projections of a building’s condition, samples should be broadlyrepresentative of the varying conditions (location, member sizes, structural demand) presentthroughout the building and should be sufficiently large to permit confidence in the projection ofoverall building damage. Three alternative methods for sample selection are provided. Whensubstantial damage is found within the sample of connections, additional connections areinspected to provide better, and more reliable information on building condition.

Once the extent of building damage is determined, the structural engineer should assess theresidual structural integrity and safety, and determine appropriate repair and/or modificationactions. General recommendations are provided, based on calculated damage indices. Directapplication of engineering analysis may also be used. For individual structures, the structuralengineer should confirm that the general recommendations are appropriate, based on evaluation ofthe specific structural characteristics of the damaged building and on engineering judgment.

4.1 Scope

This Chapter presents guidelines for:

1) identifying those WSMF structures likely to have been damaged in an earthquake;


Chapter 4 - Post-Earthquake Evaluation Design of Steel Moment Frames

4-2

2) development of a program of inspection for structures suspected of having beendamaged;

3) assessing the implications of discovered damage; and

4) determining appropriate occupancy, repair, and/or structural modification actions toprotect life safety.

Nothing in these Interim Guidelines should be deemed to preempt the judgment of thebuilding official or to prevent individual structural engineers from adopting alternative approachesbased on accepted engineering principles, rational criteria and sound reasoning. However,independent qualified third party review should be considered when such alternative approachesare adopted. Section 4.5 provides recommended criteria for such independent third partyreviews.

Commentary: This Chapter provides a basic approach and suggested criteria forpost-earthquake evaluation. This includes preliminary evaluation to determine ifa building is likely to have been damaged and detailed evaluation to determinethe actual damage experienced and the extent to which the building’s lateral-force-resisting system has been compromised. In the detailed evaluationmethodology, procedures are given for selecting a representative sample ofbuilding connections for inspection, and for interpreting the results of theseinspections. Chapter 5 provides detailed recommendations on how to performinspections. Chapter 6 provides guidance on damage repair as well as structuralmodification to improve future seismic performance.

4.2 Preliminary Evaluation

This section provides recommended criteria for determining which WSMF structures shouldbe subjected to detailed post-earthquake evaluations and suggestions for the scheduling of suchevaluations. It also provides recommendations (Section 4.2.4) for the acceptance of inspectionand evaluation programs performed prior to the publication of these Interim Guidelines.

Following an earthquake, all WSMF structures that experienced ground motion having thepotential to cause structural damage in these buildings, as indicated in Section 4.2.1, should besubjected to a detailed evaluation. Given that a detailed evaluation should be performed for abuilding, this evaluation should be completed prior to:

1) permanent occupancy of a building under construction at the time of the earthquake;

2) reoccupancy of a building closed for post-earthquake repairs that require a buildingpermit; or

3) reoccupancy of a building where occupancy was limited by the building official as aresult of apparent structural damage.


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The results of all building evaluations should be transmitted to the building owner and filedwith the building official as described in Section 4.3.9.

Commentary: This section provides guidelines for building officials andstructural engineers to determine if a WSMF building should be subjected todetailed evaluations. An evaluation includes, as a minimum, assessment as towhether the building has experienced sufficient earthquake-induced structuraldemands to cause damage, and unless this is judged not to be the case, detailedstructural evaluations should be performed. Given the high levels of uncertaintyassociated with the many issues involved in making such judgments, inspectionsshould be performed for any building suspected of having damage, even when theprovisions of these Interim Guidelines or the building official do not so require.It is particularly recommended that all buildings indicated by the preliminaryevaluation of Section 4.2.1 as likely to have been damaged be subjected todetailed evaluations, regardless of whether or not the building official so directs.

4.2.1 Evaluation Process

Preliminary evaluation is the process of determining if a building should be subjected todetailed post-earthquake evaluations. Detailed evaluations should be performed for all buildingsthought to have experienced strong ground motion, as indicated in Section 4.2.1.1 or for whichthe other indicators of Section 4.2.1.2 apply. Detailed post-earthquake evaluations include theentire process of determining if a building has experienced significant damage and if damage isfound, determining appropriate strategies for occupancy, structural repair and/or modification.Except as indicated in Section 4.2.3, detailed evaluation should include inspections of arepresentative sample of moment-resisting (and other type) connections within the building.

4.2.1.1 Ground Motion

Within UBC Seismic Zone 4 {NEHRP Map Area 7}, detailed evaluation is recommended forall WSMF buildings when an earthquake of Magnitude greater than or equal to 6.5 has producedground motion at the building site in excess of 0.20g, or when any earthquake has producedground motion at the building site in excess of 0.30g. For buildings located in zones of lowerseismicity, refer to Table 4-1, Section 4.2.2 for appropriate ground motion thresholds. Wheneverfeasible, ground motion estimates should be based on actual instrumental recordings in the vicinityof the building. When such instrumental recordings are not available, ground motion estimatesmay be based on empirical or analytical techniques. In all cases, ground motion estimates shouldreflect the site-specific soil conditions.

Commentary: A number of techniques are available for estimating thedistribution of ground motion in an area, following an earthquake. In regionswith a large number of strong motion accelerographs present, actual groundmotion recordings produce the best method of mapping contours of groundmotion. In other regions, empirical techniques, such as the use of standard



4-4

ground motion attenuation relationships (e.g. Joyner and Boore - 1994, Campbelland Bazorgnia - 1994) may be required. These can be supplemented withanalytically derived estimates such as those obtained by direct simulation of thefault rupture and ground wave propagation. It should be noted, however, thatlacking direct instrumental evidence, site-specific ground motion estimates are atbest, uncertain, and subject to wide variations depending on the assumptionsmade. Therefore, the best indicator of the severity of ground motion at a site isoften the performance of adjacent construction. The criteria of Section 4.2.1.2are provided to help assure that sites which experienced strong ground motionare not overlooked as a result of inaccurate estimates of the ground motionseverity.

4.2.1.2 Additional Indicators

Regardless of the magnitude of the earthquake event, detailed evaluation should be consideredfor a building if any of the following apply:

1) significant structural damage is observed in one or more WSMF structures locatedwithin 1 kilometer of the building, on sites with similar, or more firm soil profiles;

2) significant structural damage is observed to one or more modern, apparently well-designed structures (of any material) within 1 kilometer of the building and on siteswith similar, or more firm soil profiles;

3) for an earthquake having a magnitude of 6.5 or greater, the structure is either within 5kilometers of the trace of a surface rupture or within the vertical projection of therupture area when no surface rupture has occurred.

4) significant architectural or structural damage is observed in the building;

5) permanent interstory drift greater than 0.5% of story height is observed;

6) unexpected damage, or significant period lengthening of the building are observed inaftershocks; or

7) entry to the building has been limited by the building official because of earthquakedamage, regardless of the type or nature of the damage.

Commentary: In the above, the term “significant” has been used withoutdefinition or quantification. The intent is to use known damage as an indicator ofthe severity of ground motion experienced. Damage is dependent not only on thestrength of ground motion, but also on the quality and condition of the affectedconstruction. Relatively moderate damage to buildings having regularconfiguration and adequate lateral-force-resisting systems may be a moresignificant indicator of strong ground motion than heavy damage to construction


4-5

in poor condition or having other poor earthquake resisting characteristics. Thebuilding official and/or structural engineer should use their own judgment indetermining the significance of such damage.

The absence of significant observable damage to WSMF structures on sitesbelieved to have experienced strong ground motion, per Section 4.2.1, should notbe used as an indication that detailed evaluations are not required. Many WSMFbuildings that were structurally damaged by the Northridge Earthquake had littleapparent damage based on casual observation.

The observed behavior of a building in repeated aftershocks may providesome clues as to whether it has experienced significant structural damage. Ininstrumented buildings it may be possible to observe a period shift in theinstrumented response, as successive damage occurs. In buildings withoutinstruments, the observation of unexpectedly large amounts of architecturaldamage could indicate the presence of structural damage.

In many cases in the past, buildings have initially been posted as unsafewithout adequate investigation of their condition. Upon reconsideration andtechnical evaluation such buildings have subsequently been re-posted to allowoccupancy. In such cases and for the purposes of item 7 above, the building neednot be considered to have been posted.

4.2.2 Evaluation Schedule

When a detailed evaluation of a building is recommended, under Section 4.2.1, suchevaluation should be completed as soon as practical and in any event, within a 12-month periodfrom the date of the earthquake main shock, unless a shorter period is indicated in Table 4-1.

Table 4-1 - Suggested Time Limits for Post-Earthquake Evaluation1

EstimatedPGA Range at Site3

6.0<M<6.5 6.5<M<7.2 7.2<M

PGA>0.40g3 6 months 6 months2 6 months2

0.33<PGA<0.43 12 months 6 months 6 months2

0.23<PGA<0.33 (1) 12 months 12 months

Notes:

1. Evaluation not required unless one or more of the conditions of Section 4.2.1.2 apply.2. Buildings meeting this criteria are likely to have experienced significant damage and evaluations should be

performed rapidly. If NDT technicians are not readily available, visual inspection, in accordance with Section5.2.2 should proceed expeditiously. If in the course of such visual inspection, serious damage to connections isobserved, then consideration should be given to the safety of the occupants in possible aftershocks.

3. The indicated PGA’s are for modern buildings designed to UBC Zone 4 {NEHRP Map Area 7} criteria. Forbuildings designed to other criteria or for other seismic zones, the indicated PGA values should be reduced bythe factor Z/0.4 (for localities that have adopted the UBC) {by the factor Aa/0.4 for localities that have adoptedcodes based on the NEHRP Provisions}. The indicated PGA’s need not be reduced lower than 0.15g.



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Commentary: It is important to conduct post-earthquake evaluations as soonfollowing the earthquake as is practical. Aftershock activity in the monthsimmediately following an earthquake is likely to produce additional strongground motion at the site of a damaged building. If there is adequate reason toassume that damage has occurred, then such damage should be expeditiouslyuncovered and repaired. However, since adequate resources for post-earthquakeevaluation may be limited, a staggered schedule is presented, with those buildingshaving a greater likelihood of damage recommended for evaluation first.

Large magnitude earthquakes are often followed by large magnitudeaftershocks. Therefore, it is particularly urgent that post-earthquake evaluationsbe performed expeditiously following such events. If insufficient resources areavailable in the affected region to perform the NDT tests recommended by theGuidelines of Chapter 5, it is recommended that visual inspection, in accordancewith Section 5.2.2, proceed as soon as possible. If visual inspection revealssubstantial damage, consideration should be given to vacating the building untileither an adequate period of time has passed so as to make the likelihood of verylarge aftershocks relatively low (e.g. 4 weeks for magnitude 7 and lower, and 8weeks for magnitudes above this), complete inspections and repairs are made, ora detailed evaluation indicates that the structure retains adequate structuralstiffness and strength to resist additional strong ground shaking. Preliminaryvisual inspections should not be used as an alternative to complete evaluation.

The table relates the urgency for post-earthquake building evaluation to boththe magnitude of the earthquake and the estimated peak ground accelerationexperienced by the building site. This is because large magnitude events aremore likely to have large magnitude aftershocks and because buildings thatexperienced stronger ground accelerations are more likely to have been damaged.Except in regions with extensive strong motion instrumentation, estimates ofground motion are quite subjective. Following major damaging earthquakes,government agencies usually produce ground motion maps showing projectedacceleration contours. These maps should be used when available. When suchmaps are not available, ground motions can be estimated using any of severalattenuation relationships that have been published.

4.2.3 Connection Inspections

Detailed evaluations should include inspections of a representative sample of WSMF (andother) connections, except as indicated in Sections 4.2.3.1 and 4.2.3.2, below. Section 4.3.3provides three alternative approaches to selecting an appropriate sample of connections forinspection.


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4.2.3.1 Analytical Evaluation

Connection inspections need not be performed for buildings on sites meeting the criteria ofSection 4.2.1.1, if conditions 4,5,6, and 7 of Section 4.2.1.2 are observed not to be present and astructural analysis indicates that estimated stresses in welded moment-resisting connections duringthe earthquake were all below the beam flexural design strength. For calculation of this strength,Fy should be taken as the minimum specified yield strength for the framing members. Suchanalysis should be performed using an estimated ground motion representation (responsespectrum or acceleration time history) similar to that believed to have occurred at the site duringthe earthquake. For the purpose of this analysis, the ground motion and resulting stressescomputed in the various framing elements should not be reduced by the lateral force reductioncoefficients (Rw or R) contained in the building code.

Commentary: This section suggests that unless there is visible evidence that abuilding has been damaged, detailed connection inspections need not beperformed for buildings which can be demonstrated by analysis to haveexperienced very low levels of stress. It will be possible to demonstrate this whenground motions at a site are low, or when the ground motion spectrum at a sitewas such that little excitation would have occurred at predominant modes for thebuilding. A dynamic analysis, using site-specific estimates of the ground motionactually experienced by the building during the earthquake, is required to makesuch a determination. It should be noted that unless a building has beeninstrumented, it is very difficult to estimate the precise ground motions itexperienced, with any accuracy. Since analyses do not provide any conclusiveevidence as to whether a building has actually experienced damage, when the costof such analyses approaches that of inspections, inspections should be performed.

4.2.3.2 Buildings with Enhanced Connections

For buildings designed in accordance with the recommendations of Chapter 7 of these InterimGuidelines, and not displaying any of the conditions 4, 5, 6, or 7 of Section 4.2.1.2, the scope ofinspection may be reduced to 1/2 the number of connections recommended in Section 4.3.3. If inthe course of this reduced scope of inspection, significant structural damage is found (damage toany connection with a damage index per Table 4-3(a or b) that is greater than 5), then fullinspections in accordance with Section 4.3 should be performed.

Commentary: Structures designed in accordance with Chapter 7 of these InterimGuidelines are expected to be less susceptible to connection fractures than WSMFstructures designed with the former prescriptive connection. However, theeffectiveness of these Interim Guidelines in preventing such fractures, during realearthquakes, is not yet known. Therefore, inspection of some connections inbuildings conforming to these Interim Guidelines is recommended, even if there isno obvious evidence of damage.



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4.2.4 Previous Evaluations and Inspections

Many WSMF buildings have been evaluated prior to the publication of these InterimGuidelines. The following approach is recommended for these buildings:

1. The previous evaluation may be considered adequate if any of the following conditionsis met:

a. a building permit has been issued for repair of damaged connections; or

b. the evaluation was performed following procedures contained in SACAdvisory No. 3 (SAC - 1995) and/or City Guidelines in force at the time of theinspection; or

c. the number and distribution of connections inspected substantially complieswith the recommendations of Section 4.3.3, and no connections with damageindices dj (per Table 4-3a or b) greater than 3 were discovered.

2. Previous inspections may be considered adequate and their results interpreted usingthese Interim Guidelines if either of the following conditions is met:

a. the number of connections inspected substantially meets the recommendationsof Section 4.3.3 and the distribution of the inspected connections, as certifiedby the responsible structural engineer, is acceptable to the building official asmeeting the intent of these Interim Guidelines.

b. one and one half times the number of inspections recommended in Section4.3.3 have been performed for each group of connections, regardless of thedistribution of connections within the groups.

When a previous evaluation has been performed that does not meet the conditions of 1 or 2above, the owner should be advised that the previous work does not comply with currentrecommendations and that additional connections should be inspected to provide adequateunderstanding of the building’s condition. The additional connections should be selected so as tobring the total inspection program, including those inspections previously conducted, intosubstantial compliance with the recommendations of Section 4.3.3. Such additional inspectionsshould be performed in a manner that minimizes disruption to building occupancy, but inaccordance with a schedule acceptable to the building official.

Commentary: This section applies to buildings affected by the NorthridgeEarthquake that were evaluated, inspected, and/or repaired, prior to thepublication of these Interim Guidelines. Two different cases are addressed: 1)buildings for which the post-earthquake evaluation/repair process has beencompleted, and 2) buildings which were inspected, but for which evaluationreports and repair actions have not been submitted to and accepted by thebuilding official. If a building was evaluated and subsequently repaired under apermit issued by the building official, or evaluated according to proceduresacceptable to the building official and found not to require repair, then no further


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work is recommended. If a building was inspected, but no report was submitted toor accepted by the building official, then additional work may be appropriate ifthe inspections did not adequately address the condition of the building.

In the months immediately following the Northridge earthquake, divergentopinions were held as to how building inspections should be performed. Somestructural engineers required inspections of every connection in buildings whileothers selected a relatively small sample of connections upon which to performpreliminary inspections, proceeding to more inspections only if significantdamage was found in the initial sample. This latter approach is essentially thesame basic approach adopted by these Interim Guidelines, although theseGuidelines may recommend a larger sample than was commonly used prior totheir publication. By the fall of 1994, many structural engineering offices in theLos Angeles area had adopted the inspection procedure recommendations of aCity of Los Angeles task group. Those suggested that for buildings with 7 storiesor less, 15% of the connections should be included in the initial sample and fortaller structures, 10% of the connections. It was suggested in thoserecommendations that connections be selected on a widely distributed basis, butbiased towards those most likely to have been damaged, as indicated by rationalanalysis. Evaluations of building condition, based on inspections performed inaccordance with those procedures should be deemed to comply with item 1b ofthis section and need not be supplemented by additional work.

If a building has been subjected to minimal inspection, meaning substantiallyless than both the criteria contained in these Interim Guidelines and therecommendations of the City of Los Angeles task group, and the building officialrequires submittal of an evaluation report, it is recommended that additionalinspections be performed to meet the intent of these Interim Guidelines, prior tosubmission of the evaluation report.

If a building is not required to be inspected by the building official, butprevious inspection has been performed at the request of the owner, the structuralengineer should notify the owner if the program of inspection was not insubstantial compliance with these recommendations. Note that under Section4.3.4 of these Interim Guidelines, inspections conducted in a random manner maybe terminated following inspection of 50% of the total number of connectionsrecommended for inspection, if only minimal damage is found. It is anticipatedthat most previously conducted inspection programs, in which minimal damagewas found, would surpass this minimum recommended amount. For those caseswhere this is not so, additional inspection should be performed if the owner andstructural engineer desire reasonable confidence in their knowledge of thecondition of the building. When the inspections were voluntary, and no currentrequirement from the building official exists, any additional inspections desired



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could be performed over a long schedule, so as to result in minimal disruption oftenants.

4.3 Detailed Evaluation Procedure

Where detailed evaluation is recommended by Section 4.2, assessment of the post earthquakecondition of a building, its ability to resist additional strong ground motion and other loads, anddetermination of appropriate occupancy, structural repair and/or modification strategies should bebased on the results of a detailed inspection and assessment of the extent to which structuralsystems have been damaged.

This Section presents one approach for making such assessments. In this approach, the resultsof the inspections are used to calculate a cumulative damage index, D, for the structure as well asthe probability that the damage index at any floor of the structure has exceeded 1/3. Generaloccupancy, structural repair and modification recommendations are made based upon the valuescalculated for these damage indices. In particular, a calculated damage index of 1/3 is used toindicate, in the absence of more detailed analyses, that a potentially hazardous condition mayexist.

The structural engineer may use other procedures consistent with the principles of statisticsand structural mechanics to determine the residual strength and stiffness of the structure in the as-damaged state and the acceptability of such characteristics relative to the criteria contained in thebuilding code, or other rational criteria acceptable to the building official.

Commentary: The most reliable approach to determining the post-earthquakecondition of a building and whether unacceptable damage has occurred would beto inspect and determine the condition of each of the moment-resistingconnections in the structure. This is an expensive process, and is not warrantedunless a structure is heavily damaged. Therefore, these Interim Guidelinesrecommend a process that includes initial inspections of a representative sampleof the welded moment-resisting connections in the building. If the initialinspections indicate a significant amount of damage, then additional inspectionsare recommended. Based upon the observed condition of the total inspectedsample of connections, the probable levels of damage throughout the structureare projected.

In this procedure, each inspected connection is assigned an individualconnection damage index “dj” ranging from 0 to 10. Judgmentally derivedguidelines are provided for the assignment of these damage indices, based on thetypes of damage observed, with 10 indicating very severe damage and loss ofability to reliably participate in the lateral load resisting system, and 0 indicatingno damage. An overall building damage index D is calculated by extrapolatingthe individual connection damage indices “dj” for the connections actuallyinspected to the total number of connections in the structure. In this way, the


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damage index D represents in a very approximate and rough manner, anindication of the loss of reliable capacity of the structure to resist future strongground motion. A structural damage index of 1/3 has arbitrarily been taken as anindication, in the absence of more detailed analyses, that a potentially hazardouscondition may exist.

The procedure presented in these Interim Guidelines to estimate the level ofdamage does not include direct calculation of the remaining lateral strength andstiffness of the damaged building, or its residual displacement capacity, nor doesit attempt to compare these characteristics to the requirements of the buildingcode for new construction. Such an approach, if properly performed, should bevery useful in assisting the structural engineer to understand the probable futureperformance of the building. However, it is important to note that no consensushas been reached yet with regard to appropriate modeling assumptions for theresidual strength and stiffness of damaged connections. Also, unless allconnections within a building have been inspected, the true condition of thebuilding is subject to considerable uncertainty. Consequently, when such anapproach is taken, qualified independent third party review, in accordance withSection 4.5, is strongly recommended.

It is recognized that many WSMF buildings have lateral strength and stiffnessconsiderably in excess of that required by the building code. When analysesindicate that connection damage results in a building that still has more strengthand stiffness than is required by the code, structural engineers should be cautiousin making judgments that there is no requirement for structural repair or furthermodifications Such an approach could be permitted by these Interim Guidelinesif such an engineering analysis is performed, and the building official approves.However, if a large number of building connections have been damaged, this mayindicate the presence of conditions likely to result in excessive damage in futureearthquakes, such as poor quality construction or an unfavorable configuration.Therefore, buildings which have experienced substantial damage should becarefully considered for repair and upgrade, regardless of their pre-earthquakedesign lateral strength and stiffness.

4.3.1 Eight Step Evaluation Procedure

Post-earthquake evaluation should be carried out under the direct supervision of a structuralengineer. The following eight-step procedure may be used to determine the condition of thestructure and to develop occupancy, repair and modification strategies:

Step 1: The moment-resisting connections in the building are categorized into two or more“groups” (Section 4.3.2 and 4.4) comprised of connections expected to have similarprobabilities of being damaged.



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Complete steps 2 through 7 below, for each group of connections.

Step 2: Determine the minimum number of connections in each group that should be inspectedand select the specific sample of connections to be inspected. (Section 4.3.3)

Step 3: Inspect the selected set of connections using the technical guidelines of Section 5.2.and determine connection damage indices, dj, for each inspected connection (Section4.3.4)

Step 4: If inspected connections are found to be seriously damaged, perform additionalinspections of connections adjacent to the damaged connections. (Section 4.3.5)

Step 5: Determine the average damage index (davg) for connections in each group, and then theaverage damage index at a typical floor. (Section 4.3.6)

Step 6: Given the average damage index for connections in the group, determine theprobability, P, that the connection damage index for any group, at a floor level,exceeds 1/3, and determine the maximum estimated damage index for any floor, Dmax.(Section 4.3.7)

Step 7: Based on the calculated damage indices and statistics, determine appropriateoccupancy, structural repair and modification strategies (Section 4.3.8). If deemedappropriate, the structural engineer may conduct detailed structural analyses of thebuilding in the as-damaged state, to obtain improved understanding of its residualcondition and to confirm that the recommended strategies are appropriate or tosuggest alternative strategies.

Step 8: Report the results of the inspection and evaluation process to the building official andbuilding owner. (Section 4.3.9)

Sections 4.3.2 through 4.3.9 indicate how these steps should be performed.

Commentary: Following an earthquake structural engineers and techniciansqualified to perform these evaluations may be at a premium. Prudent owners maywant to consider having an investigation plan already developed (Steps 1 and 2)before an earthquake occurs, and to have an agreement with appropriatestructural engineering and inspection professionals and organizations to givepriority to inspecting their buildings rapidly following the occurrence of anearthquake.

4.3.2 Step 1— Categorize Connections by Groups

The welded moment-resisting connections participating in the lateral-force-resisting systemfor the building are categorized into a series of “connection groups.” Each group consists ofconnections expected to behave in a similar manner (as an example, a group may consist of all


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those connections that are highly stressed by lateral forces applied in a given direction). As aminimum, two groups of connections should be defined - each group consisting of connectionsthat primarily resist lateral movement in one of two orthogonal directions. Additional groupsshould be defined to account for unique conditions including building configuration, constructionquality, member size, grade of steel, etc., that are likely to result in substantially differentconnection behavior, as compared to other connections in the building. Each connection in thebuilding should be uniquely assigned to one of the groups, and the total number of connections ineach group determined.

In buildings that have significant torsional irregularity, it may be advisable to define at leastfour groups— one group in each orthogonal direction on each side of an assumed center ofresistance. Section 4.4 gives a procedure for defining groups where damage may accentuatetorsional response, or where the structural engineer desires a more reliable characterization of thebuilding’s degree of damage. Such procedures should be considered when a building hassignificant torsional irregularity or when there is so little redundancy that failure of one connectionat a floor level would exacerbate a torsional response.

For buildings of two or more stories, the roof connections may be excluded from the initialinspection process. However, when Table 4-5 recommends inspection of all connections within agroup or building, they should be inspected.

Commentary: Many base plates of columns in moment frames use the same basicconnection detail as do the beam/column connections. When such base plates arenot within the cast-in-place concrete floor and grade-beam system, thenconsideration should be given to their inspection. There is evidence from the1995 Kobe earthquake that column splice damage can occur, with resultingsevere impacts on the building’s stability. Consideration should be given by thestructural engineer to their inspection as well. Although these connections shouldalso be inspected, they should not be included within the statistical calculationscontained in this eight-step procedure. Any damage to such connections, shouldbe repaired.

4.3.3 Step 2— Select Samples of Connections for Inspection

Assign a unique identifier to each connection within each group. Consecutive integeridentifiers are convenient to some of the methods employed in this Section.

For each group of connections, select a representative sample for inspection in accordancewith any of Methods A, B, or C, below. A letter indicating the composition of the groups, andthe specific connections to be inspected should be submitted to the building official prior to theinitiation of inspection. The owner or structural engineer may at any time in the investigationprocess elect to investigate more connections than required by the selected method. However,the additional connections inspected may not be included in the calculation of damage statistics



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under Step 5 (Section 4.3.6) unless they are selected in adherence to the rules laid out for theoriginal sample selection, given below.

Commentary: The purpose of inspection plan submittal prior to the performanceof inspections is to prevent a structural engineer, or owner, from performing agreater number of inspections and reporting data only on those which provide afavorable economic result with regard to building disposition. The buildingofficial need not perform any action with regard to this submittal other than tofile it for later reference at the time the structural engineer’s evaluation report isfiled. During the inspection process, it may be decided to inspect additionalconnections to those originally selected as part of the sample. While additionalinspections can be made at any time, the results of these additional inspectionsshould not be included in the calculation of the damage statistics, in Step 5, astheir distribution may upset the random nature of the original sample selection.If the additional connections are selected in a manner which preserves thedistribution character of the original sample, they may be included in thecalculation of the damage statistics in Step 5.

4.3.3.1 Method A - Random Selection

Connections are selected for inspection such that a statistically adequate random sample isobtained. The minimum number of connections to be inspected for each group is determined inaccordance with Table 4-2. The following limitations apply to the selection of specificconnections:

1. Up to a maximum of 20% of the total connections in any sample may be pre-selectedas those expected by rational assessment to be the most prone to damage. Acceptablecriteria to select these connections could include:

• Connections shown by a rational analysis to have the highest demand-capacityratios or at locations experiencing the largest drift ratios.

• Connections which adjoin significant structural irregularities and which thereforemight be subjected to high localized demands. These include the followingirregularities:

- re-entrant corners

- set-backs

- soft or weak stories

- torsional irregularities (connections at perimeter columns)

- diaphragm discontinuities

• Connections incorporating the largest size framing elements.


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2. The balance of the sample should be selected randomly from the remainingconnections in the group.

Up to 10% of the connections in the sample may be replaced by other connections in thegroup to which access may more conveniently be made.

Table 4-2 - Minimum Sample Size for Connection Groups

Number of connectionsin Group1

Minimum number ofconnections to be

inspected

Number of Connectionsin Group1

Minimum number ofconnections to be

inspected6 2 200 2710 3 300 3715 4 400 4520 5 500 5330 7 750 7240 8 1000 9950 10 1250 10475 13 1500 120100 17 2000 147

Note: 1. For other connection numbers use linear interpolation between values given, rounding up to the next highest integer.

Commentary: The number of connections needed to provide a statisticallyadequate sample depends on the total number of connections in the group. Thesample sizes contained in Table 4-2 were developed from MIL-STD-105D, a wellestablished quality control approach that has been widely adopted by industry.

If relatively few connections within a group are expected, the standarddeviation for the computed damage index will be large. This may result inprediction of excessive damage when such damage does not actually exist. Thestructural engineer may elect to investigate more connections than the minimumindicated in order to reduce the standard deviation of the sample and moreaccurately estimate the total damage to the structure. These additionalinspections may be performed at any time in the investigative process. However,care should be taken to preserve the random characteristics of the sample, so thatresults are not biased either by selection of connections in unusually heavy (orlightly) damaged areas of the structure.

It is recognized that in many cases the structural engineer may wish to pre-select those connections believed to be particularly vulnerable. However, unlessthese pre-selected connections are fairly well geometrically distributed, a numberthat is more than about 20% of the total sample size will begin to erode thevalidity of the assumption of random selection of the sample. If the structuralengineer has a compelling reason for believing that certain connections are most



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likely to be damaged, and that more than 20% should be pre-selected on thisbasis, the alternative approach of Method C should be used.

It is recognized that there is often a practical incentive to select connectionsthat are in specific unoccupied or more accessible areas. It is suggested that nomore than 10% of the total sample be composed of connections pre-selected forthis reason. These connections, rather than having a higher disposition fordamage, might well have a lower than average tendency to be damaged. Anexcessive number of this type of pre-selected connection would quickly invalidatethe basic assumption of random selection. It is also recognized that during theinspection process conditions will be discovered that make it impractical toinspect a particular connection, e.g., the architectural finishes are moreexpensive to remove and replace than in other areas, or a particular tenant isunwilling to have their space disturbed. However, as discussed above, not morethan 10% of the total connections inspected should be selected based onconvenience.

There are a number of methods available for determining the randomlyselected portion of the sample. To do this, each connection in the group(excluding pre-selected connections) should be assigned a consecutive integeridentifier. The sample may then be selected with the use of computer spread sheetprograms - many of which have a routine for generation of random integersbetween specified limits, published lists of random numbers, or by drawing oflots.

4.3.3.2 Method B - Deterministic Selection

Connections are selected to satisfy the following criteria:

1. At least one connection is selected on every column face of every line of moment-resisting framing in the group;

2. At least one connection is selected on every floor from every frame;

3. No more than 50% of the connections in a sample may be selected from any floor orcolumn face than would be done if the number of inspected connections was equallyapportioned among either the column faces or floors; and

Up to 10% of the connections in the sample may be replaced by other connections in the sameframe and group to which access may more conveniently be made.

Commentary: It is recognized that in many cases the structural engineer may bereluctant to select connections in a random manner, as provided by Method A.For those cases, Method B is acceptable since it assures that every floor andevery column is inspected at least once. The structural engineer may select any


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combination of connections to be inspected that meets these criteria;notwithstanding, care should be exercised to assure that these allowances are notused to subvert the intent of the inspection process to determine the degree ofdamage to the building, if any.

4.3.3.3 Method C - Analytical Selection

Connections are selected for inspection in accordance with the following criteria:

1. The minimum number of connections within the group to be inspected is as indicatedin Table 4-2.

2. Up to 60% of the connections may be selected based on the results of rational analysisindicating those connections most likely to be damaged.

3. The remaining connections in the group to be inspected are selected such that thesample contains connections distributed throughout the building, including upper,middle and lower stories.

Prior to initiation of the inspections, the rational analysis and list of connections to beinspected should be subjected to a qualified independent third party review in accordance withSection 4.5. The peer review should consider the basis for the analysis, consistency of theassumptions employed, and to assure that overall, the resulting list of connections to be inspectedprovides an appropriate sampling of the building’s connections.

During the inspection process, up to 10% of the connections in the sample may be replaced byother connections to which access may more conveniently be made. Substitution for more than10% of the connection sample may be made provided that the independent third party reviewerconcurs with the adequacy of the resulting revised sample.

Commentary: In analyses conducted of damaged buildings, there has been agenerally poor correlation of the locations of damage and the locations of highestdemand predicted by the analysis. However, there has been some correlation.Analysis is a powerful tool to assist the structural engineer in understanding theexpected behavior of a structure. The specific analysis procedure used should betailored to the individual characteristics of the building. It should includeconsideration of all building elements that are expected to participate in thebuilding’s seismic response, including, if appropriate, elements not considered tobe part of the lateral-force-resisting system. The ground motion characteristicsused for the analysis should not be less than that required by the building codefor new construction, and to the extent practical, should contain the spectralcharacteristics of the actual ground motion experienced at the site. Qualifiedindependent review is recommended to assure that there is careful considerationof the basis for the selection of the connections to be inspected and that arepresentative sample is obtained.



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4.3.4 Step 3— Inspect the Selected Samples of Connections

Inspect the selected samples of connections in each group as indicated in Chapter 5.Characterize the damage at each inspected connection as described in Section 4.3.4.1

Inspections may be terminated when at least 50% of the connections selected for each samplehave been inspected if:

1) the inspections have progressed in a manner that retains an adequately random natureand distributed geometry for those connections that are inspected (a distributionthroughout the building that is acceptable to the building official); and

2) no connections with damage indices dj > 5 (Table 4-3a or b) are discovered; and,

3) not more than 10% of the total connections inspected are discovered to have dj > 2.

If all of these conditions are not met, then inspections should be completed for all connectionscontained in all samples.

Commentary: The sample size suggested for inspection in the methods of Section4.3.3 are based on full inspection using both visual (Section 5.3.1) and NDTtechniques (Section 5.3.2) at all connections in the sample. Other methods ofselection and inspection may be used as provided in Section 4.3, with theapproval of the building official. One such approach might be the visual-onlyinspection of the bottom girder flange to column connection, but with theinspection of a large fraction of the total connections in the group, possiblyincluding all of them. If properly performed, such an inspection procedure woulddetect almost all instances of the most severe damage but would not detect welddefects (W1a), or root cracking (W1b), nor lamellar damage in columns (C5).The occurrence of a few of these conditions, randomly scattered through thebuilding would not greatly affect the assessment of the building’s post-earthquakecondition, or the calculation of the damage index. However, if a large number ofsuch defects were present in the building, this would be significant to the overallassessment. Therefore, such an inspection approach should probably includeconfirming NDT investigations of at least a representative sample of the totalconnections investigated. If within that sample, significant incidence of visuallyhidden damage is found, then full NDT investigations should be performed, assuggested by these Interim Guidelines. Similarly, if visual damage is found at thebottom flange, then complete connection inspection should be performed todetermine if other types of damage are also present.

4.3.4.1 Damage Characterization

Characterize the observed damage at each of the inspected connections by assigning aconnection damage index, dj, obtained either from Table 4-3a or Table 4-3b. Table 4-3a presents


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damage indices for individual classes of damage and a rule for combining indices where aconnection has more than one type of damage. Table 4-3b provides combined indices for themore common combinations of damage.

Table 4-3a - Connection Damage Indices

Type Location Description1 Index2djG1 Girder Buckled Flange 4G2 Girder Yielded Flange 1G3 Girder Top or Bottom Flange fracture in HAZ 8G4 Girder Top or Bottom Flange fracture outside HAZ 8G5 Girder Top and Bottom Flange fracture 10G6 Girder Yielding or Buckling of Web 4G7 Girder Fracture of Web 10G8 Girder Lateral-torsional Buckling 8C1 Column Incipient flange crack (detectable by UT) 4C2 Column Flange tear-out or divot 8C3 Column Full or partial flange crack outside HAZ 8C4 Column Full or partial flange crack in HAZ 8C5 Column Lamellar flange tearing 6C6 Column Buckled Flange 8C7 Column Fractured column splice 8W1a CJP weld Minor root indication - thickness <3/16” or tf/4; width < bf/4 1W1b CJP weld Root indication - thickness > 3/16” or tf/4 or width > bf/4 4W2 CJP weld Crack through weld metal thickness 8W3 CJP weld Fracture at girder interface 8W4 CJP weld Fracture at column interface 8W5 CJP weld Root indication— non-rejectable 0S1a Shear tab Partial crack at weld to column (beam flanges sound) 4S1b Shear tab Partial crack at weld to column (beam flange cracked) 8S2a Shear tab Crack in Supplemental Weld (beam flanges sound) 1S2b Shear tab Crack in Supplemental Weld (beam flange cracked) 8S3 Shear tab Fracture through tab at bolt holes 10S4 Shear tab Yielding or buckling of tab 6S5 Shear tab Damaged, or missing bolts4 6S6 Shear tab Full length fracture of weld to column 10P1 Panel Zone Fracture, buckle, or yield of continuity plate3 4P2 Panel Zone Fracture of continuity plate welds3 4P3 Panel Zone Yielding or ductile deformation of web3 1P4 Panel Zone Fracture of doubler plate welds3 4P5 Panel Zone Partial depth fracture in doubler plate3 4P6 Panel Zone Partial depth fracture in web3 8P7 Panel Zone Full (or near full) depth fracture in web or doubler plate3 8P8 Panel Zone Web buckling3 6P9 Panel Zone Fully severed column 10



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Notes To Table 4-3a:1. See Figures 3-2 through 3-6 for illustrations of these types of damage.2. Where multiple damage types have occurred in a single connection, then:

a. Sum the damage indices for all types of damage with d=1 and treat as one type. If multiple types stillexist; then:

b. For two types of damage refer to Table 4-3b. If the combination is not present in Table 4-3b and thedamage indices for both types are greater than or equal to 4, use 10 as the damage index for theconnection. If one is less than 4, use the greater value as the damage index for the connection.

c. If three or more types of damage apply and at least one is greater than 4, use an index value of 10,otherwise use the greatest of the applicable individual indices.

3. Panel zone damage should be reflected in the damage index for all moment connections attached to thedamaged panel zone within the assembly.

4. Missing or loose bolts may be a result of construction error rather than damage. The condition of the metalaround the bolt holes, and the presence of fireproofing or other material in the holes can provide clues to this.Where it is determined that construction error is the cause, the condition should be corrected and a damageindex of “0” assigned.

Table 4-3b - Connection Damage Indices for Common Damage Combinations1

Girder, Columnor Weld Damage

Shear TabDamage

DamageIndex


Shear TabDamage

DamageIndex

G3 or G4 S1a 8 C5 S1a 6S1b 10 S1b 10S2a 8 S2a 6S2b 10 S2b 10S3 10 S3 10S4 10 S4 10S5 10 S5 10S6 10 S6 10

C2 S1a 8 W2, W3, or W4 S1a 8S1b 10 S1b 10S2a 8 S2a 8S2b 10 S2b 10S3 10 S3 10S4 10 S4 10S5 10 S5 10S6 10 S6 10

C3 or C4 S1a 8S1b 10S2a 8S2b 10S3 10S4 10S5 10S6 10

1. See Table 4-3a, footnote 2 for combinations other than those contained in this table.


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More complete descriptions (including sketches) of the various types of damage are providedin Section 3.1. When the engineer can show by rational analysis that other values for the relativeseverities of damage are appropriate, these may be substituted for the damage indices provided inthe tables. A full reporting of the basis for these different values should be provided to thebuilding official, upon request.

Commentary: The connection damage indices provided in Table 4-3 (rangingfrom 0 to 10) represent judgmental estimates of the relative severities of thisdamage. An index of 0 indicates no damage and an index of 10 indicates verysevere damage.

When initially developed, these connection damage indices wereconceptualized as estimates of the connection’s lost capacity to reliablyparticipate in the building’s lateral-force-resisting system in future earthquakes(with 0 indicating no loss of capacity and 10 indicating complete loss ofcapacity). However, due to the limited data available, no direct correlationbetween these damage indices and the actual residual strength and stiffness of adamaged connection was ever made. They do provide a convenient measure,however, of the extent of damage that various connections in a building haveexperienced.

4.3.5 Step 4— Inspect Connections Adjacent to Damaged Connections

Perform additional inspections of moment-resisting connections near connections withsignificant damage as follows:

1) when a connection is determined to have a damage index dj > 5, inspect all moment-resisting connections immediately adjacent (above and below, to the left and right) tothe damaged connection in the same moment frame (See Figure 4-1). Also inspect anyconnections for beams framing into the column in the transverse direction at that floorlevel, at the damaged connection.

2) when a connection is determined to have a damage index dj > 9, inspect the twomoment-resisting connections immediately adjacent (above and below, to the left andright) to the damaged connection in the same moment frame (See Figure 4-2). Alsoinspect any connections for beams framing into the column in the transverse directionat that floor level at the damaged connection.



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Frame Elevation Floor Plan

Damaged moment - resisting connection with dj > 5Adjacent moment - resisting connection - to be inspectedTransverse connection - to be inspected

Figure 4-1 - Inspection of Connections Adjacent to Damaged Connection (dj > 5)

Frame Elevation Floor Plan

Damaged moment - resisting connection with dj > 9Adjacent moment - resisting connection - to be inspectedTransverse connection - to be inspected

Figure 4-2 - Inspection of Connections Adjacent to Damaged Connection (dj > 9)

Assign damage indices, dj, per Section 4.3.3, to each additional connection inspected. Ifsignificant damage is found in these additional connections (dj > 5), then inspect the connectionsnear these additional connections, as indicated in 1) and 2) above. Continue this process, untilone of the following conditions occurs:

a) The additional connection inspections do not themselves trigger more inspections, or

b) All connections in the group have been inspected.

The results of these added connection inspections, performed in this step are not included inthe calculation of average damage index davg per Section 4.3.6 but are included in the calculation


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of the maximum likely damage index Dmax and probability of excessive damage, P, per Section4.3.7.

4.3.6 Step 5— Determine Average Damage Index for Each Group

For each group of connections, determine the estimated average value of the damage indexfor the group (davg) and its standard deviation (σ) from the equations:

d1n

d

10avgj

j 1

n

==∑

σ 2 =−

−

=

∑1n 1

d

10dj

avgj 1

n 2

(4-1)

(4-2)

where: “n” is the number of connections in the sample selected for inspection under step 2(Section 4.3.3), and

“dj” is the damage index, per Table 4-3 for the “jth” inspected connection in the sample

The additional connections selected using the procedure of Section 4.3.5 (Step 4) are notincluded in the above calculation.

4.3.7 Step 6— Determine the Probability that the Connections in a Group at a Floor Level Sustained Excessive Damage

Two procedures are provided. The first procedure (Section 4.3.7.1) is used in the typicalcase, when some connections in the group have not been inspected. In this case, the maximumdamage index at a floor “Dmax” is estimated based on the damage indices determined for theconnections actually inspected, and the probability “P” that Dmax exceeds a value of 1/3 isdetermined. The second procedure (Section 4.3.7.2) is used when all connections in the grouphave been inspected. In this case, the maximum damage index at any floor “Dmax” can becalculated directly from the known values of the damage indices of the inspected connections.

4.3.7.1 Some Connections in Group Not Inspected

If some connections in the group have not been inspected, determine the expected maximumdamage index at a floor “Dmax” and the probability that at least one floor has a damage indexexceeding 1/3.

First determine the average damage index at a typical floor “D” and its standard deviation “S”from the equations:

D d avg=

Sk

= σ(4-3)

(4-4)



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where “k” is the total number of connections (both inspected and not inspected) in the group at a typical floor.

Then, determine the probability P that the set of connections within the group at any floor hashad a cumulative damage index that is greater than or equal to 1/3. This may be done by usingthe parameters D and S to calculate a factor “b”, which represents the number of multiples of thestandard deviation of a Normal distribution above the mean that would be required to exceed 1/3.The factor “b” is calculated from the equation:

( )b 1 / 3 D S= − (4-5)

Using the value of “b” calculated from equation 4-5, determine Pf, from Table 4-4. Pf is theprobability that if all connections had been inspected, the cumulative damage index at any floorwould have been found to exceed 1/3. This strongly suggests the possibility that there has been areduction in seismic resisting capacity of a similar amount.

Then determine the probability P that if all connections within the group had been inspected,the connections within the group on at least one floor (out of “q” total floors in the group) wouldhave been found to have a cumulative damage index of 1/3 or more from the equation:

( )P 1 1 Pfq= − − (4-6)

Table 4-4 - Pf as a function of b

b Pf - (%) b Pf - (%)-1.2816 90 1.2265 11-0.8416 80 1.2816 10-0.5244 70 1.3408 9-0.2533 60 1.4051 80.0000 50 1.4395 7.50.2533 40 1.4758 70.5244 30 1.5548 60.8416 20 1.6449 50.8779 19 1.7507 40.9154 18 1.8808 30.9542 17 1.9600 2.50.9945 16 2.0537 21.0364 15 2.1701 1.51.0803 14 2.3263 11.1264 13 3.0962 .11.1750 12 3.7190 .01

* Note - Intermediate values of Pf may be determined by linear interpolation

Finally, for each floor “i” in the group for which an inspection has been performed, determinethe cumulative damage index, “Di”, from the equation:


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( )D

k m d

k1k

d

10ii i avg

i i

j

j 1

m i

=−

+

=∑ (4-7)

where: ki is the total number of connections in the group at floor “i”mi is the number of inspected connections in the group at floor “i” including

the additional connections inspected under step 4

Take “Dmax” as the largest of the “Di” values calculated for each floor of the group.

4.3.7.2 All Connections in Group Inspected

If all connections in a group have been inspected, determine the damage index for each floor“i” in the group from the equation:

D1k

d

10ii

j

j 1

k i

==∑ (4-8)

where: “ki” is the total number of connections in the group at floor “i”

Take “Dmax” as the maximum of the “Di” values calculated for each floor of the group.

Commentary: The criterion for damage evaluation used in this Guideline is toassume that a cumulative damage index of 1/3 marks the threshold at which astructure may become dangerous. Such a damage index could correspond tocases where 1/3 of the connections in a building have been severely damaged;cases where all of the connections have experienced moderate damage; or somecombination of these, and therefore represents a reasonable point at which tobegin serious consideration of a building’s residual ability to withstandadditional loads.

Given the current limited understanding of steel moment frame damage, theprobability distribution for connection damage is not known. However, since thedamage index for a floor is the sum of the damage indices for each connection atthe floor, then, by the Central Limit Theorem, as the number of connectionsincreases, the distribution tends to a normal distribution, regardless of the formof the distribution for individual connections. Therefore, the probability that adamage index of 1/3 has been exceeded at a floor, in a group with k connectionsmay be approximated by determining how many multiples (“b”) times thestandard deviation (S), when added to the mean damage index (D) equals 1/3.Or, in equation form :

D + bS = 1/3 (4-9)

Solution of this equation for the multiplier “b” results in the requiredrelationship of equation 4-5.



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Damage Indices (from Table 4-3) that are largely judgmental are used tocharacterize the loss of reliable seismic performance capability of individualconnections. These indices are added, averaged and otherwise statisticallymanipulated for use as an indication of the average damage index for groups ofconnections, entire frames and ultimately of the lateral system itself. It should beclear that use of such an approximate, judgmentally defined characterization ofstrength cannot rigorously calibrate the loss of lateral resistance, or the residualstrength and stiffness of the building.

In spite of the somewhat arbitrary nature of the 1/3 damage index criterionand the judgmental nature of the suggested way of testing whether that criteriahas been exceeded, it is believed that the results of these procedures will lead toreasonable conclusions in most cases. However, it is always the prerogative ofthe responsible structural engineer to apply other rational techniques, such asdirect analyses of the remaining structural strength, stiffness, and deformationcapacity as a verification of the conclusions provided by these procedures.Particularly in anomalous or marginal cases, such additional checks based onengineering judgment are strongly encouraged .

4.3.8 Step 7— Determine Recommended Recovery Strategies for the Building

Recommended post-earthquake recovery strategies are as indicated in Table 4-5, based on thecalculated damage indices and statistics determined in the previous steps. For those groups inwhich all connections have been inspected, the statistic P in the table is neglected.

Table 4-5 - Recommended Repair and Modification Strategies

Observation6 Recommended Strategy (Cumulative) NoteP>0 or Dmax>0 Repair all connections discovered to have dj > 5 1,2P > 5% or Dmax > 0.1 Repair all connections discovered to have dj > 2 1,2P > 10 % or

Dmax> 0.2Inspect all connections in the group. Repair all connections withdj > 2

2

P > 25 % orDmax> 0.33

A potentially unsafe condition may exist. Carefully evaluate theearthquake resistance of the building and the safety of itsoccupants and if not satisfied that adequate vertical stability,lateral strength and stiffness exists, notify the building owner ofthe potentially unsafe condition. Inspect all connections in thebuilding. Repair all connections with dj > 1. Considermodification of all repaired connections and others as appropriate.

3

Dmax > 0.50 An unsafe condition probably exists. Notify the building owner ofthis unless more detailed evaluations indicate otherwise. Inspectall connections in the building. Repair all damaged connectionsand modify all connections for better performance, or modify thebuilding’s lateral-force-resisting system for improvedperformance.

4,5


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Notes to Table 4-5:

1. Includes damage discovered either as part of Step 2 or Step 3.2. Although repair is recommended only for the more seriously damaged connections, the repair of all

connections that are damaged or otherwise deficient should be considered.3. The determination that an unsafe condition may exist should continue until either:

a. full inspection reveals that the gravity system is not compromised, and that the damage index atany floor does not exceed 1/3, or

b. detailed structural analyses indicate that a dangerous condition does not exist, orc. recommended repairs are completed for all connections having dj > 3.

4. An unsafe condition probably exists. The building is almost certainly too severely damaged toprovide adequate occupant safety in a strong earthquake. The structural engineer should eitherrecommend that the building be vacated, or, alternatively, demonstrate by analysis that the risks tooccupant safety, while repairs are conducted, are acceptable. If a decision is made to accept the short-term risks of continued occupancy, an independent third party review of the basis of this decision isrecommended.

5. Repairs required to the building are extensive. In addition to repair, strong consideration should begiven to performing systematic modifications of the building’s lateral-force-resisting system toprovide more reliable performance in the future.

6. The more restrictive observation governs the recommendation. If all connections in the group wereinspected, than do not apply the criteria pertaining to P.

Commentary: The value of P (the probability that the connections on at least onefloor have a cumulative damage index of 1/3 or more) and Dmax (the maximumdamage index at a floor level within a group) were determined in Method A byusing a random selection process, and thus represent a statistically valid basis forthe characterization of the damage index for the group of connections, and thusfor the building. Method B selects the connections by using a specifieddistribution throughout the building based on forcing selection of connections inevery column line and floor. Method C selects the connections, based onengineering characterization of those most likely to have been damaged, modifiedto reflect a distribution throughout the structure. While the connections selectedby Methods B and C are not truly random, they are widely distributed and havesome characteristics of a random distribution. Such selections are judged to besufficiently “random-like” to warrant processing as if the connections wereselected randomly. Thus regardless of whether method A, B, or C was used,decisions on disposition of the building, and the need for repair measures candefensibly be based on the values of these two key parameters, as determined foreach group of connections.

For buildings that have experienced relatively limited levels of damage, Table4-5 recommends repair of damaged connections, without further modification.This is not intended to indicate that buildings that experienced only slight damagehave been demonstrated to be seismically rugged. In fact, if a buildingexperienced light damage as a result of being subjected to relatively low levels ofground motion, it may have substantial vulnerability. This recommendation ismade based on economic considerations and the fact that modification ofbuildings which are only slightly damaged entails a significant increase in the



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required investment. It should be made clear to the owner of such buildings thateven an undamaged or fully repaired welded steel moment frame building stillcarries risk of damage, and to an uncertain extent, risk to life safety in subsequentlarge earthquakes.

When damage is moderate (Dmax < 33%) consideration should be given tomodification of those connections which are being repaired, to provide improvedreliability in the future. However, the structural engineer is cautioned thatmodification of only those connections which have been damaged couldunintentionally create an undesirable condition such as a weak story or torsionalirregularity. Therefore, care should be taken that such conditions are not createdby connection modifications. Modification of the entire structural lateral force-resisting system is strongly recommended when Dmax>0.50. This is not becausethe extent of damage indicates that the building is particularly vulnerable,although this may be the case, but because the work required to repair thebuilding is extensive enough that a relatively small incremental investment willallow substantial improvement in the building’s future potential performance.

If a decision to structurally modify a building is made, and it can bedemonstrated that the structural modifications will reduce the earthquakedemands on the existing WSMF connections from the original design levels, itmay be acceptable not to repair some conditions. In such cases, analyses shouldbe performed to demonstrate the adequacy of the modified structure assumingeither that the affected connections have no moment-resisting capacity, or byincluding an estimate of their reliable post-elastic behavior in the damaged state.In no case should conditions that affect the gravity load-carrying capacity of thestructure be left unrepaired.

Recommendations to close a damaged building to occupancy should not bemade lightly, as such decisions will have substantial economic impact, both onthe building owner and tenants. A building should be closed to occupancywhenever, in the judgment of the structural engineer, damage is such that thebuilding no longer has adequate lateral-force-resisting capacity to withstandadditional strong ground shaking, or if gravity load carrying elements of thestructure appear to be unstable.

4.3.9 Step 8 - Evaluation Report

When an evaluation of a WSMF building has been performed, the responsible structuralengineer should prepare a written evaluation report and submit it to the owner, upon completionof the evaluation. When the building official has required evaluation of a WSMF building, thisreport should also be submitted to the building official. This report should directly or by attachedreferences, document the inspection program that was performed, provide an interpretation of the


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results of the inspection program, and a recommendation as to appropriate repair and occupancystrategies. The report should include but not be limited to the following material:

1) Building Address

2) A narrative description of the building indicating plan dimensions, number of stories,total square feet, occupancy, the type and location of lateral-force-resisting elements.Include a description of the grade of steel specified for beams and columns, and ifknown, the type of welding (SMAW, FCAW, etc.) present. Indicate if momentconnections are provided with continuity plates. The narrative description should besupplemented with sketches (plans and elevations) as necessary to provide a clearunderstanding of pertinent details of the building’s construction. The descriptionshould include an indication of any structural irregularities, as defined in the BuildingCode.

3) A description of nonstructural damage observed in the building, especially as relates toevidence of the drift or shaking severity experienced by the structure.

4) If a letter was submitted to the building official before the inspection process wasinitiated, indicating how the connections were divided into groups and the specificconnections to be inspected; a copy of this letter should be included.

5) A description of the inspection and evaluation procedures used, includingdocumentation of all instructions the inspectors, and of the signed inspection forms foreach individual inspected connection.

6) A description, including engineering sketches, of the observed damage to the structureas a whole (e.g. - permanent drift) as well as at each connection, keyed to the damagetypes in Table 4-3; photographs should be included for all connections with damageindex dj>5. (Refer to Section 5.3.5)

7) Calculations of davg, Di, and Dmax for each group, and if all connections in a groupwere not inspected, Pf and P.

8) Calculations demonstrating the safety of the building where Dmax > 33% and thestructural engineer has determined that an unsafe condition does not exist.

9) A summary of the recommended actions (repair and modification measures andoccupancy restrictions). Any recommendations which represent significant departuresfrom the requirements of Section 4.3.8 should be carefully and completely explained.

The report should include identification of any potentially hazardous conditions which wereobserved, including corrosion, deterioration, earthquake damage, pre-existing rejectableconditions, and evidence of poor workmanship or deviations from the approved drawings. Inaddition, the report should include an assessment of the potential impacts of observed conditions



4-30

on future structural performance. The report should include the Field Inspection Reports ofdamaged connections, as an attachment and should bear the seal of the structural engineer inresponsible charge of the evaluation.

Commentary: Following completion of the detailed damage assessments, thestructural engineer should prepare a written report. The report should includeidentification of any potentially hazardous conditions which were observed,including earthquake damage, pre-existing rejectable conditions, and evidence ofpoor workmanship or deviations from the approved drawings. In addition, thereport should include an assessment of the potential impacts of observedconditions on future structural performance. The report should include the fieldinspection, visual inspection and NDT records, data sheets, and reports asattachments.

The nature and scope of the evaluations performed should be clearly stated.If the scope of evaluation does not permit an informed judgment to be made as tothe extent with which the building complies with the applicable building codes, oras to a statistical level of confidence that the damage has not exceeded anacceptable damage threshold, this should be stated.

4.4 Alternative Group Selection for Torsional Response

This Section provides an optional procedure to that of Section 4.3.2, Step 1, that may beappropriate in selected situations where the structural engineer wants more reliable determinationof the building’s susceptibility to excessive torsional response. If a building responds in atorsionally dominated manner, one side of the structure may experience substantially moredamage than the other side. Such a situation would result in a building that is even moresusceptible to torsional response in future strong ground shaking. In the group selectionprocedure of Section 4.3.2, the connections on opposite sides of a building are included in thesame group. If the building responds torsionally, connections on one side will experience moredamage and connections on the other side less damage, but the average damage statisticscalculated for the group will mask this behavior. In this optional procedure a connection group isestablished on each side of the building’s center of lateral resistance so that if one side of thebuilding has experienced greater damage, due to torsional response, this will be detected by thedamage statistics calculated for the different groups. Typically, under this procedure, at least 4groups of moment-resisting connections will be designated for the building, one on each of thenorth, south, east and west sides of the center of rigidity. Buildings with unusual plan shapes(triangular, hexagonal, etc.) may require more (or possibly fewer) groups of connections toadequately capture torsionally induced damage.

For each group of connections, the following assumptions are made:

1. All of the connections in a group are expected to perform in the same statisticalmanner;


4-31

2. The probability of damage to each connection is uniform over the group, that is, allconnections have the same probability of failure; and,

3. Prior to inspection, whether an individual connection in the group is damaged or not isindependent of the damage state of any other connection in the group.

The number of groups should be increased as is required to meet these objectives.

To reflect torsional response, resulting either from the structural response characteristics ofthe undamaged building or a chance concentration of damage that creates such an imbalance, eachmoment-resisting frame connection is assigned to a group according to the following procedure:

a. Determine the approximate center of rigidity for torsional response of the first floor(assuming the building is in its pre-earthquake, original condition). Draw twoorthogonal lines in the plan principal directions of the moment frames and extend thesevertically as planes. These planes should be adjusted so that all of the connectionsalong a given structural frame are assigned to the same group and that all frames onhigher floors are unambiguously assigned to a group. Where the seismic system doesnot have an orthogonal system, the principal axes can be drawn skewed, or asappropriate to give approximately equal classes of connections assigned to one or theother directions. The following discussion assumes a building with principalorthogonal axes aligned with the north-south and east-west directions.

b. All of the connections providing north-south lateral force resistance and located to thewest of the center of resistance on all floors (and expected to perform in a similarmanner) are assigned to the same group (No. 1). Both weak and strong axisconnection connections are included. Similarly all of the connections providing north-south lateral force resistance and located to the east of the center of resistance areassigned to a second group. A similar procedure is followed to assign connectionsproviding east-west lateral force resistance to one of two additional groups.

c. Sample selection from these groups may be made by any of Methods A, B, or C. Inkeeping with the suggestion in Section 4.3.3.1 paragraph 1, several of the connectionsin each group having the greatest distance from the assumed center of rotation shouldbe included in each sample.

Commentary: It is well known that torsion can play an important role in thedistribution of loads on a building’s frame. The eccentricity of the damagedbuilding, either by its design or the chance occurrence of damage to individualconnections, has major implications for its response in future earthquakes. It isalso clear that the building’s response in orthogonal directions is important.Therefore, for buildings with moment frames in both principal directions, it isrecommended that the investigation procedure include at least four distinctgroups of connections to reflect the torsional and orthogonal loading conditions.



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For buildings with moment frames in only one direction, it is recommended thatthe investigation procedure have at least two distinct groupings of connections.

4.5 Qualified Independent Engineering Review

Independent third party review, by qualified professionals, is recommended throughout theseInterim Guidelines when alternative approaches to evaluation or design are taken, or whereapproaches requiring high degrees of structural engineering knowledge and judgment are taken.Specifically, it is recommended that qualified engineering review be provided in any of thefollowing cases:

1. Where an engineer elects to select connections for inspection by a method other thanMethods A or B of Section 4.3 of these Interim Guidelines.

2. Where the calculated damage index Dmax exceeds 33% and the engineer hasdetermined that an unsafe condition does not exist.

3. Where an engineer has decided not to repair damage otherwise recommended to berepaired by these Interim Guidelines.

4. When any story of the building has experienced a permanent lateral drift exceeding 1%of the story height and proposed repairs do not correct this condition.

5. When an engineer elects to design connections for plastic rotation capacitiesdetermined by analysis.

6. When an engineer elects to design connection configurations by calculations only,without the use of, or reference to, qualification tests for a connection prototype.

Where independent review is recommended, the analysis and/or design should be subjected toan independent and objective technical review by a knowledgeable reviewer experienced in thedesign, analysis, and structural performance issues involved. The reviewer should examine theavailable information on the condition of the building, the basic engineering concepts, and therecommendations for proposed action.

Commentary: The independent reviewer may be one or more persons whosecollective experience spans the technical issues anticipated in the work. Whenmore than one person is collectively performing the independent review, one ofthese should be designated the review chair, and should act on behalf of the teamin presenting conclusions or recommendations.

Independent third party review is not a substitute for plan checking. It isintended to provide the structural engineer of record with an independentopinion, by a qualified expert, on the adequacy of structural engineeringdecisions and approaches. The seismic behavior of WSMF structures is now


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understood to be an extremely complex issue. Proper understanding of theproblem requires knowledge of structural mechanics, metallurgy, welding,fracture mechanics, earthquake engineering and statistics. Due to our limitedcurrent state of knowledge, even professionals who possess such knowledge faceconsiderable uncertainty in making design judgments. Third party review shouldnot be performed by unqualified individuals.

4.5.1 Timing of Independent Review

The independent reviewer(s) should be selected prior to the initiation of substantial portions ofthe design and/or analysis work that is to be reviewed, and review should start as soon assufficient information to define the project is available.

4.5.2 Qualifications and Terms of Employment

The reviewer should be independent from the design and construction team. The reviewershould have no interest of any kind with the work being reviewed other than the performance oftasks required by this section.

a. The reviewer should have no other involvement in the project before, during, or afterthe review.

b. The reviewer should be selected and paid by the owner and should have an equal orhigher level of technical expertise in the issues involved than the structural engineer ofrecord.

c. The reviewer (or in the case of peer review teams, the review chair) should be astructural engineer who is familiar with governing regulations for the work beingreviewed.

d. The reviewer should serve through completion of the project and should not beterminated except for failure to perform the duties specified herein. Such terminationshould be in writing with copies to the building official, owner, and the structuralengineer-of-record.

4.5.3 Scope of Review

Review activities related to evaluation of the safety condition of a building should include areview of available construction documents for the building, all inspection and testing reports, anyanalyses prepared by the structural engineer of record, the method of connection sample selectionand visual observation of the condition of the structure. Review should include consideration ofthe proposed design approach, methods, materials and details.



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4.5.4 Reports

The reviewer should prepare a written report to the owner and building official that covers allaspects of the structural engineering review performed including conclusions reached by thereviewer. Such reports should include statements on the following:

a. Scope of engineering review performed with limitations defined.

b. The status of the project documents at each review stage.

c. Ability of selected materials and framing systems to meet performance criteria withgiven loads and configuration.

d. Degree of structural system redundancy, ductility and compatibility, particularly inrelation to lateral forces.

e. Basic constructability of structural members and connections (or repairs andmodifications of these elements).

f. Other recommendations that would be appropriate to the specific project.

g. Presentation of the conclusions of the reviewer identifying any areas which needfurther review, investigation and/or clarifications.

h. Recommendations, if any.

4.5.5 Responses and Corrective Actions

The structural engineer-of-record should review the report from the reviewer and developcorrective actions and other responses as appropriate. Changes during the construction/fieldphases that affect the seismic resistance system should be reported to the reviewer in writing foraction and recommendations.

4.5.6 Distribution of Reports

All reports, responses and corrective actions prepared pursuant to this section should besubmitted to the building official and the owner along with other plans, specifications andcalculations required. If the reviewer is terminated by the owner prior to completion of theproject, then all reports prepared by the reviewer, prior to such termination, should be submittedto the building official, the owner, and the structural engineer-of-record within (10 ) ten workingdays of such termination.

4.5.7 Engineer of Record

The structural engineer-of-record should retain the full responsibility for the structural designas outlined in professional practice laws and regulations.


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4.5.8 Resolution of Differences

If the structural engineer-of-record does not agree with the recommendations of the reviewer,then such differences should be resolved by the building official in the manner specified in theapplicable Building Code.



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Interim Guidelines:Evaluation, Repair, Modification andDesign of Steel Moment Frames Chapter 5 - Post-Earthquake Building Inspection

5-1

5. POST-EARTHQUAKE INSPECTION

Post-earthquake inspection is that part of the post-earthquake evaluation process that isconducted at the building. It includes detailed visual observation of the condition of the entirestructure as well as selected individual connections and elements. Visual observation is theprimary tool for determining the damage sustained by the structure. It should be supplemented bynon-destructive testing techniques as required to detect damage that is not directly observable.

The moment-resisting connections to be inspected should be determined in accordance withChapter 4. In addition, other potentially vulnerable connections should also be inspected,particularly when evidence of damage is found in the observation of overall building condition, orin the inspection of moment-resisting connections.

Inspection should be conducted under the supervision of a structural engineer familiar with theissues involved. When lower tier personnel are used to perform the inspections, the structuralengineer should ascertain that they have adequate knowledge of the types of damage likely to beencountered, and the indicators as to its existence.

Careful recording and reporting of the results of inspections is critical to the process. Damageshould be reported using the standard classification system of Section 3.1. Care must be taken toaccurately report the location as well as the type and degree of damage, and since damage canincrease as the building is subjected to additional loads, the date at which observations weremade.

When required by the building official, or recommended by the Interim Guidelines in Chapter4, post-earthquake inspections of buildings may be conducted in accordance with the InterimGuidelines of this Chapter. An appropriate sample (or samples) of WSMF connections should beselected for inspection in accordance with the Chapter 4 Guidelines. These connections, andothers deemed appropriate by the engineer, should be subjected to visual inspection (VI) and non-destructive testing (NDT) as required by this Chapter.

5.1 Connection Types Requiring Inspection

5.1.1 Welded Steel Moment Frame (WSMF) Connections

The inspection of a WSMF connection should include the complete joint penetration (CJP)groove welds connecting both top and bottom beam flanges to the column flange, including thebacking bar and the weld access holes in the beam web; the shear tab connection, including thebolts, supplemental welds and beam web; the column's web panel zone, including doubler plates;and the continuity plates and continuity plate welds (See Figure 3-1).

Commentary: The largest concentration of reported damage following theNorthridge Earthquake occurred at the welded joint between the bottom girder


Chapter 5 - Post-Earthquake Building Inspection Design of Steel Moment Frames

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flange and column, or in the immediate vicinity of this joint. To a much lesserextent, damage was also observed in some buildings at the joint between the topgirder flange and column. If damage at either of these locations is substantial (dj

per Chapter 4 greater than 5) then damage is also commonly found in the panelzone or shear tab areas.

These Interim Guidelines recommend complete inspection, by visual and NDTassisted means, of all of these potential damage areas for a small representativesample of connections. This practice is consistent with that followed by mostengineers in the Los Angeles area, following the Northridge Earthquake. Itrequires removal of fireproofing from a relatively large surface of the steelframing, which at most connections will be undamaged.

Some engineers have suggested an alternative approach consisting of visual -only inspections, limited to the girder bottom flange to column joint, but for avery large percentage of the total connections in the building. These bottomflange joint connections can be visually inspected with much less fireproofingremoved from the framing surfaces. When significant damage is found at theexposed bottom connection, then additional fireproofing is removed to allow fullexposure of the connection and inspection of the remaining surfaces. Theseengineers feel that by inspecting more connections, albeit to a lesser scope thanrecommended in these Interim Guidelines, their ability to locate the most severeoccurrences of damage in a building is enhanced. These engineers use NDTassisted inspection on a very small sample of the total connections exposed toobtain an indication of the likelihood of hidden problems including damage types.

If properly executed, such an approach can provide sufficient information toevaluate the post-earthquake condition of a building and to make appropriateoccupancy, structural repair and/or modification decisions. It is important thatthe visual inspector be highly trained and that visual inspections be carefullyperformed, preferably by a structural engineer. Casual observation may missclues that hidden damage exists. If, as a result of the partial visual inspection,there is any reason to believe that damage exists at a connection (such as smallgaps between the CJP weld backing and column face), then complete inspectionof the suspected connection, in accordance with the recommendations of theseInterim Guidelines should be performed. If this approach is followed, it isrecommended that a significantly larger sample of connections than otherwiserecommended by these Interim Guidelines, perhaps nearly all of the connections,be inspected.


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5.1.2 Gravity Connections

In addition to the sample of moment-resisting connections recommended for inspection inChapter 4, it may be appropriate to inspect selected gravity connections. These include gravityconnections for:

1. beams framing orthogonally into a WSMF within the zone of influence of particularWSMF connections with significant damage, and

2. beams framing parallel to a WSMF where significant permanent drift has occurred.

Inspection should include any shear tabs, clip angles, or similar elements and the welds and/orbolts attaching these elements to the beam and supporting framing member.

Commentary: If little or no damage is found to the moment-resisting connectionsin a building, it is probable that the gravity connections have not sustained anysignificant damage. However, if substantial damage is found to moment-resistingconnections, some inspection of the gravity connections in the zone of influence ofthe more heavily damaged moment-resisting connections is probably warranted. For beams framing orthogonally into a WSMF, the zone of influence includesthose beams framing directly into columns with damaged connections, as shownin Figures 4-1 and 4-2. It also includes any other beams that could haveexperienced large torsional rotations as a result of flexural rotations experiencedby the WSMF members they frame to. For beams aligned parallel with theWSMF, this zone of influence includes any portion of the structure likely to haveexperienced excessive drift, as indicated by the damaged moment connection.

5.1.3 Other Connection Types

The structural engineer should review the need to inspect a representative sample of otherconnection types that exhibit negative attributes similar to the CJP beam-to-column weldconfiguration.

Commentary: These negative attributes include: the inherent residual stressconcentrations caused by the welding sequence of highly restrained CJP groovewelds used to connect WSMF beams and columns, and the particular carerequired during their execution to ensure that the welds have no material defects;the post-yield straining in the through-thickness direction of CJP welds used tojoin WSMF beams and columns; the post-yield straining in the through-thicknessdirection of WSMF column flanges in a tri-axial state of stress; the difficulty ofexecuting the WSMF beam's bottom flange CJP weld through the restrictioncreated by the web access hole; and the potential for creating a stress riser byleaving the steel backing (backing bar) in place after completing the CJP weld. Connections that are potential high priority candidates for inspection because oftheir similar connection and stiffness configuration, and because of their use of



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highly restrained CJP welds include certain eccentric braced frame (EBF)configurations, column-to-base plate connections, and certain drag and collectorelements.

In addition, selected column splices located such that stresses on the weldduring the earthquake response likely approached the minimum specified yieldstrength should be inspected, including complete joint penetration welded splicesin Group 4 and 5 shapes and partial penetration groove welded joints for allshape groups. Complete joint penetration flange welds in Group 4 and 5 Sectionshave demonstrated a vulnerability to brittle fracture under gravity loadconditions. Partial joint penetration groove welds have an inherent “notch” orstress-riser condition which can serve as the initiation point for fracture underconditions of high tensile stress demands.

5.2 Preparation

5.2.1 Preliminary Document Review and Evaluation

5.2.1.1 Document Collection and Review

Prior to performing an inspection, the original construction drawings should be reviewed (ifavailable) to identify the primary lateral and gravity load-resisting systems, typical detailing,presence of irregularities, etc. Pertinent available engineering and geotechnical reports, includingprevious damage survey reports and current ground motion estimates should also be reviewed. Specifications (including the original Welding Procedure Specifications), shop drawings, erectiondrawings, and construction records need not be reviewed.

5.2.1.2 Preliminary Building Walk-Through.

A walk-through should be conducted to note visible structural and nonstructural damage,deviations from the plans, and other conditions not evident from the document review.

Commentary: If a preliminary post-earthquake evaluation has not previouslybeen conducted, one should be performed at this time. A preliminarypost-earthquake evaluation based on ATC-20 (Applied Technology Council -1989), or a similar standard, will not necessarily indicate that damage has beensustained.

5.2.1.3 Structural Analysis

A detailed structural analysis of the building need not be performed prior to performingbuilding inspections. At the engineer’s discretion, such analyses may be performed, in order todevelop an understanding as to which connections in the building are most critical and to theextent possible, an understanding of where damage may have concentrated. Analyses used for


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this purpose should be based on rational principles of engineering mechanics and to the extentpossible, should use an actual representation of the ground motion experienced by the building.

Commentary: Detailed analytical studies of buildings damaged by theNorthridge Earthquake indicated some correlation between the actual occurrenceof damage and predicted connections with high demands. However, thiscorrelation was not large enough to warrant strong recommendations thatanalyses be performed prior to performing inspections. In fact, these analysesshowed that it is important to inspect connections throughout the structure,regardless of the demands predicted by analysis. The Interim Guidelines forselecting a representative sample of connections for inspection, presented inChapter 4, contain two methods, A and B, which do not require any prior analysisof the structure, other than to identify its structural system and the location ofmoment-resisting connections. Some engineers may feel that structural analysesare beneficial in developing a program of inspection, and will prefer to select asample of connections for inspection based on such analyses. Sample selectionMethod C, in Chapter 4, is provided for engineers who prefer such an approach.

Any rational method of analysis, including linear static, linear dynamic andnonlinear methods may be utilized. When performing dynamic analyses, it isimportant to use a representation of the ground motion that reasonably resemblesthat likely to have been experienced by the building, as opposed to a generalsmoothed response spectrum. The sharp peaks of response which occur overnarrow bands of frequencies in actual ground motion recordings can accentuatehigher mode response in some buildings, which may not be adequately detectedusing generic smoothed spectra. Analyses of the response of taller buildingsaffected by the Northridge Earthquake, as well as their damage patterns, suggestthat higher mode effects had a significant impact on the locations of severestrength and deformation demands, as well as damage.

The most reliable method of obtaining a representative ground motion is touse data directly recorded by instrumentation at the building site or a nearby site.Instruments located more than 1 km from the building site, or on sites withsignificantly different subsoil conditions should not be considered particularlyrepresentative. Seismologists have the capability to generate estimates of groundmotion using fault rupture simulation and wave propagation modeling techniquesthat may be useful for these purposes as well. However, the engineer should beadvised that great uncertainty is associated with such techniques and groundmotion representations generated in this manner are only estimates.

5.2.1.4 Vertical Plumbness Check

A rigorous vertical plumbness check is not necessary unless signs of a permanent lateral drift(e.g., elevators are not functioning, door jambs are distorted, or the building is visibly tilted) are



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observed at one or more floors. In such cases, a vertical plumbness check should be conducted bya licensed Surveyor to determine the extent that the post-earthquake out-of-plumbness exceedsAISC Frame Tolerances as defined in the AISC Code of Standard Practice Section 7.11. Ifsignificant permanent lateral drift is determined to exist, the structural engineer should determinewhether or not this drift, when superimposed with the postulated drift from a future earthquake,presents unacceptable P-∆ stress effects.

Commentary: When the plumbness check is deemed advisable, preliminarychecks can conveniently be made by the engineer from the interior of thebuilding, through the elevator shaft with the use of a plumb bob as elevatorappurtenances, such as sill plates on doors are typically constructed in closevertical alignment. When a more accurate evaluation of plumbness is required,surveying measurements should be made at each exterior principal corner of thebuilding..

5.2.2 Connection Exposure

Pre-inspection activities to expose and prepare a connection for inspection should include thelocal removal of suspended ceiling panels or (as applicable) local demolition of permanent ceilingfinish to access the connection; and cleaning of the column panel zone, the column flange,continuity plates, beam web and flanges. The extent of the removal of fireproofing should besufficient to allow adequate inspection of the surfaces to be inspected. Figure 5-1 suggests apattern that will allow both visual and NDT inspection of the top and bottom beam flange tocolumn joints, the beam web and shear connection, column panel zone and continuity plates, andcolumn flanges in the areas of highest expected demands. The maximum extent of the removal offireproofing need not be greater than a distance equal to the beam depth "d" into the beam span toexpose evidence of any yielding.

6”

6”

12”

Figure 5-1 Recommended Zone for Removal of Fireproofing


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Commentary: Cleaning of weld areas and removal of mill scale and weld spattershould be done with care, preferably using a power wire brush, to ensure a cleansurface that does not affect the accuracy of ultrasonic testing. The resultingsurface finish should be clean, free of mill scale, rust and foreign matter. The useof a chisel should be avoided to preclude scratching the steel surfaces whichcould be mistaken for yield lines. Sprayed-on fireproofing on WSMFs erectedprior to about 1980 is likely to contain asbestos and should be handled accordingto applicable standards for the removal of hazardous materials. To precludephysical exposure to hazardous materials and working conditions, the structuralengineer should require by contractual agreement with the building owner, priorto the start of the inspection program, that the building owner deliver to thestructural engineer for his/her review and files a laboratory certificate thatconfirms the absence of asbestos in structural steel fireproofing, local pipeinsulation, ceiling tiles, and drywall joint compound.

The pattern of fireproofing removal indicated in Figure 5-1 is adequate toallow visual and UT inspection of the top and bottom girder flange to columnjoints, the beam web and shear connection and the column panel zone. Asdiscussed in the commentary to Section 5.1.1, some engineers prefer to initiallyinspect only the bottom beam flange to column joint. In such cases, the initialremoval of fireproofing can be more limited than indicated in the figure. If afterinitial inspection, damage at a connection is suspected, then full removal, asindicated in the figure, should be performed to allow inspection of all areas of theconnection.

5.3 Inspection Program

5.3.1 Visual Inspection (VI)

Visual Inspection is the primary means of determining the condition of the structure. It shouldbe performed by, or under the direction of, a structural engineer, and in as many locations as ispractical. As a minimum, it should be performed in those locations selected in accordance withone of the methods of Chapter 4. It may be performed and documented by other competentpersons, but should be performed with a structural engineer's written instructions and guidance. When VI is performed by a testing agency, the agency and personnel performing the work shouldconform to the Interim Guidelines of Chapter 10. As a minimum, the structural engineer incharge should visit the site as needed during the performance of visual inspection to confirm thathis/her instructions are understood and followed, and to provide a spot check of the adequacy ofsurface preparation of the connection for VI and NDT, that the recorded locations of damage arecorrect, and that damage is accurately reported.

The presence or absence of damage should be recorded in a consistent and objective manneron a uniform data sheet that will allow later interpretation of the conditions and assessment of itsseverity and the types of repair which may be warranted. Severe damage should be documented



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with photographs. Data sheets should include a sketch of the connection and locations of anysignificant non-conforming or damage conditions noted. Damage should be classified inaccordance with the system indicated in Section 3.1.

Commentary: The presence or absence of the following conditions should berecorded for each inspected connection:

a) Deviations from Construction Documents or Specifications.

b) Continuity plates.

c) Doubler plates (on one or both sides of the web).

d) Supplemental web welds (from beam-web-to-shear-tab).

e) Flange weld backing bar and runoff tabs.

f) Flange weld end dams.

g) Poor Fit-up of backing bar.

h) Evidence of weld spatter (must be removed prior to performing Ultrasonic Testing).

i) Smooth (or rough) beam web cope for weld access holes.

j) Evidence of poor quality welding workmanship per AWS D1.1 Section 6.5.1 and 8.15.1.

k) Undercut, underfill or excessive concavity/convexity of welds.

l) Undersized fillet welds.

The presence or absence of damage should be recorded. For purposes ofvisual inspection, backing bars need not be removed. If damage is discovered, itshould be recorded by type, per the classification system of Chapter 3.

When full inspection of a connection is conducted, both sides of the beam,column, and panel zone should be inspected. If one side of the connection isobstructed (e.g., by exterior walls), such obstructions need not be removed if theaccessible side of the connection appears undamaged. Beam top flangeconnection welds may be inspected without local removal of the floor diaphragmfinish if there is no apparent significant damage at the bottom beam flange,adjacent column flange, column web, or shear connection. If severe beam bottomflange damage is observed, removal of diaphragm materials to allow directobservation of the beam top flange is recommended. More information on VImay be found in AWS B1.11.

5.3.1.1 Top Flange

The exposed root of this "T" joint should be inspected to note any possible separation of theedge of the backup bar from the face of the column flange. The exposed surface of the beamflange and column flange should be observed to note any cracks which may have occurred. Beam


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flange base metal at the intersection of the weld access hole in the web to the beam flange shouldbe inspected to note any visible cracks.

5.3.1.2 Bottom Flange

The possible separation of the backup bar to column flange should be inspected as above. Theface of the weld should be inspected to note any possible cracks in the weld, the toe of the weldor in the adjacent base metal. The base metal of the column flange above, below and each side ofthe intersecting beam flange should be inspected to note any possible visible cracks as should thebeam flange in the vicinity of the web access hole, as above.

5.3.1.3 Column and Continuity Plates

Column base metal and the continuity plates and their welds to the inside face of the columnflanges should be visually inspected. The column web above and below the continuity plateshould be inspected to note any visible cracks.

5.3.1.4 Beam Web Shear Connection

The shear connection plate, beam web, and corresponding bolts should be inspected to noteany possible rotation. The base metal around the bolt head and nut including washers if used, mayshow signs of bright metal if rotation has occurred. Observation should include examination forbolts which may have loosened as well as any welds used in combination with the boltedconnection. The exposed surface of the shear plate to column flange weld should be visuallyinspected with primary attention being paid to the termination of this weld near the beam's bottomflange.

5.3.2 Nondestructive Testing (NDT)

NDT should be used to supplement the visual inspection of connections selected inaccordance with the Interim Guidelines of Chapter 4. The testing agency and NDT personnelperforming this work should conform to the qualifications indicated in Chapter 11 of these InterimGuidelines. The following NDT techniques should be used at the top and bottom of eachconnection, where accessible, to supplement visual inspection:

a) Magnetic particle testing (MT) of the beam flange - to column flange weld surfaces. Allsurfaces which were visually inspected should be tested using the magnetic particletechnique.

Commentary: The color of powder should be selected to achieve maximumcontrast to the base and weld metal under examination. The test may be furtherenhanced by applying a white coating made specifically for MT or by applyingpenetrant developer prior to the MT examination. This background coatingshould be allowed to thoroughly dry before performing the MT.



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b) Ultrasonic testing (UT) of all faces at the beam flange welds and adjacent column flanges(extending at least 3 inches above and below the location of the CJP weld, along the faceof the column, but not less than 1-1/2 times the column flange thickness).

Commentary: The purpose of UT is to 1) locate and describe the extent ofinternal defects not visible on the surface and 2) to determine the extent of cracksobserved visually and by MT.

Requirements and acceptance criteria for NDT should be as given in AWS D1.1 Sections 6and 8. Acceptance or rejection of planar weld discontinuity (cracks, slag inclusion, or lack offusion), including root indications, should, as a minimum, be consistent with AWS DiscontinuitiesSeverity Class designations of cracks and defects per Table 8.2 of AWS D1.1 for StaticStructures. Beam flange welds should be tested as "tension welds" per AWS D1.1 Table 8.15.3,Note 3. Backing bars need not be removed prior to performing UT.

Commentary: The value of UT for locating small discontinuities at the root ofbeam flange to column flange welds when the backing is left in place is notuniversally accepted. The reliability of this technique is particularly questionableat the center of the joint, where the beam web obscures the signal. There havebeen a number of reported instances of UT detected indications which were notfound upon removal of the backing, and similarly, there have been reportedinstances of defects which were missed by UT examination but were evident uponremoval of the backing. The smaller the defect, the less likely it is that UT alonewill reliably detect its presence.

Despite the potential inaccuracies of this technique, it is the only methodcurrently available, short of removal of the backing, to find subsurface damage inthe welds. It is also the most reliable method for finding lamellar problems in thecolumn flange (type C4 and C5 damage) opposite the girder flange. Removal ofweld backing at these connections results in a significant cost increase that isprobably not warranted unless UT indicates widespread, significant defectsand/or damage in the building.

The proper scanning techniques, beam angle(s) and transducer sizes should be used asspecified in the written UT procedure contained in the Written Practice, prepared in accordancewith Section 5.3.3 of these Interim Guidelines. The acceptance standard should be that specifiedin the original contract documents, but in no case should it be less than the acceptance criteria ofAWS D1.1, Chapter 8, for Statically Loaded Structures.

The base metal should be scanned with UT for cracks. Cracks which have propagated to thesurface of the weld or beam and column base metal will probably have been detected by visualinspection and magnetic particle tests performed earlier. The purpose of ultrasonic testing of thebase metal is to:

1. Locate and describe the extent of internal indications not apparent on the surface and,


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2. Determine the extent of cracks found visually and by magnetic particle test.

Commentary: Liquid dye penetrant testing (PT) may be used where MT isprecluded due to geometrical conditions or restricted access. Note that morestringent requirements for surface preparation are required for PT than MT, perAWS D1.1.

If practical, NDT should be performed across the full width of the bottombeam flange joint. However, if there are no discontinuity signals from UT ofaccessible faces on one side of the bottom flange weld, obstructions on the otherside of the connection need not be removed for testing of the bottom flange weld.

Slabs, flooring and roofing need not be removed to permit NDT of the topflange joint unless there is significant visible damage at the bottom beam flange,adjacent column flange, column web, or shear connection. Unless such damageis present, NDT of the top flange should be performed as permitted, without localremoval of the diaphragms or perimeter wall obstructions.

It should be noted that UT is not 100% effective in locating discontinuitiesand defects in CJP beam flange to column flange welds. The ability of UT toreliably detect such defects is very dependent on the skill of the operator and thecare taken in the inspection. Even under perfect conditions, it is difficult toobtain reliable readings of conditions at the center of the beam flange to columnflange connection as return signals are obscured by the presence of the beamweb. If backing is left in place on the welds, UT becomes even less reliable. There have been a number of reported instances in which UT indicated apparentdefects, that were found not to exist upon removal of the backing. Similarly, UThas failed in some cases to locate defects that were later discovered upon removalof the backing. Additional information on UT may be found in AWS B1.10.

5.3.3 Inspector Qualification

Testing shall be performed and supervised by qualified and properly certified technicians. It isrecommended that the structural engineer (or his/her agent) observe the inspection proceduresdirectly until such a time that confidence is developed that the inspections are being made inaccordance with the given instructions. It is the responsibility of the structural engineer in chargeto confirm that only certified Level II technicians, certified in accordance with AWS D1.1-1994Section 6.1.3.1 and 6.7.8 and SNT-TC-1A, are allowed to execute VI and NDT, under thesupervision of a Level III technician with current certification by examination. All Level IIcertifications should be current, having been issued within 3 years prior to the start of theinspection program. The structural engineer should require that the inspector provide a WrittenPractice for his/her review and approval, and the building owner's file, in accordance with therequirements of SNT-TC-1A. The Written Practice should as a minimum provide 1) allcertification records for all technicians executing VI and NDT on the project and 2) the detailed



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UT, MT and PT procedures used to perform NDT of WSMF connections, as well as other typesof connections, in accordance with the instructions of the structural engineer.

Commentary: Special care is recommended in the selection of inspectors becauseVI and NDT methods are highly dependent on the skill and integrity of theoperator for proper interpretation of the results.

5.3.4 Post-Earthquake Field Inspection Report

A Field Inspection Report for both VI and NDT should be completed for each connectioninspected and/or tested, regardless of whether damage or rejectable defects are detected. Thereare two (2) CJP welds for each WSMF beam-to-column connection. A standard form should beused to ensure complete documentation. Sample Field Inspection Report forms are provided inFigure 5-2 through 5-5. The technician should record the depth, length and location of observedindications, should characterize the discontinuities as planar (cracks or lack of fusion) orvolumetric (porosity, slag, etc.), should classify the weld as acceptable or rejectable according topredetermined criteria, and should note any uncertainties. In addition, the form should record thedate of testing, the person responsible, connection location and orientation, and descriptions ofitems not tested due to limited access. The Field Inspection Report form should, as a minimum,objectively identify for each CJP weld and shear tab tested the following information, asapplicable:

a) Damage/Defect type classification/description (per Section 3-1, and summarized forconvenience at the rear of this Chapter as Table 5-1).

b) AWS Discontinuity Severity Class of crack/defect per Table 8.2 of AWS D1.1-94.

c) Depth of crack/defect.

d) Length of crack/defect/damaged material.

e) Location of crack/defect/damaged material.

f) Identification of NDT procedure used.

g) Possible inclusion of photographs.

Commentary: To ensure a correct understanding and identification of reportedconnection damage, the format of the Field Inspection Report form shouldinclude an easy to understand graphic description of what face of the connectionis being inspected (e.g., north face, south face, east or west) and at what framingelevation the connection is located (e.g., inspector is standing on the 4th floorand looking at a connection located at the 5th floor framing). In addition, anyidentified significant damage should be recorded on a generic sketch of theWSMF standard joint detail to facilitate consistent reporting and correctinterpretation and assessment by the structural engineer.


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5.3.5 Written ReportFollowing completion of the detailed damage assessments, the structural engineer should

prepare a written report, in accordance with Section 4.3.9.



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Table 5-1 - Connection Damage ClassificationType Location DescriptionG1 Girder Buckled FlangeG2 Girder Yielded FlangeG3 Girder Top or Bottom Flange fracture in HAZG4 Girder Top or Bottom Flange fracture outside HAZG5 Girder Top and Bottom Flange fractureG6 Girder Yielding or Buckling of WebG7 Girder Fracture of WebG8 Girder Lateral-torsional BucklingC1 Column Incipient flange crack (detectable by UT)C2 Column Flange tear-out or divotC3 Column Full or partial flange crack outside HAZC4 Column Full or partial flange crack in HAZC5 Column Lamellar flange tearingC6 Column Buckled FlangeC7 Column Fractured column spliceW1a CJP weld Minor root indication, thickness < 3/16” or tf/4;

width < bf/4W1b CJP weld Root indication, thickness > 3/16” or tf/4;

width > bf/4W2 CJP weld Crack through weld metal thicknessW3 CJP weld Fracture at girder interfaceW4 CJP weld Fracture at column interfaceW5 CJP weld UT detectable indication— non-rejectableS1a Shear tab Partial crack at weld to column (beam flanges sound)S1b Shear tab Partial crack at weld to column (beam flange cracked)S2a Shear tab Crack in Supplemental Weld (beam flanges sound)S2b Shear tab Crack in Supplemental Weld (beam flange cracked)S3 Shear tab Fracture through tab at bolt holesS4 Shear tab Yielding or buckling of tabS5 Shear tab Damaged, or missing boltsS6 Shear tab Full length fracture of weld to columnP1 Panel Zone Fracture, buckle, or yield of continuity plateP2 Panel Zone Fracture of continuity plate weldsP3 Panel Zone Yielding or ductile deformation of webP4 Panel Zone Fracture of doubler plate weldsP5 Panel Zone Partial depth fracture in doubler plateP6 Panel Zone Partial depth fracture in webP7 Panel Zone Full (or near full) depth fracture in web or doubler

plateP8 Panel Zone Web bucklingP9 Panel Zone Fully severed column


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Figure 5-2 Inspection Form - Major Axis Column Connection



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Figure 5-3 Inspection Form - Large Discontinuities - Major Axis


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Figure 5-4 Inspection Form - Major Axis Column Connection



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Figure 5-5 Inspection Form - Large Discontinuities - Minor Axis

Interim Guidelines:Evaluation, Repair, Modification andDesign of Steel Moment Frames Chapter 6 - Post-Earthquake Repair and Modification

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6. POST-EARTHQUAKE REPAIR AND MODIFICATION

As used in these Interim Guidelines, repair means restoration of the strength, stiffness anddeformation capacity of structural elements that have been damaged or have constructiondefects. Modification means actions taken to enhance the strength, stiffness or deformationcapacity of either damaged or undamaged elements, or of the structure as a whole.

Based on the observed behavior of actual buildings in the Northridge Earthquake, as well asrecent test data, WSMF structures constructed with the typical pre-Northridge detailing andconstruction practice prevalent prior to the Northridge Earthquake do not have the samedeformation capacity they were presumed to possess at the time of their design. The seismic riskassociated with these structures is higher than typically judged as acceptable for buildings of newconstruction. When these buildings are damaged or have excessive construction defects, the riskis higher.

Based on limited testing, it appears that the repair recommendations contained in this Chaptercan be effective in restoring a building’s pre-earthquake condition. This does not imply,however, that the repaired building will be an acceptable seismic risk. As a minimum, it shouldbe assumed that buildings that are repaired, but not modified, can sustain similar and possiblymore severe damage in future earthquakes than they did in the present event. If this isunacceptable, either to the owner or the building official, then the building should be modified toprovide improved future performance. Modification can consist of local reinforcement ofindividual moment connections as well as alteration of the basic lateral-force-resistingcharacteristics of the structure through addition of braced frames, shear walls, base isolation,energy dissipation devices, etc.

6.1 Scope

This section provides interim guidelines for structural repair of earthquake damage andmodification of structures to improve future earthquake performance. Repair constitutes anymeasure(s) taken to restore earthquake damaged joints, connections, elements of the building, orthe building as whole, to their original strength, stiffness and deformation capacity. It does notinclude routine correction of non-conforming conditions during original fabrication. InterimGuidelines for acceptable methods of repair are provided in Sections 6.2 through 6.5 below.These Interim Guidelines are not intended to be used for the routine repairs of non-conformancecommonly encountered in fabrication and erection work. Industry standard practices areacceptable for such repairs.

Work that increases structural stiffness or strength of an element or the structure as a wholeby more than 5% is classified as modification. Guidelines for some methods of modification arecontained in Section 6.6.


Chapter 6 - Post-Earthquake Repair & Modification Design of Steel Moment Frames

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6.2 Shoring

6.2.1 Investigation

The structural engineer responsible for designing any damage repair should investigate theentire building for imminent collapse or life safety hazard conditions, regardless of jointconsiderations. Such conditions should be shored prior to commencement of any repairs.

Commentary: In projects relating to construction of new buildings, it is commonpractice to delegate all responsibility for temporary shoring and bracing of thestructure to the contractor. Such practice may not be appropriate for severelydamaged buildings. The structural engineer should work closely with thecontractor to define shoring and bracing requirements. Some structuralengineers may wish to perform the design of temporary bracing systems. If thecontractor performs such design, the structural engineer should review thedesigns for adequacy and potential effects on the structure prior toimplementation.

6.2.2 Special Requirements.

Conditions which may become collapse or life safety hazards during the repair operationsshould be considered in the development of repair details and specifications, whether theyinvolve the connection area directly or indirectly. These conditions should be brought to theattention of the contractor by the structural engineer, and adequate means of shoring theseconditions should be provided. Consideration should be given to sequencing of repairprocedures for proper design of any required shoring. For column repair details that requireremoval of 20% or more of the damaged cross section, consideration should be given to the needfor shoring to prevent overstress of elements due to redistribution of loads.

Commentary: In general, contractors will not have adequate resources to definewhen such shoring is necessary. Therefore, the Contract Documents shouldclearly indicate when and where shoring is required. Design of this shoring maybe provided by the structural engineer, or the contractor may submit a shoringdesign to the structural engineer for review.

6.3 Repair Details

The scope of repair work should be shown on drawings and specifications prepared by astructural engineer. The drawings should clearly indicate the areas requiring repair, as well as allrepair procedures, details, and specifications necessary to properly implement the proposedrepair. Sample repair details for various types of damage are included in these InterimGuidelines, for reference, only.

Commentary: Examples of repair details are provided for some classes ofdamage, based on previous repairs performed in the field for specific projects.Limited testing indicates these repair methods can be effective. Details are not


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complete in all respects and should not be used verbatim, as constructiondocuments. Many repairs will require the application of more than oneoperation, as represented by a given detail. The sample details indicated may notbe directly applicable to specific repair conditions. The structural engineer iscautioned to thoroughly review the conditions at each damaged element,connection or joint, and to determine the applicability and suitability of thesedetails based on sound structural engineering judgment, prior to employing themon projects.

6.3.1 Approach

Based on the nature and extent of damage several alternative approaches to repair should beconsidered. Repair approaches may include, but should not be limited to:

a) replacement of portions of base metal (i.e. column and beam section),

b) replacement of connection elements,

c) replacement of connection weld, or

d) repairs to portions of any of the aforementioned components.

Any or all of these techniques may be appropriate. The approach(s) used should consideradjacent structural components which may be affected by the repair or the effects of the repair.

Where base material is to be removed and replaced with plates, clear direction should begiven to orient the plates with the direction of rolling of the plate parallel to the direction ofapplication of major axial loads to be resisted by the plate.

6.3.2 Weld Fractures - Type W Damage

All fractures and rejectable defects found in weld material, either between girder and columnor between connection element and structural member, should have sufficient material removedto completely eliminate any discontinuity or defect. NDT should be used to determine the extentof fracture or defect and sufficient material should be removed to encompass the damaged area.It is suggested that material removal extend 2 inches beyond the apparent end of the fracture ordefect. Simple fillet welds may be repaired by backgouging to eliminate unsound weld materialand replacing the damaged weld with sound material. Complete joint penetration (CJP) weldsfractured through the full thickness should be replaced with sound material deposited in strictaccordance with the Welding Procedure Specification (WPS) and project specifications. The useof weld dams on new welds is prohibited. Weld backing (backup bars), existing dams, and weldtabs should be removed from all welds that are being repaired. After backing is removed, theroot should be backgouged to sound material, rewelded and a reinforcing fillet added.

The structural engineer is cautioned to observe the provisions of AISC regarding intermixingof weld metals deposited by different weld processes (see AISC LRFD Manual of Steel



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Construction, second edition, page 6-77, and AISC ASD Steel Construction Manual, ninthedition, page 5-69). As an example, E7018 stick electrodes should not be used to weld over self-shielded flux cored arc welding deposits. Removed weld material from fractures not penetratingthe full weld thickness should be replaced in the same manner as full thickness fractures. Forother types of W damage, existing backing, end dams, and weld tabs should also be removed in alike manner to CJP weld replacement. Table 6-1 provides an index to suggested repair detailsfor type W damage.

Table 6-1 - Reference Details for Type W Damage

Damage Class FigureW1a, W1b Figure 6-1, Figure 6-

2W2 Figure 6-3W3 Figure 6-3W4 Figure 6-3W5 Figure 6-3

Commentary: FCAW-ss utilizes approximately 1-2% aluminum in the electrode toprotect the weld from mixing with atmospheric nitrogen and oxygen. By itself,aluminum can reduce the toughness and ductility of weld metal. The design ofFCAW-ss electrodes requires the balance of other alloys in the deposit tocompensate for the effects of aluminum. Other welding processes rely on fluxesand/or gasses to protect the weld metal from the atmosphere, relieving them ofany requirement to contain aluminum or other elements that offset the effects ofaluminum. If the original weld that is being repaired consists of FCAW-ss andsubsequent repair welds are made with SMAW (stick) using E7018, for example,the SMA arc will penetrate into the FCAW-ss deposit, resulting in the addition ofsome aluminum into the SMAW deposit. The notch toughness and/or ductility ofthe resultant weld metal may be substantially reduced as compared to pure E7018weld metal, based on the depth of penetration into the FCAW-ss material.

Various types of FCAW-ss electrodes may be mixed one with the other withoutpotentially harmful effect. Further, FCAW-ss may be used to weld over othertypes of weld deposits without potentially harmful interaction. The structuralengineer could specify all repairs on FCAW-ss deposits be made with FCAW-ss.Alternately, intermixing of FCAW-ss and other processes could be permittedprovided the subsequent composition is demonstrated to meet materialspecification requirements.


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20o Min.Arc - Gouge

1/4” radius min.

Existing column flange

Removed backing

Existing beam flange

Reweld & reinforcew/ fillet

Notes:

1 Remove existing backing.2 Taper the depth of grinding or air arc gouging at each end to the face of flange with a minimum 2:1

(horizontal/vertical) taper. Provide a minimum root radius of 1/4.”3 Grind all surfaces on which weld metal will be deposited. Surfaces should be smooth, uniform and free

from fins, tears, fractures and other discontinuities which would adversely affect weld strength.4 A fillet weld should be applied to reinforce the joint. The size of the reinforcing fillet should be equal to

1/4 of the beam flange thickness, but not less than 1/4.” It need nor be more than 3/8.”5 On joints to be repaired, remove all remaining weld tabs and excess weld metal beyond the length of the

joint and grind smooth. Imperfection less than 1/16" should be removed by grinding. Repair as necessary.

Figure 6-1 - Gouge & Re-weld of Root Defect or Damage - W1


Removed backing


Backgouge , repair andreinforce per Figure 6-1.

Air-arc gougeReweld

Notes:1. Remove the entire fracture plus 2” of sound metal beyond each end.2. For additional notes, refer to Figure 6-1

Figure 6-2 - Gouge & Re-weld of Fractured Weld - W1



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Remove backing aftercompleting top welding,Backgouge , repair andreinforce, per Figure 6-1.

Air-arc gougeReweld

For notes see Figure 6-1 and 6-2.

Figure 6-3 - Backgouge and Reweld repair

6.3.3 Column fractures - Type C1 - C5 and P1 - P6

Any column fracture observable with the naked eye or found by NDT and classified asrejectable in accordance with the AWS D1.1 criteria for Static Structures should be repaired.Repairs should include removing the fracture such that no sign of rejectable discontinuity ordefect within a six (6) inch radius around the fracture remains. Removal should includeeliminating any zones of fracture propagation, with a minimum of heat used in the removalprocess. Following removal of material, MT and PT should be used to confirm that all fracturedmaterial has been removed. Repairs of removed material may consist of replacement of portionsof column section, build-up with weld material where small portions of column were removed,or local replacement of removed base metal with weld material. Procedures of weld fracturerepair should be applied to limit the heat affected area and to provide adequate ductility to therepaired joint. Tables 6-2 and 6-3 indicate representative details for these repairs. In manycases, it may be necessary to remove a portion of the girder framing to a column, in order toattain necessary access to perform repair work, per Figure 6-4. Refer to Section 6.3.5 for repairof girders.

Shore Beam

Remove Shear TabReplace upon completion

Remove portion of existing beam. Provide minimum 2” radius.

New web platethickness = tw +1/8”

tw


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Figure 6-4 - Temporary Removal of Beam Section for Access

When the size of divot (type C2) or transverse column fractures (types C1, C3, C4) dictate atotal cut-out of a portion of a column flange or web (types P6, P7), the replacement material shouldbe ultrasonically tested in accordance with ASTM A578-92, “Straight-Beam UltrasonicExamination of Plane and Clad Steel Plates for Special Applications,” in conjunction with AWSK6.3 “Shearwave Calibration.” Acceptance criteria should be that of Level III. The replacementmaterial should be aligned with the rolling direction matching that of the column.

Table 6-2 - Reference Details for Type C and P Damage

Damage Class FigureBeam Access Figure 6-4C1 Figure 6-4, 6-5C2 Figure 6-4, 6-6C3 Figure 6-4, 6-5C4 Figure 6-4, 6-5C5 Figure 6-4, 6-6P1 remove, prepare, replaceP2 arc-gouge and reweldP4 arc-gouge and reweldP5 Figure 6-7P6 Figure 6-7P7 Figure 6-7P8 Figure 6-8

Portion of E beamflange removedWeld access hole

in column web 45o

per AWS D1.1section 3.2.5,and Figure 3.2

Backgouge and reweld

10o

1 Investigate extent of fracture by UT to confirm that fractures are contained with the 45 degree angle zoneof a standard pre-qualified CJP groove weld as defined by AWS D1.1, Figure 2.4, Joint Designation B-U4a-G

2 Provide 10o bevel on lower flange plate, to channel slag out of joint.3 Grind all surfaces upon which weld metal will be deposited to smooth, uniform surface.



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Figure 6-5 - Backgouge and reweld of column flange

New flangesplice plate

Weld accesshole and backing

10o

6” m

inim

um

Note: Provide new flange plate material of the same strength, and width as the existing column flange. Alignrolling direction of plate with that of column flange. New plate should be of the same thickness as theexisting flange with a tolerance of -0”/+1/4.” The welding should be sequenced to connect the columnflange to new flange plate welds prior to welding the column web to new flange plate. Bevel the loweredge of the column flange, and upper edge of the splice plate down 10o, to channel slag out of joint.

Figure 6-6 - Replacement of Column Flange Repair

Doubler PlateColumn web

Typical

Web with Doubler Plate Web without Doubler Plate

Notes:1. Prepare fractured section of doubler by air-arc gouging, grind and reweld, using web as backing2. Prepare fractured section of web by air-arc gouging, grind and reweld, using doubler as backing or

backgouge and reweld from reverse side, if no doubler present.


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Figure 6-7 - Reweld Repair of Web plate and Doubler plate

Flange removal and replacmentper Figure6-6, if required

Weld access holes as requiredfor weld terminations

Notes:1. Sequence removal of portions of column and provide shoring as required to safely support existing column

loads.2. Thickness of new web plate to match existing column web (tolerance +1/8”, -0”).

Figure 6-8 - Alternate Column Web Repair - Columns without DoublerPlates

Commentary: Special attention should be given to conditions where more than20% of the column cross section will be removed at one time, as specialtemporary shoring may be warranted. In addition, care should be taken whenapplying heat to a flange or web containing a fracture, as fractures have beenobserved to propagate with the application of heat. This can be prevented bydrilling a small diameter hole at the end of the fracture, to prevent it fromrunning.

6.3.4 Column splice fractures - Type C7

Any fractures detected in column splices should be repaired by removing the fracturedmaterial and replacing it with sound weld material. For partial joint penetration groove welds,remove up to one half of the material thickness from one side and replace with sound material.Where complete joint penetration groove welds are required, it may be preferable to provide adouble bevel weld, repairing one half of the material thickness completely prior to preparing andrepairing the other half. Alternatively, if calculations indicate that column loads may safely beresisted with the entire section of column flange removed, or if suitable shoring is provided, itmay be preferable to use a single bevel weld.

Commentary: Special attention should be given to these conditions, as theremoval of material may require special temporary shoring. Also, since partialpenetration groove welds can serve as fracture initiators in tension applications,



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consideration should be given to replacing such damaged splice areas withcomplete joint penetration welds.

6.3.5 Girder Flange Fractures - Type G3-G5

Repair of fractures in girder flanges may be performed by several methods. One method is toremove the fracture by air arc gouging such that no sign of discontinuity or defect within a six(6) inch radius around the fracture remains, preparing the surface by grinding and welding newmaterial back. Alternatively, damaged portions of the girder flange may be removed andreplaced with new plate as shown in Figure 6-9 or Figure 6-10.

Typ.

New beam flange plate

New web stiffeners, near side and far side

Weld accesshole

Notes:1. New plate thickness to match beam flange thickness + height of removed web fillet.2. Weld sequence - a) weld of new flange plate to column; b)weld of flange plate stiffeners to web and flange

plate; c) weld of new flange plate to beam flange. d)weld of stiffener plate to beam flange and web

Figure 6-9 - Beam Flange Plate Replacement


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New beam flange plate

Figure 6-10 - Alternative Beam Flange Plate Replacement

Commentary: Due to accessibility difficulties or excessive weld build-uprequirements, it may become necessary to remove a portion of the girder flange toproperly complete the joint repair. A minimum of six inches of girder flange maybe removed to facilitate the joint repair, with the optimum length being equal tothe flange width. After removal of the portion of flange, the face of column andcut edge of girder flange may then be prepared to receive a splice plate matchingthe flange in grade and width. Thickness should be adjusted as required to make-up the depth of the girder web and fillet removed as part of the preparationprocess.

It is recommended that a double bevel joint be utilized in replacing theremoved plate to eliminate the need for backup bars, consequently alsoeliminating the removal of these backup bars. A suggested joint detail is a B-U3/TC-U5, per AWS D1.1, with 1/3 tflange-2/3 tflange bevels on the plate. The webof the girder should be prepared at the column and butt weld areas to allowwelding access. Weld tabs may be used at the column and butt weld. The weldbetween the splice plate and the column flange should be completed first. If adouble bevel weld is selected, the welder may choose to weld the first few passesfrom one face, then backgouge and weld from the second side. This may help tokeep the interpass temperature below the maximum without down time oftenencountered in waiting for the weld to cool.

6.3.6 Buckled Girder Flanges - Type G1

Where the top or bottom flange of a girder has buckled, and the rotation between the flangeand web is less than or equal to the mill rolling tolerance given in the AISC Manual of SteelConstruction (AISC-1994 or AISC-1989) the flange need not be repaired. Where the angle isgreater than mill rolling tolerance, repair should be performed and may consist of adding full



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height stiffener plates on the web over each portion of buckled flange, contacting the flange atthe center of the buckle, (Figure 6-11) or using heat straightening procedures. Another availableapproach is to remove the buckled portion of flange and replace it with plate, similar to Figures6-9 and 6-10.

New stiffener plateseach side,tplate = tweb

Note: Provide stiffeners at beginning of buckle and at center of buckle

Figure 6-11 - Addition of Stiffeners at Buckled Girder Flange

Commentary: Should flange buckling occur on only one side of the web, and thebuckle repair consists of adding stiffener plates, only the side that has buckledneed be stiffened. In case of partial flange replacement, special shoringrequirements should be considered by the design engineer.

6.3.7 Buckled column flanges - Type C6

Any column flange or portion of a flange that has buckled to the point where it exceeds therolling tolerances given in the AISC Manual of Steel Construction should be repaired. Flangerepair may consist either of flame straightening or of removing the entire buckled portion offlange and replacing it with material with yield properties similar to the actual yield properties ofthe damaged material similar to Figure 6-6. If workers with the appropriate skill to performflame straightening are available, this is the preferred method.

Commentary: For flange replacement, shoring is normally required. Thisshoring should be designed by the structural engineer, or may be designed by thecontractor provided the design is reviewed by the structural engineer.

Flame straightening can be an extremely effective method of repairingbuckled members. It is performed by applying heat to the member in a triangularpattern, in order to induce thermal strains that straighten the member out. Verylarge bends can be straightened by this technique. However, the practice of thistechnique is not routine and there are no standard specifications available forcontrolling the work. Consequently, the success of the technique is dependent onthe availability of workers who have the appropriate training and experience toperform the work. During the heat application process, the damaged member islocally heated to very high temperatures. Consequently, shoring may be requiredfor members being straightened in this manner.


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A number of references are available that provide more information on thisprocess and its applications, published by AISC and others (Avent - 1992), (Avent- 1995), (Shonafelt and Horn - 1984)

6.3.8 Gravity connections

Connections not part of the lateral load-resisting system may also be found to require repairdue to excessive rotation or demand caused by distress of the lateral load-resisting system in thezone of influence. These connections should be repaired to a capacity at least equivalent to thepre-damaged connection capacity. Shear connections that are part of the lateral load resistingsystem should be repaired in a similar manner, with special consideration given to the nature andsignificance of the overall structural damage. In buildings which are repaired, but not modified,future earthquakes may cause moment connection failures with resulting large buildingdeflections and high rotation demands at gravity connections. When repairing gravityconnections, consideration should be given to providing connections with the ability to rotatewith little or no reduction in vertical load carrying capacity, possibly by dissipating energy(through the use of slip critical bolts with horizontal short slotted holes).

Commentary: In many cases, shear connections which were not a part of thelateral-force-resisting system provided an unanticipated redundancy afterdamage occurred to the primary WSMF lateral system. While repair detailscould provide for rotation to minimize damage, such details should not eliminatethe beneficial effect of the extra strength and stiffness these shear connectionsprovide. This is especially important in framing systems with low moment frameredundancy.

The suggestion of providing gravity connections with slotted holes and slipcritical bolts may be a reasonable compromise. Such a connection would becapable of providing some additional, unintended, strength and stiffness for thebuilding but would also be able to withstand relatively large rotations withoutjeopardizing the gravity support the connection is actually intended to provide.

6.3.9 Reuse of Bolts

Bolts in a connection displaying bolt damage or plate slippage should not be re-used. Asindicated in the AISC Specification for Structural Joints using ASTM A325 or A490 Bolts(American Institute of Steel Construction - 1985), A490 bolts and galvanized A325 bolts shouldnot be retightened and re-used under any circumstances. Other A325 bolts may be reused ifdetermined to be in good condition. Touching up or retightening previously tightened boltswhich may have been loosened by the tightening of adjacent bolts need not be considered asreuse provided the snugging up continues from the initial position and does not require greaterrotation, including the tolerance, than that required by Table 5 of the AISC Specification. Boltsin connections displaying bolt or plate slippage should not be reused.

Commentary: Proper performance of high strength bolts used in slip criticalapplications requires proper tensioning of the bolt. Although a number of



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methods are available to ensure that bolts are correctly tensioned, the mostcommon methods relate to torquing of the nut on the bolt. When a bolt has beendamaged, the torquing characteristics will be altered. As a result, damaged boltsmay either be over-tightened or under-tightened, if reinstalled. The threads ofASTM A-490 bolts and galvanized ASTM A-325 bolts become slightly damagedwhen tightened, and consequently, should not be reused. To determine if anungalvanized ASTM A-325 bolt is suitable for re-use, a nut should be run up thethreads of the bolt. If this can be done smoothly, without binding, then the boltmay be re-used.

6.3.10 Welding Specification

Welded repairs involving thick plates and conditions of high restraint should be specifiedwith caution. These conditions can lead to large residual stresses and in some cases, initiation ofcracking before the structure is loaded. The potential for problems can be reduced by specifyingappropriate joint configurations, welding processes, control of preheat, heat input during weldingand cooldown, as well as selecting electrodes appropriate to the application. Engineers who donot have adequate knowledge to confidently specify these parameters should seek consultationfrom a person with the required expertise.

6.4 Preparation

6.4.1 Welding Procedure Specifications

A separate Welding Procedure Specification (WPS) should be established for every differentweld configuration, welding position, and material specification. Two categories of qualifiedwelding procedures are given in AWS D1.1-94. The WPS should be reviewed by the structuralengineer responsible for the repairs. The WPS is a set of focused instructions to the welders andinspectors stating how the welding is to be accomplished. Each type of weld should have its ownWPS solely for the purpose of that weld. The WPS should include instructions for jointpreparation based on material property and thickness, as well as welding parameters. Weldprocess, electrode type, diameter, stick-out, voltage, current, and interpass temperature should beclearly defined. In addition, joint preheat and postheat requirements should be specified asappropriate, including insulation guidelines if applicable. The WPS should also list appropriateinterim specification requirements that are mandated by the project specification.

Commentary: Preparation of the WPS is normally the responsibility of thefabricator/erector. Sample formats for WPS preparation and submission areincluded in AWS D1.1. Some contractors fill out the WPS by inserting referencesto the various AWS D1.1 tables rather than the actual data. This does not meetthe intent of the WPS which is to provide specific instructions to the welder andinspector on how the weld is to be performed. The actual values of theparameters to be used should be included in the WPS submittal.


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6.4.2 Welder Training

Training of welders should take place at the outset of the repair operations. Welders andinspectors should be familiar with the WPS, and should be capable of demonstrating familiaritywith each of its aspects. A copy of the WPS should be located on site, preferably at theconnection under repair, accessible to all parties involved in the repair.

6.4.3 Welder Qualifications

Welders must be qualified and capable of successfully making the repair welds required. Allwelders should be qualified to the AWS D1.1 requirements for the particular welding processand position in which the welding is to be performed. Successful qualification to theserequirements, however, does not automatically demonstrate a welder’s ability to make repairwelds for all the configurations that may be encountered. Specific additional training and/orexperience may be required for repair situations. Welders performing repairs should have aminimum of two years of verifiable field experience for the welding process that is employed, aswell as experience in arc-gouging and thermal cutting of material. Inexperienced welders shoulddemonstrate their ability to make proper repair welds. This may be done by welding on a mock-up assembly (see Section 6.4.4) that duplicates the types of conditions that would be encounteredon the actual project. Alternatively, the welder could demonstrate proficient performance on theactual project, providing this performance is continuously monitored, start to finish, during theconstruction of at least the first weld repair. This observation should be made by a qualifiedwelding inspector or Welding Engineer.

6.4.4 Joint Mock-ups

A joint mock-up should be considered as a training and qualification tool for each type ofrepair the welder is to perform that is more challenging than work in which he/she has previouslydemonstrated competence, or at the discretion of the structural engineer. This will allow thewelder to become familiar with atypical welds, and will give the inspector the opportunity toclearly observe the performance of each welder. An entire mock-up is recommended for eachsuch case, rather than only a single pass or portion of the weld as all welding positions and typesof weld would be experienced, thus showing the welder capable of successfully completing theweld in all required positions, and applying all heating requirements.

6.4.5 Repair Sequence

Repair sequence should be considered in the design of repairs, and any sequencingrequirements should be clearly indicated on the drawings and WPS. Structural instabilities orhigh residual stresses could arise from improper sequencing. The order of repair of flanges,shear plates, fractured columns, etc. should be indicated on the drawings to reduce possibleresidual stresses.



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6.4.6 Concurrent Work

The maximum number of connections permitted to be repaired concurrently should beindicated on the drawings or in the project specifications.

Commentary: Although a connection is damaged, it may still posses significantability to participate in the structure’s lateral load resisting system.Consideration should be given to limiting the total number of connections beingrepaired at any one time, as the overall lateral load resistance of the structuremay be temporarily reduced by some repair operations. If many connections areunder repair simultaneously, the overall lateral resistance of the remaining frameconnections may not be adequate to protect the structure’s stability. Althoughthis appears to fall under the category of means and methods, the typicalcontractor would have no way of determining the maximum number ofconnections that can be repaired at any one time without requiring supplementallateral bracing of the building during construction. Therefore, the structuralengineer should take a pro-active role in determining this.

6.4.7 Quality Control/Quality Assurance:

Quality control and quality assurance should follow the guidelines set forth in Section 6.6and Chapters 9, 10 and 11 of these Interim Guidelines.

6.5 Execution

6.5.1 Introduction

Recommended general requirements should include the following:

1. Strict enforcement of the welding requirements in AWS D1.1 as modified in 1994UBC Chapter 22, Division VIII or IX.

Commentary: Following the 1994 Northridge Earthquake, the AWS established apresidential task group to determine if deficiencies in the D1.1 code contributedto the unexpected damage, and to determine if modifications to the code should bemade. That task group noted some areas of practice, related to steel momentframes in seismic zones, that could be improved relative to D1.1. These includedthe following recommendations:

a) the root pass of the complete joint penetration welds of beam tocolumn flanges should not exceed 1/4 inch in size, for prequalifiedprocedures.

b) where notch tough weld metal is desired, such as at the criticalcomplete joint penetration welds of beam flanges to columns, themaximum interpass temperature should not exceed 550o.


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c) when a FCAW process is used, the welding procedure specificationshould conform to the electrode manufacturer’s recommendations.

d) the criteria for joints loaded in tension should apply to both top andbottom flange connections in moment frames.

Future editions of the AWS D1.1 code may adopt some or all of theserecommendations. In the interim period, the structural engineer should considerincluding these recommendations in the project welding specifications, tosupplement the standard AWS D1.1 requirements.

2. Implementation of the special inspection requirements in 1994 UBC section 1701{NEHRP-91 Section 1.6.2.6} and AWS D1.1. Visual inspection means that theinspector inspects the welding periodically for adherence to the approved WeldingProcedure Specification (WPS) and AWS D1.1 starting with preliminary tackwelding and fit-up and proceeding through the welding process. Reliance on the useof nondestructive testing (NDT) at the end of the welding process alone should beavoided. Use visual inspection in conjunction with NDT to improve the chances ofachieving a sound weld.

3. Require the fabricator to prepare and submit a WPS with at least the informationrequired by AWS D1.1 as discussed in Section 4.

4. Welding electrodes should be capable of depositing weld metal with a minimumnotch toughness as described in Chapter 8.

5. All welds for the frame girder-column joints should be started and ended on weld run-off tabs where practical. All weld tabs should be removed, the affected area groundsmooth and tested for defects using the magnetic particle method. Acceptancecriteria should be AWS D1.1, section 8.15.1. Imperfections less than 1/16” should beremoved by grinding. Deeper gouges, areas of lack of fusion, slag inclusions, etc.,should be removed by gouging or grinding and rewelding following the proceduresoutlined above.

6. Weld dams do not meet the intent of weld tabs, are not permitted by AWS D1.1, andshould not be permitted in the work. Dams are not necessary when proper bead sizelimitations are observed.

7. Steel backing (backing bars), if used, should be removed from new and/or repairedwelds at the girder bottom flange, the weld root back-gouged by air arcing and thearea tested for defects using the magnetic particle method, as described above. Theweld should be completed and reinforced with a fillet weld. Removal of the weldbacking at repairs of the top girder flange weld may be considered, at the discretion ofthe structural engineer.



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Prior to removing weld backing, the contractor should prepare and submit a written WPS forreview by the structural engineer. The WPS should conform to the requirements of AWS D1.1.In addition, a WPS should be prepared for each welding process to be used on the project andshould include minimum preheat, maximum interpass temperatures, and the as-gouged crosssection which must simulate a prequalified joint design of D1.1. If for any reason the WPS doesnot meet the prequalified limits of AWS it should be qualified by test, in accordance withSection 5.2 of AWS D1.1 In addition the contractor should propose the method(s) which will beused to remove the weld backing, back gouge to sound metal and when during this process hewill apply preheat.

Although project conditions may vary, the following general guidelines may be considered:

The steel backing may be removed by either grinding or by the use of air arc, or oxy-fuel gouging. The zone just beyond the theoretical 90 degree intersection of the beamto column flange should be removed by either air arc or oxy-fuel gouging followed bya thin grinding disk, or by a grinding disk alone. This shallow gouged depth of weldand base metal should then be tested by MT to determine if any linear indicationsremain. If the area is free of indications the area may then be re-welded. The preheatshould be maintained and monitored throughout the process. If no furthermodification is to be made or if the modification will not be affected by a reinforcingfillet weld, the reinforcing fillet may be welded while the connection remains at orabove the minimum preheat temperature and below the maximum interpasstemperature.

If weld tabs were used and are to be removed in conjunction with the removal of the weldbacking, the tabs should be removed after the weld backing has been removed and fillet added.If cover plates are to be added, the removal of the weld tabs may occur before or after the plate isadded depending on the width and configuration of the plate. This sequence should be submittedto the structural engineer for his/her approval prior to the beginning of the work.

The weld tabs may be removed by air arc or oxy-fuel gouging followed by grinding or bygrinding alone. The resulting contour should blend smoothly with the face of the column flangeand the edge of the beam flange and should have a radius of 1/4-3/8 inch.

The finished surface should be visually inspected for contour and any visually apparentindications. This should be followed by magnetic particle testing (MT). Linear indicationsfound in this location of the weld may be detrimental. They may be the result of the final residueof defects commonly found in the weld tab area. Linear indications should be removed bylightly grinding or using a cutting tool until the indication is removed. If after removal of thedefect the ground area can be tapered and is not beyond the theoretical 90 degree intersection ofthe beam flange edge and column flange, weld repair may not be necessary and should beavoided if possible.

If the defect removal has extended into the theoretical weld section, then weld repair may benecessary. The weld repair should be performed in accordance with the contractor’s WPS, withstrict adherence to the preheat requirements.


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The surface should receive a final visual inspection and MT after all repairs and surfaceconditioning has been completed.

End dams, if present, should be removed if UT indicates rejectable flaws in the area of theend dam. Prior to removal of end dams, the contractor should submit a removal / repair planwhich lists the method of dam removal, defect removal, welding procedure including, process,preheat, and joint configuration. The tab may be removed by grinding, air arc or oxy-fuel torch.

Any weld defects should be removed by grinding or cutting tools, or by air arc gougingfollowed by grinding. The individual performing defect removal should be furnished the UTresults which describe the location depth and extent of the defect(s).

When the individual removing the defects has completed this operation, and has visuallyconfirmed that no remnants remain, the surface should be tested by MT. Additional defectremoval and MT may occur until the MT tests reveal that the defects have been removed.

The contour of the surface at this point may be too irregular in profile to allow welding tobegin. The surface should be conditioned by grinding or using a cutting tool to develop a jointprofile which conforms to the WPS. Prior to welding MT should be performed to determine ifany additional defects have been exposed.

Based upon a satisfactory MT the joint may be prepared for welding. Weld tabs (andbacking if necessary) should be added. The welding may begin and proceed in accordance withthe WPS. The theoretical weld must be completed for its full height and length. Carefulattention should be paid to ensure that weld bead size does not exceed that permitted by theWPS.

If specified, the weld tabs and backing should be removed in accordance with the guidelinesection describing this technique. The final weld should be inspected by MT and UT.

Commentary: Removal of the weld backing from the top flange may be difficult,particularly along perimeter frames where access to the outer side is restricted.Since the potential stress riser produced by the unwelded portions of the weldbacking are not located on the extreme outer fiber of the frame girder, thebenefits of removal may be limited in repair situations. Nevertheless, there maybe benefits to providing a weld with a more favorable contour (i.e. that producedby the reinforcing fillet). Tests conducted to date have not been conclusive withregard to the benefit of top flange weld backing removal. At this time, there is nodirect evidence that removal of weld backing from continuity plates in the columnpanel zone is required.

The decision to remove end dams should be based upon the results of UT.Since numerous stop - starts have occurred in this section of the theoretical weld,rejectable edge indications may reduce the integrity of the weld, especially duringdynamic or seismic loading. If, however the area is found acceptable by UTremoval is not necessary.



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Excessive weaving of the weld bead, which can lead to unacceptable stressesat the toe of each weave, should not be allowed. However, some oscillation of theelectrode may be required to obtain good fusion.

6.5.2 Girder Repair

If at bottom flange repairs back gouging removes sufficient material such that a weld backingis required for the repair, after welding the backing should be removed from the girder.Alternatively, a double-beveled joint may be used The weld root should be inspected and testedfor imperfections, which if found, should be removed by back-gouging to sound material. Areinforcing fillet weld should be placed at “T” joints equal to one-quarter of the girder flangethickness. It need not exceed 3/8 inch (see Note J, Figure 2.4 of AWS D1.1.)

If the bottom flange weld requires repair, the following procedure may be considered:

1. The root pass should not exceed a 1/4 inch bead size.

2. The first half-length root pass should be made with one of the following techniques,at the option of the contractor:

a) The root pass may be initiated near the center of the joint. If this approach isused, the welder should extend the electrode through the weld access hole,approximately 1” beyond the opposite side of the girder web. This is to allowadequate access for clearing and inspection of the initiation point of the weldbefore the second half-length of the root pass is applied. It is not desirable toinitiate the arc in the exact center of the girder width since this will limit access tothe start of the weld during post-weld operations. After the arc is initiated, travelshould progress towards the end of the joint (outboard beam flange edge), and theweld should be terminated on a weld tab.

b) The weld may be initiated on the weld tab, with travel progressing toward thecenter of the girder flange width. When this approach is used, the welder shouldstop the weld approximately 1” before the beam web. It is not advisable to leavethe weld crater directly in the center of the beam flange width since this willhinder post-weld operations.

3. The half length root pass should be thoroughly slagged and cleaned.

4. The end of the half length root pass that is near the center of the beam flange shouldbe visually inspected to ensure fusion, soundness, freedom from slag inclusions andexcessive porosity. The resulting bead profile should be suitable for obtaining fusionby the subsequent pass to be initiated on the opposite side of the girder web. If theprofile is not conducive to good fusion, the start of the first root pass should beground, gouged, chipped or otherwise prepared to ensure adequate fusion.


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5. The second half of the weld joint should have the root pass applied before any otherweld passes are performed. The arc should be initiated at the end of the half lengthroot pass that is near the center of the beam flange, and travel should progress to theoutboard end of the joint, terminating on the weld tab.

6. Each weld layer should be completed on both sides of the joint before a new layer isdeposited.

7. Weld tabs should be removed and ground flush to the beam flange. Imperfectionsless than 1/16” should be removed by grinding. Deeper gouges, areas of lack offusion, slag inclusions, etc. should be removed by gouging or grinding and reweldingfollowing the procedures outlined above.

6.5.3 Weld Repair (Types W1, W2, or W3)

When W1, W2, or W3 cracks are found, the column base metal should be evaluated usingUT to determine if fractures have progressed into the flange. This testing should be performedboth during the period of discovery and during repair.

When a linear planar-type defect such as a crack or lack of fusion can be determined toextend beyond one-half the thickness of the beam flange, it is generally preferred to use adouble-sided weld for repair (even though the fracture may not extend all the way to the oppositesurface.) This is because the net volume of material that needs to be removed and restored isgenerally less when a double-sided joint is utilized. It also results in a better distribution ofresidual stresses since they are roughly balanced on either side of the center of the flangethickness.

Repair of these cracks may warrant total removal of the original weld, particularly if multiplecracks are present. If the entire weld plus some base metal is removed care must be taken not toexceed the root opening and bevel limits of AWS D1.1 unless a qualified by test WPS is used. Ifthis cannot be avoided one of two options is available:

1. The beveled face of the beam and/or the column face may be built up (buttered) until the desired root opening and angle is obtained.

2. A section of the flange may be removed and a splice plate inserted.

Commentary: Building up base metal with welding is a less intrusive techniquethan removing large sections of the base metal and replacing with new plate.However, this technique should not be used if the length of build-up exceeds thethickness of the plate.

6.5.4 Column Flange Repairs - Type C2

Damage type C2 is a pullout type failure of the column flange material. The zone should beconditioned to a concave surface by grinding and inspected for soundness using MT. The



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concave area may then be built up by welding. The joint contour described in the WPS shouldspecify a "boat shaped" section with a "U" shaped cross section and tapered ends. The weldpasses should be horizontal stringers placed in accordance with the WPS. Since stop/starts willoccur in the finished weld, care must be taken to condition each stop/start to removediscontinuities and provide an adequate contour for subsequent passes. The final surface shouldbe ground smooth and flush with the column face. This surface and immediate surrounding areashould be subjected to MT and UT.

6.6 STRUCTURAL MODIFICATION

6.6.1 Definition of Modification

Within the context of damage to WSMF connections, the term "structural modification"refers to alteration of the connection to improve its earthquake performance and that of thestructure as a whole. This typically involves substantial changes to the connection's geometry,capacity, or relevant limit states (e.g. flexural or shear strength or stiffness). Work that includesremoval of existing welds and replacement with welds of improved toughness and/orworkmanship is not considered modification under these Interim Guidelines.

Commentary: This term is contrasted with "repair," wherein the essentialbehavior of the connection is unchanged as a result of the repair effort.Geometrical or stiffness changes can involve spatial alterations to the elements ofthe connection, such as adding column stiffeners or the addition of newconnection elements, such as cover plates, upstanding ribs, side plates orhaunches. Changes to the connection's capacity, either in flexure or shear, mayoccur as a result of the addition of new connection components. Altering theconnection's relevant limit states may occur, for example, when the location of theplastic hinge is shifted away from its original location or the shear capacity of theconnection or one of its elements determines the behavior of the connection.

Much of the damage that occurred in the Northridge Earthquake has beenattributed to the presence of “crack like” conditions at the root of the completejoint penetration beam flange to column flange welds. These crack likeconditions included lack of fusion at the weld root as well as the presence ofpartially fused weld backing. Some engineers believe that if these crack-likeconditions are removed, substantial improvement in connection performance canbe obtained. SAC conducted specific testing in the Phase 1 program in whichsuch “dressing up” of these welds was performed. The performance of theconnection in these tests was mixed, and often not substantially improved relativeto that of connections in which the backing was left in place. Based on these tests,removal of weld backing, backgouging and repairing welds, and reinforcing witha fillet is not recommended as a means of connection modification, although it isan acceptable means of repair for joints with type W1 and W2 damage.


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Several engineers and researchers knowledgeable in fracture mechanics havesuggested that the standard, unreinforced moment connection could performacceptably if weld metal and base metal with adequate toughness wereincorporated, and beam flange to column joints are executed in such a mannerthat large crack-like discontinuities are not present (removal of backing and weldtabs, backgouging, and reinforcing with a fillet). Other engineers knowledgeablein mechanics of materials (Blodgett - 1994) believe that regardless of thetoughness of the weld metal employed, the connection configuration is such thatreliable performance is unlikely.

If joints with adequate weld metal toughness can provide substantially morereliable performance, then, removal of existing low-toughness welds andreplacement with new tough material may be an acceptable means ofmodification. To date, only limited testing of such assemblies have beenconducted. In one test (Popov - 1995) an assembly consisting of a W36 x 150beam connected to a W14 x 257 column and originally fabricated using E70T4electrodes (not having rated notch toughness) was repaired following initialtesting by completely removing the complete joint penetration welds of the beamflanges to column flanges and replacing them with new welds made withelectrodes having specified notch toughness. Weld backing and weld tabs wereremoved and the welds were reinforced with a fillet. The specimen wassuccessfully tested to a plastic rotation of 0.04 radians. However, until additionalresearch can be performed to quantify the reliability obtained through the use ofnotch tough weld metal, this is not recommended by itself as a method ofmodification in these Interim Guidelines.

Modification of the structure as a whole, as opposed to individual connectionmodifications, can be an effective means of obtaining more reliable performance.The addition of braced frames, shear walls, energy dissipation systems, baseisolation, etc., can be used to reduce the total deformation demand induced in thestructure by earthquakes, and consequently the need for the moment-resistingconnections to resist large plastic rotation demands. Interim Guidelines for thesetypes of modifications are not directly included in this document. However,sections on connection qualification presented below provide information thatcan be used to determine the plastic rotation capacity of existing connections inthe building. Once this is determined, the effectiveness of proposed globalmodification measures can be assessed, as part of the design process.

6.6.2 Damaged vs. Undamaged Connections

Engineers should inform building owners that substantial improvements in the reliability offuture earthquake performance of a WSMF building can be obtained by structural modification.Modification can be made at connections that have sustained damage as well as those that areundamaged. On the basis of cost, some owners may elect to modify those connections whichhave been damaged, and which will be repaired, but not other, undamaged connections. If a



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building has had only a few scattered connections damaged, such an approach will not result inany significant improvement in future building performance, and is not recommended. If asubstantial number of connections in a building have been damaged and will be repaired,modification of these damaged connections may improve future building performance,depending on the distribution of damaged connections, throughout the building. Therefore,consideration of such an approach has been recommended in Chapter 4 of these InterimGuidelines.

If possible, it is recommended that the modification of connections follow a rational spatialdistribution, so a to distribute the enhanced energy dissipation capacity (and ductility) throughoutthe building. As a minimum, structural modification should consider the effect of thosemodifications on the performance of the lateral system as well as on the performance ofindividual components of the frames. An appropriate analysis should be performed of thebuilding, considering the modifications, to ensure that undesirable stiffness irregularities are notintroduced or made more severe, and that excessive demand is not concentrated in connectionsunable to resist the applied loads or deformations. The effects of connection modifications oninelastic demands in adjacent columns and panel zones should be considered.

Commentary: Structural modification of connections will normally be performedas a means of enhancing the expected performance of the building in futureearthquakes, by minimizing the potential for fractures. The intent of modificationis to make the connection sufficiently strong that inelastic behavior of the framewill be controlled by the formation of plastic hinges within the girder spans.

Evaluation of statistical data on the types and distribution of damageexperienced by 89 buildings affected by the Northridge Earthquake (Bonowitz &Youssef - 1995) indicates that the spatial distribution of damage other than smallroot indications (Type W1) has modest correlation with the distribution of highseismic demands predicted by traditional analytical approaches. The distributionof type W1 indications appear to be random. A modification scheme that selectsconnections on the basis of existing damage could therefore result in a randomdistribution of connections with improved performance characteristics. In suchan approach, connections that may undergo high plastic rotation demands or maybe part of a lateral system with limited redundancy might not be modified in favorof connections damaged as a result of poor workmanship. The result of this couldbe a modified system with only marginally improved behavior. Connections thathave not been modified can be expected to have a significant failure rate insubsequent earthquakes, at near-elastic demand levels. Therefore, the amount ofimprovement obtained by modifying only the damaged connections is not directlyquantifiable. Generally, as more connections in the building are modified, thepotential performance of the building should improve.

An alternative approach, and one that appears to represent a more reliablemethod of ensuring that the earthquake performance of the lateral system isequivalent to that assumed at the time the WSMF was designed, is to modify all of


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the connections. Tests on girder-column connections similar to those found inmany buildings suggest that the traditional welded flange/bolted web connectioncannot develop the rotational demands implicit in building code designs.Modified connections appear to represent one approach to achieve the requiredlevel of deformation capacity.

Modification of only selected connections may be a cost-effective approach ifthe analysis can accurately predict the demand on the connections as well as theconsequences of future connection failures in the modified and unmodifiedconnections. The structural engineer should inform the building owner of theassumed benefits as well as the potential disadvantages of a scheme that modifiesonly a selected number of the connections. The reliability of analyses used tojustify such a partial modification scheme is sensitive to the modelingassumptions and the ground motion input.

6.6.3 Criteria

Connection modification intended to permit inelastic frame behavior should be proportionedso that the required plastic deformation of the frame may be accommodated through thedevelopment of plastic hinges at pre-determined locations within the girder spans, as indicated inFigure 6-12. Beam-column connections should be designed with sufficient strength (through theuse of cover plates, haunches, side plates, etc.) to force development of the plastic hinge awayfrom the column face. This condition may also be attained through local weakening of the beamsection, at the desired location for plastic hinge formation. All elements of the connectionshould have adequate strength to develop the forces resulting from the formation of the plastichinge at the predetermined location, together with forces resulting from gravity loads.

Plastic Hinges

Deformed frame shapeUndeformedframe

L’

L

h

drift angle - θ

Figure 6-12 - Desired Plastic Frame Behavior

Commentary: Nonlinear deformation of frame structures is accommodatedthrough the development of inelastic flexural or shear strains within discreteregions of the structure. At large inelastic strains these regions can develop into



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plastic hinges, which can accommodate significant concentrated rotations atconstant (or nearly constant) load through yielding at tensile fibers and bucklingat compressive fibers. If a sufficient number of plastic hinges develop in a frame,a mechanism is formed and the frame can deform laterally in a plastic manner.This behavior is accompanied by significant energy dissipation, particularly if anumber of members are involved in the plastic behavior, as well as substantiallocal damage to the highly strained elements. The formation of hinges incolumns, as opposed to beams, is undesirable, as this results in the formation ofmechanisms with relatively few elements participating, and consequently littleenergy dissipation occurring. In addition, such mechanisms also result in localdamage to critical gravity load bearing elements.

The prescriptive connection contained in the UBC and NEHRP RecommendedProvisions prior to the Northridge Earthquake was based on the development ofplastic hinges within the beams at the face of the column, or within the columnpanel zone itself. If the plastic hinge develops in the column panel zone, theresulting column deformation results in very large secondary stresses on thebeam flange to column flange joint, a condition which can contribute to brittlefailure. If the plastic hinge forms in the beam, at the face of the column, this canresult in very large through-thickness strain demands on the column flangematerial and large inelastic strain demands on the weld metal and surroundingheat affected zones. These conditions can also lead to brittle joint failure. Inorder to achieve more reliable performance, it is recommended that theconnection of the beam to the column be modified to be sufficiently strong to forcethe inelastic action (plastic hinge) away from the column face. Plastic hinges insteel beams have finite length, typically on the order of half the beam depth.Therefore, the location for the plastic hinge should be shifted at least thatdistance away from the face of the column. When this is done, the flexuraldemands on the columns are increased. Care must be taken to assure that weakcolumn conditions are not inadvertently created by local strengthening of theconnections.

It should be noted that connection modifications of the type described above,while believed to be effective in preventing brittle connection fractures, will notprevent structural damage from occurring. Brittle connection fractures areundesirable because they result in a substantial reduction in the lateral-force-resisting strength of the structure which, in extreme cases, can result in instabilityand collapse. Connections modified as described in these Interim Guidelinesshould experience many fewer such brittle fractures than unmodified connections.However, the formation of a plastic hinge within the span of a beam is not acompletely benign event. Beams which have formed such hinges may exhibitlarge buckling and yielding deformation, damage which typically must berepaired. The cost of such repairs could be comparable to the costs incurred inrepairing fracture damage experienced in the Northridge Earthquake. Theprimary difference is that life safety protection will be significantly enhanced and


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most structures that have experienced such plastic deformation damage shouldcontinue to be safe for occupancy, while repairs are made.

If the types of damage described above are unacceptable for a given building,then alternative methods of structural modification should be considered, that willreduce the plastic deformation demands on the structure during a strongearthquake. Appropriate methods of achieving such goals include the installationof supplemental braced frames, energy dissipation systems, and similarsystematic modifications of the building’s basic lateral force resisting system.

6.6.4 Strength

When these Interim Guidelines require determination of the strength of a framing element orcomponent, this shall be calculated in accordance with the criteria contained in UBC-94, Section2211.4.2 {NEHRP-91 Section 10.2, except that the factor φ should be taken as 1.0}, restated asfollows:

2211.4.1 Member strength. Where this section requires that the strength of the member bedeveloped, the following shall be used:

Flexure Ms = Z FyShear Vs = 0.55 Fy d tAxial compression Psc = 1.7 Fa AAxial tension Pst = Fy AConnectors Full Penetration welds Fy A Partial Penetration welds 1.7 allowable (see commentary) Bolts and fillet welds 1.7 allowable

Commentary: Partial penetration welds are not recommended for tensionapplications in critical connections resisting seismic induced stresses. Thegeometry of partial penetration welds creates a notch-like condition that caninitiate brittle fracture under conditions of high tensile strain.

6.6.5 Plastic Rotation Capacity

The plastic rotation capacity of modified connections should reflect realistic estimates of therequired level of plastic rotation demand. In the absence of detailed calculations of rotationdemand, connections should be shown to be capable of developing a minimum plastic rotationcapacity on the order of 0.025 to 0.030 radian. The demand may be lower when braced frames,supplemental damping, base isolation, or other elements are introduced into the moment framesystem, to control its lateral deformation; when the design ground motion is relatively low in therange of predominant periods for the structure; and when the frame is sufficiently strong.

If calculations are performed to determine the required connection plastic rotation capacity,the capacity should be taken somewhat greater than the calculated deformation demand, due tothe high variability and uncertainty inherent in predictions of inelastic seismic response. Until



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better guidelines become available, a required plastic rotation capacity on the order of 0.005radians greater than the demand calculated for the design basis earthquake (or if greaterconservatism is desired - the maximum capable earthquake) is recommended. Rotation demandcalculations should consider the effect of plastic hinge location within the beam span, asindicated in Figure 6-12, on plastic rotation demand. Calculations should be performed to thesame level of detail specified for nonlinear dynamic analysis for base isolated structures in UBC-94 Section 1655 {NEHRP-94 Section 2.6.4.4}. Ground motion time histories utilized for thesenonlinear analyses should satisfy the scaling requirements of UBC-94 Section 1655.4.2{NEHRP-94 Section 2.6.4.4} except that instead of the base isolated period, TI, the structureperiod, T, calculated in accordance with UBC-94 Section 1628 {NEHRP-94 Section 2.3.3.1}should be used.

Commentary. Traditionally, structural engineers have calculated demand inmoment frames by sizing the members for strength and drift using code forces(either equivalent static or reduced dynamic forces) and then "developing thestrength of the members." Since 1988, "developing the strength" has beenaccomplished by prescriptive means. It was assumed that the prescribedconnections would be strong enough so that the girder would yield (in bending),or the panel zone would yield (in shear) in a nearly perfectly plastic mannerproducing the plastic rotations necessary to dissipate the energy of theearthquake. It is now known that the prescriptive connection is often incapable ofbehaving in this manner.

In the 1994 Northridge earthquake, many moment-frame connectionsfractured with little evidence of plastic hinging of the girders or yielding of thecolumn panel zones. Testing of moment frame connections both prior to andsubsequent to the earthquake suggests that the standard welded flange-bolted webconnection is unable to reliably provide plastic rotations beyond about 0.005radian for all ranges of girder depths and often fails below that level. Thus, forframes designed for code forces and for the code drift limits, new connectionconfigurations must be developed to reliably accommodate such rotation withoutbrittle fracture.

In order to develop reasonable estimates of the plastic rotation demands on aframe’s connections, it is necessary to perform inelastic time history analyses.For regular structures, approximations of the plastic rotation demands can beobtained from linear elastic analyses. Analytical research (Newmark and Hall -1982) suggests that for structures having the dynamic characteristics of mostWSMF buildings, and for the ground motions typical of western US earthquakes,the total frame deflections obtained from an unreduced (no R or Rw factor)dynamic analysis provide an approximate estimate of those which would beexperienced by the inelastic structure. For the typical spectra contained in thebuilding code, this would indicate expected drift ratios on the order of 1%. Thedrift demands in a real structure, responding inelastically tend to concentrate in afew stories, rather than being uniformly distributed throughout the structure’s


6-29

height. Therefore, it is reasonable to expect typical drift demands in individualstories on the order of 1.5% to 2% of the story height. As a rough approximation,the drift demand may be equated to the joint rotation demand, yielding expectedrotation demands on the order of perhaps 2%. Since there is considerablevariation in ground motion intensity and spectra, as well as the inelastic responseof buildings to these ground motions, conservatism in selection of an appropriateconnection rotation demand is warranted.

In recent testing of large scale subassemblies incorporating modifiedconnection details, conducted by SAC and others, when the connection designwas able to achieve a plastic rotation demand of 0.025 radians or more forseveral cycles, the ultimate failure of the subassembly generally did not occur inthe connection, but rather in the members themselves. Therefore, the statedconnection capacity criteria would appear to result in connections capable ofproviding reliable performance.

It should be noted that the connection assembly capacity criteria for themodification of existing buildings, recommended by these Interim Guidelines, issomewhat reduced compared to that recommended for new buildings (Chapter 7).This is typical of approaches normally taken for existing structures. For newbuildings, these Interim Guidelines discourage building-specific calculation ofrequired plastic rotation capacity for connections and instead, encourage thedevelopment of highly ductile connection designs. For existing buildings, such anapproach may lead to modification designs that are excessively costly, as well asthe modification of structures which do not require such modification.Consequently, an approach which permits the development of semi-ductileconnection designs, with sufficient plastic rotation capacity to withstand theexpected demands from a design earthquake is adopted. It should be understoodthat buildings modified to this reduced criteria will not have the same reliabilityas new buildings, designed in accordance with the recommendations of Chapter7. The criteria of Chapter 7 could be applied to existing buildings, if superiorreliability is desired.

When performing inelastic frame analysis, in order to determine the requiredconnection plastic rotation capacity, it is important to accurately account for thelocations at which the plastic hinges will occur. Simplified models, whichrepresent the hinge as occurring at the face of the column, will underestimate theplastic rotation demand. This problem becomes more severe as the columnspacing, L, becomes shorter and the distance between plastic hinges, L’, agreater portion of the total beam span. In extreme cases, the girder will not formplastic hinges at all, but instead, will develop a shear yield, similar to aneccentric braced frame.



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6.6.6 Connection Qualification and Design

Modified girder-column connections may be qualified by testing or designed usingcalculations. Qualification by testing is the preferred approach. Preliminary designs ofconnections to be qualified by test may be obtained using the calculation procedures of Section6.6.6.3. The procedures of that section may also be used to calibrate previous tests of similarconnection configurations to slightly different applications, by extrapolation. Extrapolation oftest results should be limited to connections of elements having similar geometries and materialspecifications as the tested connections. Designs based on calculation alone should be subject toqualified independent third party review.

6.6.6.1 Qualification Test Protocol

Unless future testing programs reveal significant effects of dynamic loading rate or timehistory loading, a testing protocol similar to ATC-24, Guidelines for Cyclic Seismic Testing ofComponents of Steel Structures (Applied Technology Council - 1992), is recommended as thebasis for qualification tests.

The testing program should replicate as closely as practical the anticipated conditions in thefield, including such factors as:

a) Member sizes.

b) Material specifications.

c) Welding process, details and construction conditions.

d) Cover-plates, continuity plates, web tabs, bolts, and doubler plates.

e) Connection configuration (e.g., beams on both sides).

f) Induced stresses because of restraint conditions on the welds and connectionmembers.

g) Axial load, where pertinent.

h) Gravity load, where significant.

The testing program should be organized to provide as much information as possible aboutthe capability of the connections selected. The following minimum program is recommended:

a) Test two full size specimens of the largest representative beam/column assembly inthe project.

b) Test one additional full size specimen for each beam/column assembly withsignificantly different interaction properties, such as beam flange width-thickness(b/t) ratio, panel zone stress/distortion, etc.

If any of the specimens fails to meet the qualification criteria, the connection should beredesigned and retested.


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Where two-sided connections are used in the structure, and the type of connection being usedcan be expected to perform differently in a two-sided use than in one-sided use, it should betested in the two-sided configuration as well as the one-sided. Two-sided connection assembliescan be expected to behave differently than one-sided assemblies, for example, when panel zonedistortions will be significantly different, or when systems involve transfer of stress to thecolumn by plates, welds, or other elements which are connected to the beams on both sides of thecolumn.

Testing to include axial load should be considered when analysis indicates that significanttension can be expected to occur in a significant number of the columns represented by thespecimen and where the connection type relies on the through-thickness strength of the columnflanges. If the presence of a floor slab is anticipated to have significant influence on either thelocation or mechanism of the plastic hinge formed, than this should also be included in the testspecimen.

Commentary: The use of connection configurations that have been qualified bytest is the preferred approach. While the testing of all connection geometries andmember combinations in any given building is not practical, the number of testsmust be large enough to be meaningful yet small enough to not be unreasonablycostly. Testing, within the limitations of test specimen simplification, has theadvantage of being able to replicate fabrication and welding procedures, jointgeometry and member size, and potential modes of failure. If the testing is donein a manner consistent with other testing programs, reasonable comparisons canbe made. On the other hand, testing is expensive and it is difficult to realisticallytest the girder-column connection using actual restraint conditions andearthquake loading rates. Calculations offer an economical alternative to testingthat can accommodate different girder and column sizes, altered connectiongeometries, and member properties. Nevertheless, recent testing on girder-column connections from WSMFs casts doubt on some fundamental assumptionsupon which the calculations are based and therefore, they should be used withcaution.

Since the level of confidence in connections developed strictly on the basis ofcalculations may not be as high as those based on tests, the use of testing isencouraged. Tests are, however, relatively expensive and a reasonable degree offlexibility in interpreting the results of limited testing programs must beacknowledged.

How much extrapolation should be accepted is a difficult decision. Asadditional testing is done, more information may be available on what constitutes"conservative" testing conditions, thereby allowing easier decisions relative toextrapolating tests to actual conditions which are likely to be less demanding thanthe tests. For example, it is hypothesized that connections of shallower, thinnerflanged members are likely to be more reliable than similar connectionsconsisting of deeper, thicker flanged members. Thus, it may be possible to test the



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largest assemblages of similar details and extrapolate to the smaller membersizes -- at least in comparable member group families.

6.6.6.2 Acceptance Criteria

The minimum acceptance criteria for connection qualification for specimens tested inaccordance with these Interim Guidelines should be as follows:

a) The connection should develop beam plastic rotations as indicated in Section 6.6.5,for at least one complete cycle.

b) The connection should develop a minimum strength equal to 80% of the plasticstrength of the girder, calculated using minimum specified yield strength Fy,throughout the loading history required to achieve the required plastic rotationcapacity, as indicated in a), above.

Commentary: Many connection configurations will be able to withstand plasticrotations on the order of 0.025 radians or more, but will have sustainedsignificant damage and degradation of stiffness and strength in achieving thisdeformation. The intent of the acceptance criteria presented in this Section is toassure that when connections experience the required plastic rotation demand,they will still have significant remaining ability to participate in the structure’slateral load resisting system.

6.6.6.3 Calculations

All connections designs should be based on test data and the use of connections based uponcalculations only is not recommended. An approved program of variations on the testedproto-typical connections may use calculations to assist in extrapolation of results.

Calculations should be correlated to tested material properties for base metals and welds.The properties should be those corresponding to the axes of loading of the base metal and weldin the joints and to the welding processes and materials intended for use. The tested propertiesmay be specific to the materials and processes to be used in the project, or based on astatistically-based testing program. Use of properties inferred from other testing programs mustbe done with appropriate care and, where such inferred properties are used, designs shouldreflect the uncertainty inherent in such an indirect approach.

Calculations should initiate with the selection of a connection configuration, such as one ofthose indicated in Section 6.6.7, that will permit the formation of a plastic hinge within the beamspan, away from the face of the column, when the frame is subjected to gravity and lateral loads.

6.6.6.3.1 Material Strength Properties

In the absence of project specific material property information (for example, mill testreports), the values listed in Table 6-3 should be used to determine the strength of steel shape and


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plate for purposes of calculation. The permissible strength for weld metal should be taken inaccordance with the building code.

Table 6-3 - Properties for Use in Connection Modification Design

Material Fy (ksi) Fy m (ksi) Fu (ksi)A36 Beam 36 1 1

Dual Certified Beam Axial, Flexural Shape Group 1 Shape Group 2 Shape Group 3 Shape Group 4 Through-Thickness

50

-

552

582

572

542

-

65 min.

Note 3A572 Column/Beam Axial, Flexural Shape Group 1 Shape Group 2 Shape Group 3 Shape Group 4 Shape Group 5 Through-Thickness

50

-

582

582

572

572

552

-

65 min.

Note 3,

Notes:1. See Commentary2. Based on coupons from web. For thick flanges,

the Fy flange is approximately 0.95 Fy web.3. See Commentary

Commentary: Table 6-3, Note 1 - The material properties for steel nominallydesignated on the construction documents as ASTM A36 can be highly variableand in recent years, steel meeting the specified requirements for both ASTM A36and A572 has routinely been incorporated in projects calling for A36 steel.Consequently, unless project specific data is available to indicate the actualstrength of material incorporated into the project, the properties for ASTM A572steel should be assumed when ASTM A36 is indicated on the drawings, and theassumption of a higher yield stress results in a more severe design condition.

Table 6-3, Note 3 - The causes for through-thickness failures of columnflanges (types C2, C4, and C5), observed both in buildings damaged by theNorthridge Earthquake and in some test specimens, are not well understood.They are thought to be a function of the metallurgy and “purity” of the steel;conditions of loading including the presence of axial load and rate of loadingapplication; conditions of tri-axial restraint; conditions of local hardening andembrittlement within the weld’s heat affected zone; and by the relationship of theconnection components as they may affect flange bending stresses and flangecurvature induced by panel zone yielding. Given the many complex factors whichcan affect the through-thickness strength of the column flange, determination of areliable basis upon which to set permissible design stresses will requiresignificant research.



6-34

Interim Recommendation No. 2 (SEAOC-1995) included a value of 40 ksi,applied to the projected area of beam flange attachment, for the through-thickness strength to be used in calculations. This value was selected because itwas consistent with the successful tests of cover plated assemblies conducted atthe University of Texas at Austin (Engelhardt and Sabol - 1994). However,because of the probable influence of all the factors noted above, this value canonly be considered to reflect the specific conditions of those tests and specimens.

Although reduced stresses at the column face produced acceptable results inthe University of Texas tests, the key to that success was more likely the result offorcing the plastic hinge away from the column than reduction of the through-thickness stress by the cover plates. Reduction of through-thickness columnflange stress to ever lower levels by the use of thicker cover plates is notrecommended, since such cover plates will result in ever higher forces on the faceof the column flange.

Notwithstanding all of the above, successful tests using cover plates and othermeasures of moving hinges (and coincidentally reducing through-thickness stress)continue to be performed. In the interim, structural engineers choosing to utilizeconnections relying on through-thickness strength should recognize that despitethe successful testing, connections relying on through-thickness strength can notbe considered to be fully reliable until the influence of the other parametersdiscussed above can be fully understood. A high amount of structural redundancyis recommended for frames employing connections which rely on through-thickness strength of the column flange.

6.6.6.3.2 Determine Plastic Hinge Location

The desired location for the formation of plastic hinges should be determined as a basicparameter for the calculations. For beams with gravity loads representing a small portion of thetotal flexural demand, the plastic hinge may be assumed to occur at a distance equal to 1/3 of thebeam depth from the edge of the reinforced connection (or start of the weakened beam section),unless specific test data for the connection indicates that a different value is appropriate. Referto Figure 6-13.


6-35

LB

eam

dep

th -

d

Edge

of r

einf

orce

dco

nnec

tion

Edge

of r

einf

orce

dco

nnec

tion

d/3

L’

Plastichinge Connection

reinforcement

Figure 6-13 - Location of Plastic Hinge

Commentary: The suggested location for the plastic hinge, at a distance d/3 awayfrom the end of the reinforced section is based on the observed behavior of testspecimens, with no significant gravity load present. If the significant gravity loadis present, this can shift the locations of the plastic hinges, and in the extremecase, even change the form of the collapse mechanism. If flexural demand on thegirder due to gravity load is less than about 30% of the girder plastic capacity,this effect can safely be neglected, and the plastic hinge locations taken asindicated. If gravity demands significantly exceed this level then plastic analysisof the girder should be performed to determine the appropriate hinge locations.Note that in zones of high seismicity (UBC Zones 3 and 4, and NEHRP MapAreas 6 and 7) gravity loading on the girders of earthquake resisting framestypically has a very small effect.

6.6.6.3.3 Determine Probable Plastic Moment at Hinges

The probable value of the plastic moment, Mpr, at the location of the plastic hinges should bedetermined from the equation:

M 0.95 Z Fpr b ya= α (6-1)

where: α is a coefficient that accounts for the effects of strain hardening and modelinguncertainty, taken as:

1.1 when qualification testing is performed or calculations arecorrelated with previous qualification testing

1.3 when design is based on calculations, alone.



6-36

Fya is the actual yield stress of the material, as identified from mill test reports. Where milltest data for the project is not traceable to specific framing elements, the averageof mill test data for the project for the given shape may be used. When mill testdata for the project is not available, the value of Fym, fromtable 6-3 may be used.

Zb is the plastic modulus of the section

Commentary: The 0.95 factor, in equation 6-1, is used to adjust the yield stress inthe beam web, where coupons for mill certification tests are normally extracted,to the value in the beam flange. Beam flanges, being comprised of thickermaterial, typically have somewhat lower yield strengths than do beam webmaterial.

The factor of 1.1 recommended to account for strain hardening, or othersources of strength above yield, agrees fairly well with available test results. Itshould be noted that the 1.1 factor could underestimate the over-strength wheresignificant flange buckling does not act as the gradual limit on the connection.Nevertheless, the 1.1 factor seems a reasonable expectation of over-strengthconsidering the complexities involved.

Connection designs that result in excessive strength in the girder connectionrelative to the column or excessive demands on the column panel zone are notexpected to produce superior performance. There is a careful balance that mustbe maintained between developing connections that provide for an appropriateallowance for girder overstrength and those that arbitrarily increase connectiondemand in the quest for a “conservative” connection design. The factorssuggested above were chosen in an attempt to achieve this balance, and arbitraryincreases in these values are not recommended.

6.6.6.3.4 Determine Beam Shear

The shear in the beam, at the location of the plastic hinge should be determined. A free bodydiagram of that portion of the beam located between plastic hinges is a useful tool for obtainingthe shear at each plastic hinge. Figure 6-14 provides an example of such a calculation.


6-37

L

d/3L’

Plastichinge P

L/2

P

Mpr MprL’Vp

taking the sum of moments about “A” = 0Vp ={Mpr + Mpr + P L’/2 + wL’2/2}/L’

“A”

VAw

Note: if 2Mpr /L’ is less then the gravity shear in the free body (in thiscase P/2 + wL’/2),then the plastic hinge location will shift and L’must be adjusted, accordingly

Figure 6-14 - Sample Calculation of Shear at Plastic Hinge

6.6.6.3.5 Determine Strength Demands on Connection

In order to complete the design of the connection, including sizing the various plates andjoining welds which make up the connection, it is necessary to determine the shear and flexuralstrength demands at each critical section. These demands may be calculated by taking a freebody of that portion of the connection assembly located between the critical section and theplastic hinge. Figure 6-15 demonstrates this procedure for two critical sections, for the beamshown in Figure 6-14.

Plastichinge

Vp

Mpr

Plastichinge

Vp

Mpr

x

Mf

x+dc/2

dc

Mf=Mpr +Vpx

Mc

Mf=Mpr +Vp(x+dc/2)

Critical Section at Column Face Critical Section at Column Centerline

Figure 6-15 - Calculation of Demands at Critical Sections

Commentary: Each unique connection configuration may have different criticalsections. The vertical plane that passes through the joint between the beamflanges and column (if such joining occurs) will typically define at least one such



6-38

critical section, used for designing the joint of the beam flanges to the column, aswell as evaluating shear demands on the column panel zone. A second criticalsection occurs at the center line of the column. Moments calculated at this pointare used to check weak beam - strong column conditions. Other critical sectionsshould be selected as appropriate to the connection configuration.

6.6.6.3.6 Check for Strong Column - Weak Beam Condition

Buildings which form sidesway mechanisms through the formation of plastic hinges in thebeams can dissipate more energy than buildings that develop mechanisms consisting primarily ofplastic hinges in the columns. Therefore, if an existing building’s original design was such thathinging would occur in the beams rather than the columns, care should be taken not to alter thisbehavior with the addition of connection reinforcement. To determine if the desired strongcolumn - weak beam condition exists, the connection assembly should be checked to determineif the following equation is satisfied:

Z (F f ) M 1.0c yc a c− >∑ ∑ (6-2)

where: Zc is the plastic modulus of the column section above and below the connectionFyc is the minimum specified yield stress for the column above and belowfa is the axial load in the column above and belowMc is the moment calculated at the center of the column in accordance with Section 6.6.6.3.5

Commentary: Equation 6-2 is based on the building code provisions for strongcolumn - weak beam design. The building code provisions for evaluating strongcolumn - weak beam conditions presume that the flexural stiffness of the columnsabove and below the beam are approximately equal. If non-symmetricalconnection configurations are used, such as a haunch on the bottom side of thebeam, this can result in an uneven distribution of stiffness between the two columnsegments. In such cases, a plastic analysis should be considered to determine ifan undesirable story mechanism is likely to form in the building.

6.6.6.3.7 Check Column Panel Zone

The adequacy of the shear strength of the column panel zone should be checked. For thispurpose, the term 0.8ΣMf should be substituted for the term 0.8ΣMs in UBC-94 Section2211.7.2.1 {0.9ΣφbMp in NEHRP-91 Section 10.10.3.1} repeated below for convenience ofreference. Mf is the calculated moment at the face of the column, when the beam mechanismforms, calculated as indicated in Section 6.6.6.3.5, above.

2211.7.2.1 Strength. The panel zone of the joint shall be capable of resisting the shear induced bybeam bending moments due to gravity loads plus 1.85 times the prescribed seismic forces, but theshear strength need not exceed that required to develop 0.8ΣMs of the girders framing into thecolumn flanges at the joint. The joint panel zone shear strength may be obtained from thefollowing formula:


6-39

V 0.55F d t 13b td d ty c

c c f2

b c

= +

(11-1)

where: bc = width of column flangedb = the depth of the beam (including haunches or cover plates)dc = the depth of the columnt = the total thickness of the panel zone including doubler platestcf = the thickness of the column flange

6.6.7 Modification Details

There are many potential details that can be used to modify the performance of girder-column joints in existing WSMF structures. Several of these have been tested as part of the SACPhase 1 effort. While these repair and modification configurations do not represent all potentialgeometries and the number of replicates is very limited, these tests do provide important insightinto the behavior of the modified connection configurations. Figures shown below presentconceptual connection configurations that have been subjected to limited testing and have shownan acceptable level of performance.

Reference to laboratory testing is provided for those connection configurations for whichresearch has been reported. However, it should be noted that none of these connections has beentested sufficiently at this time to permit unqualified use of the connection.

The figures provided in the following sections are schematic, indicating the general type ofconnection configuration being described. When designing connections patterned after the reportedtest data, the test specimen details included in the references should be reviewed to determinespecific details not shown.

The SAC Joint Venture does not endorse or specifically recommend any of the connectiondetails shown in this Section. These are presented only to acquaint the reader with availableinformation on representative testing of different connection configurations that have beenperformed by various parties.

6.6.7.1 Haunch at Bottom Flange

Figure 6-16 illustrates two alternative configurations of this detail that have been tested(Uang - 1995). The basic concept is to reinforce the connection with the provision of a triangularhaunch at the bottom flange. The intended behavior of both configurations is to shift the plastichinge from the face of the column and to reduce the demand on the CJP weld by increasing theeffective depth of the section. In one test, shown on the left of Figure 6-16, the joint between thegirder bottom flange and column was cut free, to simulate a condition which might occur if thebottom joint had been damaged, but not repaired. In a second tested configuration, the bottomflange joint was repaired and the top flange was replaced with a locally thickened plate, similarto the detail shown in Figure 6-9.



6-40

WT, trimmed

12

dd/

4

Thickened flange

Bottom Flange Not AttachedTop Flange Not Reinforced

Bottom Flange AttachedTop Flange Reinforced

oror

Figure 6-16 - Bottom Haunch Connection Modification

Design Issues: This approach developed acceptable levels of plastic rotation. Acceptable levelsof connection strength were also maintained during large inelastic deformations of the plastichinge. This approach does not require that the top flange be modified, or slab disturbed, unlessother conditions require repair of the top flange, as in the detail on the left of Figure 6-16. Thebottom flange is generally far more accessible than the top flange because a slab does not haveto be removed. In addition, the haunch can be installed at perimeter frames without removal ofthe exterior building cladding. There did not appear to be any appreciable degradation inperformance when the bottom beam flange was not re-welded to the face of the column.Eliminating this additional welding should help reduce the cost of the repair.

Performance is dependent on properly executed complete joint penetration welds at the columnface and at the attachment of the haunch to the girder bottom flange. The joint can be subject tothrough-thickness flaws in the column flange; however, this connection may not be as sensitiveto this potential problem because of the significant increase in the effective depth of the beamsection which can be achieved. Welding of the bottom haunch requires overhead welding. Theskewed groove welds of the haunch flanges to the girder and column flanges may be difficult toexecute.

Experimental Results: This approach developed excellent levels of plastic rotation. In Specimen1, the bottom flange CJP weld was damaged in a prior test but was not repaired: only the bottomhaunch was added. During the test of specimen 1, a slowly growing crack developed at theunderside of the top flange-web intersection, perhaps exacerbated by significant local bucklingof the top flange. Some of the buckling may be attributed to lateral torsional buckling thatoccurred because the bottom flange was not restrained by a CJP weld. A significant portion ofthe flexural strength was lost during the cycles of large plastic rotation. In the second specimen,the bottom girder flange weld was intact during the haunch testing, and its performance wassignificantly improved compared with the first specimen. The test was stopped when significantlocal buckling led to a slowly growing crack at the beam flange and web intersection. At thistime, it appears that repairing damaged bottom flange welds in this configuration can produce


6-41

better performance. Acceptable levels of flexural strength were maintained during largeinelastic deformations of the plastic hinge for both specimens.

Quantitative Results: No. of specimens tested: 2Girder Size: W30 x 99Column Size: W14 x 176Plastic Rotation achieved-

Specimen 1:0.04 radian (w/o bottom flange weld)Specimen 2:0.05 radian (with bottom flange weld)

6.6.7.2 Top and Bottom Haunch

Figure 6-17 illustrates the basic geometry of the top and bottom haunch detail. The intendedbehavior of this modification is to shift the plastic hinge from the face of the column and toreduce the demand on the CJP weld to the column flange by increasing the effective depth of thesection. As opposed to the bottom-only haunch, of Section 6.6.7.1, this detail further reduces thedemand on all CJP welds and allows for the structural engineer to introduce filler metal withbetter toughness properties into all critical joints, without necessarily having to remove the topflange CJP weld.

WT

12

dd/

3

or

Figure 6-17 - Top and Bottom Haunch Modification Detail

Two specimens for this detail have been tested to date, with excellent results. Possiblevariations that have not yet been tested include using a shallower haunch at the top flange,substitution of a flat cover plate for the top haunch, and not rewelding either of the originalgirder flanges to the column, if these have been damaged.

Design Issues: The haunches can be installed at perimeter frames without removal of theexterior building cladding. Performance is dependent on the proper execution of the CJP weldsfrom the haunch to the girder and column flanges, which can be difficult. The joint at thecolumn flange is subject to through-thickness flaws in the column flange, however, due to theadditional depth of the section at this joint, and the resulting reduced stresses, this design maynot be particularly sensitive to this.



6-42

Experimental Results: This approach developed excellent levels of plastic rotation in twospecimens. The tests were terminated when fractures across the width of the column flangesdeveloped at the locations of severe buckling in these flanges. Acceptable levels of connectionstrength were maintained throughout the test.


Specimen 1:0.07 radianSpecimen 2:0.07 radian

6.6.7.3 Cover Plate Sections

Figure 6-18 illustrates the basic configurations of cover plate connections. The assumptionbehind the cover plate is that it reduces the demand on the weld at the column flange and shiftsthe plastic hinge away from the column face. Only the connection with cover plates on the topof the top flange has been tested. There are no quantitative results for cover plates on the bottomside of the top flange, such as might be used in repair. It is likely that thicker plates would berequired where the plates are installed on the underside of the top flange. The implications ofthis deviation from the tested configuration should be considered.

Top &Bottom

Near and Far Sides

Top &Bottom

d

d/2, typical

Figure 6-18 - Cover Plate Connection Modification

Design Issues: Approximately eight connections similar that shown in Figure 6-18 have beentested (Engelhardt & Sabol - 1994), and have demonstrated the ability to achieve acceptablelevels of plastic rotation provided that the beam flange to column flange welding is correctlyexecuted and through-thickness problems in the column flange are avoided. The option with thetop flange cover plate located on top of the flange can be used on perimeter frames where accessto the outer side of the beam is restricted by existing building cladding. The option with thecover plate for the top flange located beneath the flange can be installed without requiringmodification of the slab. In the figures shown, the bottom cover plate is rectangular, and sized


6-43

slightly wider than the beam flange to allow downhand fillet welding of the joint between the twoplates. Some configurations using triangular plates at the bottom flange, similar to the topflange have also been tested.

Designers using this detail are cautioned to be mindful of not making cover plates so thickthat excessively large welds of the beam flange combination to column flange result. As thecover plates increase in size, the weld size must also increase. Larger welds invariably result ingreater shrinkage stresses and increased potential for cracking prior to actual loading. Inaddition, larger welds will lead to larger heat affected zones in the column flange, a potentiallybrittle area.

Performance is dependent on properly executed girder flange welds. The joint can be subjectto through-thickness failures in the column flange. Access to the top of the top flange requiresdemolition of the existing slab. Access to the bottom of the top flange requires overhead weldingand may be problematic for perimeter frames. Costs are greater than those associated withapproaches that concentrate modifications on the bottom flange

Experimental Results: Six of eight connections tested by the University of Texas at Austin wereable to achieve plastic rotations of at least 0.025 radians, or better. These tests were performedusing heavy column sections which forced nearly all of the plastic deformation into the beamplastic hinge; very little column panel zone deformation occurred. Strength loss at the extremelevels of plastic rotation did not reduce the flexural capacity to less than the plastic momentcapacity of the section based on minimum specified yield strength. One specimen achievedplastic rotations of 0.015 radians when a brittle fracture of the CJP weld (type W2 failure)occurred. This may partially be the result of a weld that was not executed in conformance withthe specified welding procedure specification. The second unsuccessful test specimen achievedplastic rotations of 0.005 radian when a section of the column flange pulled out (type C2failure). The successful tests were terminated either when twisting of the specimen threatened todamage the test setup or the maximum stroke of the loading ram was achieved.

Quantitative Results: No. of specimens tested: 8Girder Size: W36 x 150Column Size: W14 x 455, and 426Plastic Rotation achieved-

6 Specimens : >.025 radian to 0.05 radian1 Specimen: 0.015 radian (W2 failure)1 Specimen: 0.005 radian (C2 failure)

6.6.7.4 Upstanding Ribs

Figures 6-19 illustrates the basic configuration of connections with upstanding ribs. Theassumption behind the rib plate is that it reduces the demand on the weld at the column flangeand shifts the plastic hinge from the column face. The figure indicates alternative configurationsusing either one centered rib, or two spaced ribs on each flange. Test data is available only forthe case with two ribs.



6-44

21

Typical

Typical

tested configuration

alternate configuration

dd/2

Figure 6-19 - Upstanding Rib Connection Modification

Design Issues: Two connections similar to Figure 6-19, with two spaced ribs at each flangehave been tested (Engelhardt & Sabol - 1994), and demonstrated the ability to achieveacceptable levels of plastic rotation provided that the girder flange welding is correctlyexecuted. This modification can be used on perimeter frames where access to the outer side ofthe girder is restricted by existing building cladding.

Performance is dependent on properly executed girder flange welds. The joint can be subject tothrough-thickness failures in the column flange. Access to the top of the top flange requiresdemolition of the existing slab. Access to the bottom of the top flange requires overhead weldingand may be problematic for perimeter frames. The size of the specimens tested required the useof two upstanding ribs per flange. This increased the costs significantly above those designs thatuse only one rib per flange, located above the girder center line. However, limited testing of thedesign with one rib at the girder centerline, performed as part of a program related to eccentricbraced frames, indicated the potential for premature failure of the weld of the rib to the girder atthe outstanding edge.

Experimental Results: Two connections have been tested (Engelhardt & Sabol - 1994) using twoplates on the top and bottom flanges. The columns used in the test were very heavy and theflanges were able to resist the applied loads from the ribs without distorting. Similarperformance might not occur with lighter column sections. In addition, the size of the columnsforced all of the plastic deformation into the beam plastic hinge; very little column panel zonedeformation occurred. Strength loss at the extreme levels of plastic rotation did not reduce theflexural capacity to less than the plastic moment capacity of the section based on minimumspecified yield strength, however, strength loss occurred more quickly than with the cover platedspecimens. The tests were terminated when a slow tear of the beam bottom flange occurred atthe tips of the ribs.

Quantitative Results: No. of specimens tested: 2Girder Size: W36 x 150Column Size: W14 x 426


6-45

Plastic Rotation achieved-2 Specimens : >.025 radian

6.6.7.5 Side-Plate Connections

This approach eliminates loading the column in the through-thickness direction by removingthe CJP welds at the girder flange and by shifting the plastic hinge from the column face. Thetension and compression forces are transferred from the girder flanges into the column throughfillet welds. A mechanism to provide a direct connection between the column panel zone and thebeam flanges is required; the difficulty appears to be equalizing the width of the beam andcolumn flanges.

Experimental Results: At least two configurations of side-plated connections have been tested.One set, shown in Figure 6-20, utilized flat bars at the top and bottom girder flanges, to transferflange forces to the column (Engelhardt & Sabol - 1994). The girder was widened to the widthof the column with the use of filler plates. The specimens achieved plastic rotations of 0.015radians, however, fractures developed within the welds connecting the beam flange to thetransfer plates. Failure of the shear tab, and finally the side plates themselves followed theinitiation of these fractures. It is believed that the unsuccessful behavior of this particularspecimen was related to the method used to increase the width of the beam flange to equal that ofthe column flange, using a combination of a filler bar and welding. Other approaches that relyon a flat filler plate to transfer the forces may perform better.

Possible AlternativeTested Configuration

Figure 6-20 - Side Plate Connection Modification

Quantitative Results:Separate Top & Bottom Side PlatesNo. of specimens tested: 2Girder Size: W36 x 150Column Size: W14 x 426Plastic Rotation achieved-

2 Specimens : >.015 radian



6-46

A second, proprietary configuration, is shown in Figure 6-21. Three specimensrepresentative of the “new structure” configuration have undergone full-scale testing to date andachieved large plastic rotations. Loss of strength at large plastic rotation demands wascomparable to that of other successful connections. No tests have yet been conducted of therepair configuration. The developer of this connection has applied for US and foreign patents.Further information on technical data for this configuration, and license fees, may be obtainedfrom the developer.

New Building Configuration Repair Configuration

WARNING: The information presented in this figure is PROPRIETARY. US and Foreign Patents havebeen applied for. Use of this information is strictly prohibited except as authorized in writing by thedeveloper. Violators shall be prosecuted in accordance with US and Foreign Patent Intellectual PropertyLaws.

Figure 6-21 - Proprietary Side Plate Connection Modification

Design Issues: Testing of three prototype specimens (Uang & Latham - 1995) indicates that thisconnection has the ability to achieve very satisfactory levels of plastic rotation without relyingon sensitive CJP welds between the column and girder flanges, or requiring specification ofnotch-tough weld material. The elimination of the through-thickness loading of the flange mayresult in higher levels of connection reliability. Due to the exclusive use of fillet welds, specialinspection requirements for welding and bolting can be reduced significantly with thisconnection.

This connection is proprietary (patent pending) and not in the public domain. It has not beentested in a repair condition. Access to the top of the top flange of the girder might requiredemolition of the existing slab. The cost of the connection may be greater than some of the othermodification methods discussed above; however, this cost differential may not be as great ondouble-sided connections because much of the cost is associated with the side plates which aresimilar for both single-sided and double-sided connections. Publicly bid projects may have todevelop performance specifications to permit other connections to be considered for use unless astrong case for sole-sourcing the connection can be made.

Quantitative Results:No. of specimens tested: 3Girder Size: W36 x 150Column Size: W14 x 426


6-47

Plastic Rotation achieved-3 Specimens : >.042 to 0.06 radian



6-48


Interim Guidelines:Evaluation, Repair, Modification andDesign of Steel Moment Frames Chapter 7 - New Construction

7-1

7. NEW CONSTRUCTION

The building code provisions for earthquake resistive design of Special Moment-Resisting Frames(SMRFs) assume that these structures are extremely ductile and therefore are capable of largeplastic rotations at, or near to, their beam-column connections. Based on limited research, andobservations of damage experienced in the Northridge Earthquake, it appears that conventionallydesigned connection assemblies configured such that plastic deformation concentrates at thebeam-column connection are not capable of reliably withstanding large plastic rotation demands. The reliability appears to decrease as the size of the connected members increases. Other factorsaffecting this reliability appear to include the quality of workmanship, joint detailing, toughness ofthe base and weld metals, relative strengths of the connection elements, and the combined stressespresent on these elements. Unfortunately, the quantitative relationship between these factors andconnection reliability is not well defined at this time.

In order to attain frames that can reliably perform in a ductile manner, these Interim Guidelinesrecommend that SMRF connections be configured with sufficient strength so that plastic hingesoccur within the beam span and away from the face of the column. All elements of the frame, andthe connection itself, should be designed with adequate strength to develop these plastic hinges. The resulting connection assemblies are somewhat complex and the factors limiting their behaviornot always evident. Therefore, qualification of connection designs through prototype testing, orby reference to tests of similar connection configurations is recommended.

These procedures should also be applied to the design of Ordinary Moment-Resisting Frames(OMRFs) located in zones of higher seismicity, or for which highly reliable earthquakeperformance is desired, unless it can be demonstrated that the connections can resist the actualdemands from a design earthquake and remain elastic. Interim Guidelines for determining if adesign meets this condition are provided. Light, single-story, frame structures, the design ofwhich is predominated by wind loads, have performed well in past earthquakes and may continueto be designed using conventional approaches, regardless of the seismic zone they are located in.

Materials and workmanship are critical to frame behavior and careful specification and control ofthese factors is essential. Interim Guidelines for the specification of materials and control ofworkmanship are provided in this Chapter, as well as in Chapters 8, 9, 10 and 11.

7.1 Scope

This Chapter presents interim design guidelines for new welded steel moment frames(WSMFs) intended to resist seismic demands through inelastic behavior. The criteria apply to allSMRF structures designed for earthquake resistance and those OMRF structures located inUniform Building Code (UBC) Seismic Zones 3 and 4 {National Earthquake Hazards ReductionProgram (NEHRP) Map Areas 6 and 7}. Light, single-story buildings, the design of which isgoverned by wind, need not consider these Interim Guidelines. Frames with bolted connections,either fully restrained (type FR) or partially restrained (type PR), are beyond the scope of this

Interim Guidelines:Evaluation, Repair, Modification and

Chapter 7 - New Construction Design of Steel Moment Frames

7-2

document. However, the acceptance criteria for connections may be applied to type FR boltedconnections as well.

Commentary: Observation of damage experienced by WSMF buildings in theNorthridge Earthquake and subsequent laboratory testing of large scale beam-column assemblies has demonstrated that the standard details for WSMFconnections commonly used in the past are not capable of providing reliableservice in the post-elastic range. Therefore, structures which are expected toexperience significant post-elastic demands from design earthquakes, or forwhich highly reliable seismic performance is desired, should be designed usingthe Interim Guidelines presented herein.

In order to determine if a structure will experience significant inelasticbehavior in a design earthquake, it is necessary to perform strength checks of theframe components for the combination of dead and live loads expected to bepresent, together with the full earthquake load. Except for structures with specialperformance goals, or structures located within the near field (within 10kilometers) of known active earthquake faults, the full earthquake load may betaken as the minimum design earthquake load specified in the building code, butcalculated using a lateral force reduction coefficient (Rw or R) of unity. If allcomponents of the structure and its connections have adequate strength to resistthese loads, or nearly so, then the structure may be considered to be able to resistthe design earthquake, elastically.

Design of frames to remain elastic under unreduced (Rw {R} taken as unity)earthquake forces may not be an overly oppressive requirement, particularly inmore moderate seismic zones. Most frame designs are currently controlled bydrift considerations and have substantially more strength than the minimumspecified for design by the building code. As part of the SAC Phase 1 research, anumber of modern frame buildings designed with large lateral force reductioncoefficients (Rw = 12, {R = 8}) were evaluated for unreduced forces calculatedusing the standard building code spectra. It was determined that despite thenominally large lateral force reduction coefficients used in the original design,the maximum computed demands from the dynamic analyses were only on theorder of 2 to 3 times those which would cause yielding of the real structures(Krawinkler, et. al. - 1995; Uang, et. al. - 1995; Engelhardt, et. al. - 1995, Hart,et. al. - 1995; Kariotis and Eimani - 1995). Therefore, it is not unreasonable toexpect that OMRF structures (nominally designed with a lateral force reductioncoefficient Rw = 6 {R = 4.5}) could resist the design earthquakes with near elasticbehavior. Regardless of these considerations, better seismic performance can beexpected by designing structures with greater ductility rather than less andengineers are not encouraged to design structures for elastic behavior usingbrittle or unreliable details..


7-3

For structures designed to meet special performance goals, and buildingslocated within the near field of major active faults, full earthquake loadscalculated in accordance with the above procedure may not be adequate. Forsuch structures, the full earthquake load should be determined using a sitespecific ground motion characterization and a suitable analysis procedure. Recent research (Heaton, et. al. - 1995) suggests that the elastic responsespectrum technique, typically used for determining seismic forces for structuraldesign, may not provide an adequate indication of the true earthquake demandsproduced by the large impulsive ground motions common in the near field oflarge earthquake events. Further, this research indicates that frame structures,subjected to such impulsive ground motions can experience very large drifts, andpotential collapse. Direct nonlinear time history analysis, using an appropriateground motion representation would be one method of more accuratelydetermining the demands on structures located in the near field. Additionalresearch on these effects is required.

As an alternative to use of the criteria contained in these Interim Guidelines,OMRF structures in zones of high seismicity (UBC seismic zones 3 and 4 andNEHRP map areas 6 and 7) may be designed for the connections to remainelastic (Rw or R taken as 1.0) while the beams and columns are designed using thestandard lateral force reduction coefficients specified by the building code. Although this is an acceptable approach, it may result in much larger connectionsthan would be obtained by following these Interim Guidelines.

The use of partially restrained connections may be an attractive andeconomical alternative to the design of frames with fully restrained connections. However, the design of frames with partially restrained connections is beyond thescope of this document. The AISC is currently working on development ofpractical design guidelines for frames with partially restrained connections.

7.2 General - Welded Steel Frame Design Criteria

7.2.1 Criteria

Welded Steel Moment Frame (WSMF) systems should, as a minimum, be designed for theprovisions of the prevailing building code and these Interim Guidelines. Special Moment-Resisting Frames (SMRF)s and Ordinary Moment-Resisting Frames (OMRF)s with FRconnections, should additionally be designed in accordance with the emergency code change tothe 1994 UBC {NEHRP-1994}, restated as follows:

2211.7.1.1. Required Strength {NEHRP-1994 Section 5.2, revision to Ref. 8.2c of Ref. 5.3}

The girder-to-column connections shall be adequate to develop the lesser of the following:


2. The moment corresponding to development of the panel zone shear strength as determined by Formula (11-1).



7-4

2211.7.1.3-2 Connection Strength

Connection configurations utilizing welds and high strength bolts shall demonstrate, by approved cyclic test results orcalculation, the ability to sustain inelastic rotations and to develop the strength criteria in Section 2211.7.1.1considering the effects of steel overstrength and strain hardening.

Commentary: At this time, no recommendations are made to change theminimum lateral forces, drift limitations or strength calculations which determinemember sizing and overall performance of moment frame systems, except asrecommended in Sections 7.2.2, 7.2.3 and 7.2.4. The design of joints andconnections is discussed in Section 7.3. The UBC permits OMRF structures withFR connections, designed for 3/8Rw times the earthquake forces otherwiserequired, to be designed without conforming to Section 2211.7.1. However, thisis not recommended.

7.2.2 Strength



Flexure Ms = Z Fy

Shear Vs = 0.55 Fy d tAxial compression Psc = 1.7 Fa AAxial tension Pst = Fy AConnectors Full Penetration welds Fy A Partial Penetration welds 1.7 allowable (see commentary) Bolts and fillet welds 1.7 allowable

Commentary: Partial penetration welds are not recommended for tensionapplications in critical connections resisting seismic induced stresses. Thegeometry of partial penetration welds creates a notch-like condition that caninitiate brittle fracture under conditions of high tensile strain.

7.2.3 Configuration

Frames should be proportioned so that the required plastic deformation of the frame may beaccommodated through the development of plastic hinges at pre-determined locations within thegirder spans, as indicated in Figure 7-1. Beam-column connections should be designed withsufficient strength (through the use of cover plates, haunches, side plates, etc.) to forcedevelopment of the plastic hinge away from the column face. This condition may also be attainedthrough local weakening of the beam section at the desired location for plastic hinge formation.


7-5

Plastic Hinges


L’

L

h

drift angle - θ

Figure 7-1 - Desired Plastic Frame Behavior

Commentary: Nonlinear deformation of frame structures is accommodatedthrough the development of inelastic flexural or shear strains within discreteregions of the structure. At large inelastic strains these regions can develop intoplastic hinges, which can accommodate significant concentrated rotations atconstant (or nearly constant) load through yielding at tensile fibers and bucklingat compressive fibers. If a sufficient number of plastic hinges develop in a frame,a mechanism is formed and the frame can deform laterally in a plastic manner. This behavior is accompanied by significant energy dissipation, particularly if anumber of members are involved in the plastic behavior, as well as substantiallocal damage to the highly strained elements. The formation of hinges incolumns, as opposed to beams, is undesirable, as this results in the formation ofmechanisms with relatively few elements participating, so called “storymechanisms” and consequently little energy dissipation occurring. In addition,such mechanisms also result in local damage to critical gravity load bearingelements.

The prescriptive connection contained in the UBC and NEHRP RecommendedProvisions prior to the Northridge Earthquake was based on the development ofplastic hinges within the beams at the face of the column, or within the columnpanel zone itself. If the plastic hinge develops in the column panel zone, theresulting column deformation results in very large secondary stresses on the beamflange to column flange joint, a condition which can contribute to brittle failure. If the plastic hinge forms in the beam, at the face of the column, this can result invery large through-thickness strain demands on the column flange material andlarge inelastic strain demands on the weld metal and surrounding heat affectedzones. These conditions can also lead to brittle joint failure. In order to achievemore reliable performance, it is recommended that the connection of the beam tothe column be configured to force the inelastic action (plastic hinge) away fromthe column face. This can be done either by local reinforcement of the



7-6

connection, or locally reducing the cross section of the beam, at a distance awayfrom the connection. Plastic hinges in steel beams have finite length, typically onthe order of half the beam depth. Therefore, the location for the plastic hingeshould be shifted at least that distance away from the face of the column. Whenthis is done through reinforcement of the connection, the flexural demands on thecolumns, for a given beam size, are increased. Care must be taken to assure thatweak column conditions are not inadvertently created by local strengthening ofthe connections.

It should be noted that some professionals and researchers believe thatconfigurations which permit plastic hinging to occur adjacent to the column facemay still provide reliable service under some conditions. These conditions mayinclude limitations on the size of the connected sections, the use of base and weldmetals with adequate notch toughness, joint detailing that minimizes notch effects,and appropriate control of the relative strength of the beam and columnmaterials. Sufficient research has not been performed to date either to confirmthese suggestions or define the conditions in which they are valid. Researchhowever does indicate that reliable performance can be attained if the plastichinge is shifted away from the column face, as suggested above. Consequently,these Interim Guidelines make a general recommendation that this approach betaken. Additional research should be performed to determine the acceptability ofother approaches.

It should also be noted that reinforced connection (or reduced beam section)configurations of the type described above, while believed to be effective inpreventing brittle connection fractures, will not prevent structural damage fromoccurring. Brittle connection fractures are undesirable because they result in asubstantial reduction in the lateral-force-resisting strength of the structure which,in extreme cases, can result in instability and collapse. Connections configuredas described in these Interim Guidelines should experience many fewer suchbrittle fractures than unmodified connections. However, the formation of aplastic hinge within the span of a beam is not a completely benign event. Beamswhich have formed such hinges may exhibit large buckling and yieldingdeformation, damage which typically must be repaired. The cost of such repairscould be comparable to the costs incurred in repairing fracture damageexperienced in the Northridge Earthquake. The primary difference is that lifesafety protection will be significantly enhanced and most structures that haveexperienced such plastic deformation damage should continue to be safe foroccupancy, while repairs are made.

If the types of damage described above are unacceptable for a given building,then alternative structural systems should be considered, that will reduce theplastic deformation demands on the structure during a strong earthquake. Appropriate methods of achieving such goals include the installation ofsupplemental braced frames, energy dissipation systems, base isolation systems


7-7

and similar structural systems. Framing systems incorporating partiallyrestrained connections may also be quite effective in resisting large earthquakeinduced deformation with limited damage.


The plastic rotation capacity of connection assemblies should reflect realistic estimates of thetotal (elastic and plastic) drift likely to be induced in the frame by earthquake ground shaking, andthe geometric configuration of the frame. For frames of typical configuration, and for groundshaking of the levels anticipated by the building code, a minimum plastic rotation capacity of 0.03radian is recommended.

When the configuration of a frame is such that the ratio L/L’ is greater than 1.25, the plasticrotation demand should be taken as follows:

( )( )θ = + −0.025 1 L L' L' (7-1)

where: L is the center to center spacing of columns, andL’ is the center to center spacing of plastic hinges in the bay under consideration

The indicated rotation demands may be reduced when positive means, such as the use of baseisolation or energy dissipation devices, are introduced into the design, to control the building’sresponse. When such measurers are taken, nonlinear dynamic analyses should be performed andthe connection demands taken as 0.005 radians greater than the rotations calculated in theanalyses. The nonlinear analyses should conform to the criteria specified in UBC-94 Section 1655{NEHRP-94 Section 2.6.4.2} for nonlinear dynamic analysis of base isolated structures. Groundmotion time histories utilized for these nonlinear analyses should satisfy the scaling requirementsof UBC-94 Section 1655.4.2 {NEHRP-94 Section 2.6.4.4}, except that if the building is not baseisolated, the structure period T, calculated in accordance with UBC-94 Section 1628 {NEHRP-94Section 2.3.3.1} should be substituted for TI.

Commentary: Traditionally, engineers have calculated demand in momentframes by sizing the members for strength and drift using code forces (eitherequivalent static or reduced dynamic forces) and then "developing the strength ofthe members." Since 1988, "developing the strength" has been accomplished byprescriptive means based on a review of testing of moment frame connections tothat date. It was assumed that the prescribed connections would be strongenough that the beam or girder would yield (in bending), or the panel zone wouldyield (in shear) in a nearly perfectly plastic manner producing the plasticrotations necessary to dissipate the energy of the earthquake.

A realistic estimate of the interstory drift demand for most structures and mostearthquakes is on the order of 0.015 to 0.025 times the story height for WSMFstructures designed to code allowable drift limits. In such frames, a portion ofthe drift will be due to elastic deformations of the frame, while the balance must



7-8

be provided by inelastic rotations of the beam plastic hinges, by yielding of thecolumn panel zone, or by a combination of the two.

In the 1994 Northridge Earthquake, many moment-frame connectionsfractured with little evidence of plastic hinging of the beams or yielding of thecolumn panel zones. Testing of moment frame connections both prior to andsubsequent to the earthquake suggests that the standard, pre-Northridge, weldedflange-bolted web connection is unable to reliably provide plastic rotationsbeyond about 0.005 radian for all ranges of beam depths and often fails belowthat level. Since the elastic contribution to drift may approach 0.01 radian, thenecessary inelastic contributions will exceed the capability of the standardconnection in many cases. For frames designed for code forces and for the codedrift, the necessary plastic rotational demand may be expected to be on the orderof 0.02 radian or more and new connection configurations should be developed toaccommodate such rotation without brittle fracture.

The recommended connection demand of 0.03 radians was selected both toprovide a comfortable margin against the demands actually expected in mostcases and because in recent testing of connection assemblies, specimens capableof achieving this demand behaved in a ductile manner through the formation ofplastic hinges.

For a given building design, and known earthquake hazard, it is possible tomore accurately estimate plastic rotation demands on frame connections. Thisrequires the use of nonlinear analysis techniques. Analysis software, capable ofperforming such analyses is becoming more available and many design officeswill have the ability to perform such analyses and develop more accurateestimates of inelastic demands for specific building designs. However, whenperforming such analyses, care should be taken to evaluate building response formultiple earthquake time histories, representative of realistic ground motions forsites having similar geologic characteristics and proximity to faults, as the actualbuilding site. Relatively minor differences in the ground motion time history usedas input in such an analysis can significantly alter the results. Since there issignificant uncertainty involved in any ground motion estimate, it isrecommended that analysis not be used to justify the design of structures withnon-ductile connections, unless positive measures such as the use of baseisolation or energy dissipation devices are taken, to provide reliable behavior ofthe structure.

It has been pointed out that it is not only the total plastic rotation demand thatis important to connection and frame performance, but also the connectionmechanism (for example - panel zone yielding, girder flange yielding/buckling,etc.) and hysteretic loading history. These are matters for further study in thecontinuing research on connection and joint performance.


7-9

7.2.5 Redundancy

The frame system should be designed and arranged to incorporate as many moment-resistingconnections as is reasonable into the moment frame.

Commentary: Early moment frame designs were highly redundant and nearlyevery column was designed to participate in the lateral-force-resisting system. Inan attempt to produce economical designs, recent practice often produced designswhich utilized only a few large columns and beams in a small proportion of thebuilding’s frames for lateral resistance, with the balance of the building columnsdesigned not to participate in lateral resistance. This practice led to the need forlarge welds at the connections and to reliance on only a few connections for thelateral stability of the building. The resulting large framing elements andconnections are believed to have exacerbated the poor performance of the pre-Northridge connection. Further, if only a few framing elements are available toresist lateral demands, then failure of only a few connections has the potential toresult in a significant loss of earthquake resisting strength. Together, theseeffects are not beneficial to building performance.

The importance of redundancy to building performance can not be over-emphasized. Even connections designed and constructed according to theimproved procedures recommended by these Interim Guidelines will have somepotential, albeit greatly reduced, for brittle failures. As the number of individualbeams and columns incorporated into the lateral-force-resisting system isincreased, the consequences of isolated connection failures significantly reduces.Further, as more framing elements are activated in the building’s response toearthquake ground motion, the building develops greater potential for energyabsorption and dissipation, and ability to control earthquake induceddeformations to acceptable levels.

Incorporation of more of the building framing into the lateral-force-resistingsystem will lead to smaller members and therefore an anticipated increase in the reliability of individual connections. It will almost certainly lead to improvedoverall system reliability. Further, recent studies conducted by designers indicatethat under some conditions, redundant framing systems can be constructed aseconomically as non-redundant systems. In these studies, the additional costsincurred in making a greater number of field-welded moment-resistingconnections in the more redundant frame were balanced by a reduced totaltonnage of steel in the lateral-force-resisting systems and sometimes, reducedfoundation costs as well.

The 1994 UBC requirements limit the relative number of weak column/strongbeam connections in the moment frame system. There is a divergence of opinionamong structural engineers on the desirability of frames in which all beam-column connections are made moment-resisting, including those of beams



7-10

framing to the minor axis of columns. Use of such systems as a means ofsatisfying these Interim Guidelines requires careful consideration by thestructural engineer. Limited testing in the past has indicated that momentconnections made to the minor axis of wide flange columns are subject to thesame types of fracture damage experienced by major axis connections. As of thistime, there has not been sufficient research to suggest methods of making reliableconnections to the column minor axis.

7.2.6 System Performance

WSMF design should consider all effects of connection modifications on the response andperformance of the frame.

Commentary: Methods developed thus far for improving performance ofbeam/column connections involve shifting of the hinge point away from thecolumn face either by reinforcement of the connection (e.g. haunches, coverplates, etc.), or reducing the relative strength of the beam locally. Thesemodifications affect the overall stiffness of the frame and, therefore, its seismicresponse. In fact, it can be shown that the use of smaller beam sizes andhaunched connections will result in the same overall frame stiffness as the use oflarger beams and unstiffened connections. Additionally, haunching orreinforcement results in magnified moments and shears at the column face whichshould be included in the strong column/weak beam calculations, panel zone andweb connection calculations, and column axial demand calculations. Unsymmetrical haunches, placed on only the bottom (or top) of the beam can alsochange the relative stiffness of columns above and below the beam resulting inunexpected formation of plastic hinges in one of the columns. In addition, ifplastic hinges are forced out into the beam span, away from the column face, thelocal lateral stability of beams at plastic hinges away from the column should beconsidered.

7.2.7 Special Systems

When WSMFs are used as components of "Tube" type buildings with beams yielding in shearrather than bending, or in Dual System structures, appropriate consideration should be given tothe differences in plastic rotation demands expected (as compared to pure moment frame designs)when applying these provisions. (See discussion in Section 7.10.)

Commentary: Moment frames which are employed in dual systems in low-risebuildings or in the lower levels of taller buildings may have significantly lowerrotation demands than those in pure frame buildings. Engineers may consider itappropriate to use less conservative connection designs or qualificationrequirements for such frames, or for portions of such frames. Appropriateanalytical substantiation should be provided for any alternative criteria utilized.


7-11

For tube frames with shear-yielding beams, qualification by testing isrecommended, but designs and requirements may differ from those presented inthese Interim Guidelines. Again appropriate analytical substantiation should beprovided for the selected criteria.

7.3 Connection Design & Qualification Procedures - General

7.3.1 Connection Performance Intent

The intent of connection design should be to force the plastic hinge away from the face of thecolumn to a pre-determined location within the beam span. This may be accomplished by localreinforcement of the connection itself (cover plates, haunches, side plates, etc.) or by localreductions of the beam section (drilled holes, trimmed flanges, etc.). All elements of theconnection should have adequate strength to develop the forces resulting from the formation ofthe plastic hinge at the predetermined location, together with forces resulting from gravity loads.

7.3.2 Qualification by Testing

Connection strength and plastic rotation capacity should be demonstrated by approved cyclictesting as described in Section 7.4, except as indicated in paragraph 7.3.3. It is recommended thatpreliminary design of specimens to be tested be developed using the Interim Guidelines of Section7.5. Extrapolation and interpolation of test results using the calculation procedures of Section 7.5is acceptable for connections of elements having similar geometries and material specifications astested connections.

Commentary: Cyclic testing of connections matching the essential features ofthose to be used in the actual design is the most reliable method of assuring thatthe expected connection performance can be attained. Section 7.4 describestesting guidelines in detail. Guidelines for extrapolation by calculation are givenin Section 7.5.

7.3.3 Design by Calculation

Connection design by calculations alone may be acceptable under the following conditions:

a) Calculations are based on comparison with previously tested assemblies, or withprototype connections tested for the project;

b) Conditions of the calculated detail, including member property relationships,material properties, welding materials, processes and procedures, and constructionsequence, mirror those of the tested detail as closely as possible; or

c) Qualified third party review, in accordance with Section 4.5 is performed.



7-12

Commentary: Use of calculations based on engineering principles alone, or toextrapolate data from tests performed on assemblies which do not preciselymirror the conditions of the calculated assembly, requires caution and judgment. Subjective factors affecting the acceptability of such an approach should include:

a) The importance of the structure: Greater caution in applying a calculation-onlyapproach should be exercised for more important facilities, particularly when thefacility is expected to remain functional after a major earthquake.

b) Confidence in the lateral forces used in design: Projects which carefully applyseismic forces based on extensively researched, site-specific seismic hazardstudies, and having resulting designs with low calculated rotation demands, maywarrant more confidence in the application of connection designs usingcalculations only, as opposed to those that do not use this type of information. Most structures are designed to satisfy the minimum code seismic forces. Structures that are designed assuming higher levels of seismic demand (bothstrength and stiffness) than found in typical projects, could also possibly bedemonstrated to warrant greater latitude in applying a calculation-onlyapproach.

c) The degree of redundancy, regularity and potential over-strength in the structure: Greater care in applying a calculation-only approach should be considered instructures with a limited number of lateral-force-resisting elements in eachdirection or those with unusual building geometries. Structures with a highdegree of redundancy may be demonstrated to be better able to tolerate limitedinstances of marginal connection performance. Frames designed to limit therotational demand by relying on elastic or near-elastic behavior may also bemore amenable to a calculation-only approach than those that depend on highlevels of plastic rotation to dissipate anticipated seismic demands. However, ithas not been shown that superior seismic performance results when strength issubstituted for ductility, and overly strong, frames with non ductile connectionsare not the intent of these guidelines.

d) Proximity to active faults: Ground motion records from recent earthquakesclearly demonstrate that sites located close to a fault rupture experiencesubstantially more severe ground motion than is explicitly provided for in currentcode design provisions. When a building is located within 5 km of an active fault,the plastic rotation demands on connections may exceed those provided for inthese Guidelines, and additional caution in design procedures is warranted.

For structures that are essential, contain hazardous materials, are designedwith a low degree of conservatism or redundancy, connections qualification bytest (either through reference to tests from other projects or project-specific


7-13

testing of connections) is strongly recommended. This recommendation should beconsidered until such time as SAC or other research develops sufficient data toallow formulation of analytical design guidelines for general application.

For non-essential structures designed with a reasonable degree of redundancyor overstrength and incorporating enhanced welding requirements and qualitycontrol, calculations as described above, using proportioning and stress levelscompatible with previously completed test programs, may provide sufficientassurance of reliability.

7.4 Guidelines for Connection Qualification by Testing

7.4.1 Testing Protocol

Unless future testing programs reveal significant effects of dynamic loading rate or timehistory loading, and unless the effects of other factors (e.g., restraint conditions and compositeslab effects) are found to be compelling, a testing protocol similar to ATC-24, Guidelines forCyclic Seismic Testing of Components of Steel Structures (Applied Technology Council - 1992),is recommended as the basis for qualification tests.

The testing program should replicate as closely as practical the anticipated conditions in thefield, including such factors as:

a) Member sizes.

b) Material specifications.

c) Welding process, details and construction conditions.

d) Cover plates, continuity plates, web tabs, bolts, and doubler plates.

e) Connection configuration (e.g., beams on both sides).

f) Induced stresses because of restraint conditions on the welds and connection members.

g) Axial load, where pertinent.

h) Gravity load, where significant.

The testing program should be organized to provide as much information as possible aboutthe capability of the connections selected. The following program is recommended:

a) Test at least two full size specimens representative of the larger beam/columnassemblies in the project.



7-14

b) Test one additional full size specimen representative of other beam/column assemblieswith significantly different interaction properties, such as beam b/t, panel zonestress/distortion, etc.

If any of the specimens fails to meet the qualification criteria, the connection should beredesigned and retested.

Where two-sided connections are used in the structure, and the type of connection being usedcan be expected to perform differently in a two-sided use than in one-sided use, it should be testedin the two-sided configuration as well as the one-sided. Two-sided connection assemblies can beexpected to behave differently than one-sided assemblies, for example, when panel zonedistortions will be significantly different, or when systems involve transfer of stress to the columnby plates, welds, or other elements which are connected to the beams on both sides of the column.

The inclusion of axial load should be considered when analysis indicates that significanttension can be expected to occur in a significant number of the columns represented by thespecimen and where the connection type relies on the through-thickness strength of the columnflanges. If the presence of a floor slab is anticipated to have significant influence on either thelocation or mechanism of the plastic hinge formed, than this should also be included in the testspecimen.

7.4.2 Acceptance Criteria


a) The connection should develop beam plastic rotations as indicated in Section 7.2.4, forat least one complete cycle.

b) The connection should develop a minimum strength equal to the plastic strength of thegirder, calculated using minimum specified yield strength Fy, throughout the loadinghistory required to achieve the required plastic rotation capacity, as indicated in a),above. If the load limiting mechanism in the test is buckling of the girder flanges, theengineer, upon consideration of the effect of strength degradation on the structure,may consider a minimum of 80% of the nominal strength as acceptable.

Commentary: While the testing of all connection geometries and membercombinations in any given building might be desirable, it would not be verypractical nor necessary. Test specimens should replicate, within the limitationsassociated with test specimen simplification, the fabrication and weldingprocedures, connection geometry and member size, and potential modes offailure. If the testing is done in a manner consistent with other testing programs,reasonable comparisons can be made. On the other hand, testing is expensiveand it is difficult to realistically test the beam-column connection using actualboundary conditions and earthquake loading histories and rates.


7-15

It was suggested in Interim Recommendation No. 2 by the SEAOC SeismologyCommittee that three tested specimens be the minimum for qualification of aconnection. Further consideration has led to the recognition that while three testsmay be desirable, the actual testing program selected should consider theconditions of the project. Since the purpose of the testing program is to "qualifythe connection", and since it is not practical for a given project to do enough teststo be statistically meaningful considering random factors such as material,welder skills, and other variables, arguments can be made for fewer tests ofidentical specimens, and concentration on testing specimens which represent therange of different properties which may occur in the project. Once a connectionis qualified, that is, once it has been confirmed that the connection can work,monitoring of actual materials and quality control to assure emulation of thetested design becomes most important.

Because of the cost of testing, use of calculations for interpolation orextrapolation of test results is desirable. How much extrapolation should beaccepted is a difficult decision. As additional testing is done, more informationmay be available on what constitutes "conservative" testing conditions, therebyallowing easier decisions relative to extrapolating tests to actual conditions whichare likely to be less demanding than the tests. For example, it is hypothesizedthat connections of shallower, thinner flanged members are likely to be morereliable than similar connections consisting of deeper, thicker flanged members. Thus, it may be possible to test the largest assemblages of similar details andextrapolate to the smaller member sizes - at least within comparable membergroup families. Extrapolation or interpolation of results with differences inwelding procedures, details or material properties is more difficult.

7.5 Guidelines for Connection Design by Calculation

In conditions where it has been determined that design of connections by calculation issufficient, or when calculations are used for interpolation or extrapolation, the followingguidelines should be used.

7.5.1 Material Strength Properties

In the absence of project specific material property information, the values listed in Table 7-1should be used to determine the strength of steel shape and plate for purposes of calculation. Thepermissible strength for weld metal should be taken in accordance with the building code. Additional information on material properties may be found in the Interim Guidelines of Chapter 8.



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Table 7-1 - Properties for Use in Connection Design

Material Fy (ksi) Fy m (ksi) Fu (ksi)A36 36 use values for

Dual Certified58

Dual Certified Beam Axial, Flexural3

Shape Group 1 Shape Group 2 Shape Group 3 Shape Group 4 Through-Thickness

50

-

551

581

571

541

-

65 min.

Note 2A572 Column/Beam

Axial, Flexural3

Shape Group 1 Shape Group 2 Shape Group 3 Shape Group 4 Shape Group 5 Through-Thickness

50

-

581

581

571

571

551

-

65 min.

Note 2,

A913-50 Axial, Flexural Through-thickness

50-

581

-65 min.Note 2,

A913--65 Axial, Flexural 65 751 80 min.Notes:1. Based on coupons from web. For thick flanges,

the Fy flange is approximately 0.95 Fy web.2. See Commentary3. Values based on (SSPC-1994)

Commentary: The causes for through-thickness failures of column flanges (typesC2, C4, and C5), observed both in buildings damaged by the NorthridgeEarthquake and in some test specimens, are not well understood. They arethought to be a function of the metallurgy and “purity” of the steel; conditions ofloading including the presence of axial load and rate of loading application;conditions of tri-axial restraint; conditions of local hardening and embrittlementwithin the weld’s heat affected zone; and by the relationship of the connectioncomponents as they may affect flange bending stresses and flange curvatureinduced by panel zone yielding. Given the many complex factors which can affectthe through-thickness strength of the column flange, determination of a reliablebasis upon which to set permissible design stresses will require significantresearch.

Interim Recommendation No. 2 (SEAOC-1995) included a value of 40 ksi,applied to the projected area of beam flange attachment, for the through-thickness strength to be used in calculations. This value was selected because itwas consistent with the successful tests of assemblies with cover plates conducted


7-17

at the University of Texas at Austin (Engelhardt and Sabol - 1994). However,because of the probable influence of all the factors noted above, this value canonly be considered to reflect the specific conditions of those tests and specimens.

Although reduced stresses at the column face produced acceptable results inthe University of Texas tests, the key to that success was more likely the result offorcing the plastic hinge away from the column than reduction of the through-thickness stress by the cover plates. Reduction of through-thickness columnflange stress to ever lower levels by the use of thicker cover plates is notrecommended, since such cover plates will result in ever higher forces on the faceof the column flange.

Notwithstanding all of the above, successful tests using cover plates and othermeasures of moving hinges (and coincidentally reducing through-thickness stress)continue to be performed. In the interim, engineers choosing to utilizeconnections relying on through-thickness strength should recognize that despitethe successful testing, connections relying on through-thickness strength can notbe considered to be fully reliable until the influence of the other parametersdiscussed above can be fully understood. A high amount of structuralredundancy is recommended for frames employing connections which rely onthrough-thickness strength of the column flange.

7.5.2 Design Procedure

Select a connection configuration, such as one of those indicated in Section 7.9, that willpermit the formation of a plastic hinge within the beam span, away from the face of the column,when the frame is subjected to gravity and lateral loads. The following procedure should befollowed to size the various elements of the connection assembly:

7.5.2.1 Determine Plastic Hinge Locations

For beams with gravity loads representing a small portion of the total flexural demand, theplastic hinge may be assumed to occur at a distance equal to 1/3 of the beam depth from the edgeof the reinforced connection (or start of the reduced beam section), unless specific test data forthe connection indicates that a different value is appropriate. Refer to Figure 7-2.



7-18

L

Bea

m d

epth

- d

Edge

of r

einf

orce

dco

nnec

tion

Edge

of r

einf

orce

dco

nnec

tion

d/3

L’

Plastichinge

Figure 7-2 - Location of Plastic Hinge

Commentary: The suggested location for the plastic hinge, at a distance d/3away from the end of the reinforced section (or beginning of reduced section) isbased on the observed behavior of test specimens, with no significant gravity loadpresent. If significant gravity load is present, this can shift the locations of theplastic hinges, and in the extreme case, even change the form of the collapsemechanism. If flexural demand on the girder due to gravity load is less thanabout 30% of the girder plastic capacity, this effect can safely be neglected, andthe plastic hinge locations taken as indicated. If gravity demands significantlyexceed this level then plastic analysis of the girder should be performed todetermine the appropriate hinge locations. In zones of high seismicity (UBCZones 3 and 4, and NEHRP Map Areas 6 and 7) gravity loading on the girders ofearthquake resisting frames typically has a very small effect.

7.5.2.2 Determine Probable Plastic Moment at Hinges

Determine the probable value of the plastic moment, Mpr, at the location of the plastic hingesas:

M M Z Fpr p b y= =β β (7-2)

where: ß is a coefficient that adjusts the nominal plastic moment to the estimated hingemoment based on the mean yield stress of the beam material and the estimatedstrain hardening. When designs are based upon calculations alone, an additionalfactor is recommended to account for uncertainty. In the absence of adequatetesting of the type described above, ß should be taken as 1.4 for ASTM A572 andfor A913, Grades 50 and 65 steels. Where adequate testing has been performed ßshould be permitted to be taken as 1.2 for these materials.


7-19


Commentary: In order to compute β, the expected yield strength, strainhardening and an appropriate uncertainty factor need to be determined. Thefollowing assumed strengths are recommended:

Expected Yield: The expected yield strength, for purposes of computing (Mpr) may be taken as:

Fye = 0.95 Fym (7-3)

The 0.95 factor is used to adjust the yield stress in the beam web, where couponsfor mill certification tests are normally extracted, to the value in the beam flange.Beam flanges, being comprised of thicker material, typically have somewhatlower yield strengths than do beam web material.

Fy m for various steels are as shown in Table 7-1, based on a survey of webcoupon tensile tests (Steel Shape Producers Council - 1994). The engineer iscautioned that there is no upper limit on the yield point for ASTM A36 steel andconsequently, dual-certification steel having properties consistent with ASTMA572, Grade 50 is routinely supplied when ASTM A36 is specified. Consequently, it is the recommendation here that the design of connections bebased on an assumption of Grade 50 properties, even when A36 steel is specifiedfor beams. It should be noted that at least one producer offers A36 steel with amaximum yield point of 50 ksi in shape sizes ranging up to W 24x62.

Strain Hardening: A factor of 1.1 is recommended for use with the mean yieldstress in the foregoing table when calculating the probable plastic momentcapacity Mpr.. The 1.1 factor for strain hardening, or other sources of strengthabove yield, agrees fairly well with available test results. The 1.1 factor couldunderestimate the over-strength where significant flange buckling does not act asa gradual limit on the beam strength. Nevertheless, the 1.1 factor seems areasonable expectation of over-strength considering the complexities involved.

Modeling Uncertainty: Where a design is based on approved cyclic testing, themodeling uncertainty may be taken as 1.0, otherwise the recommended value is1.2.

In summary, for Grade 50 steel, we have:

β = [0.95 (54 ksi to 58 ksi)/50 ksi] (1.1) 1.2) = 1.35 t0 1.45, say 1.4

7.5.2.3 Determine Shear at the Plastic Hinge

The shear at the plastic hinge should be determined by statics, considering gravity loads actingon the beam. A free body diagram of that portion of the beam between plastic hinges, is a useful



7-20

tool for obtaining the shear at each plastic hinge. Figure 7-3 provides an example of such acalculation. For the purposes of such calculations, gravity load should be based on the loadcombinations required by the building code in use.

L

d/3L’

Plastichinge P

L/2

P

Mpr MprL’Vp


“A”

VAw


Figure 7-3 - Sample Calculation of Shear at Plastic Hinge

Commentary: The UBC gives no specific guidance on the load combinations touse with strength level calculations while the NEHRP Recommended Provisionsdo specify specific load factors for the various dead, live and earthquakecomponents of load. For designs performed in accordance with the UBC it iscustomary to use unfactored gravity loads when checking the strength ofelements.

7.5.2.4 Determine Strength Demands at Each Critical Section

In order to complete the design of the connection, including sizing the various plates andjoining welds which make up the connection, it is necessary to determine the shear and flexuralstrength demands at each critical section. These demands may be calculated by taking a free bodyof that portion of the connection assembly located between the critical section and the plastichinge. Figure 7-4 demonstrates this procedure for two critical sections, for the beam shown inFigure 7-3.


7-21

Plastichinge

Vp

Mpr

Plastichinge

Vp

Mpr

x

Mf

x+dc/2

dc

Mf=Mpr +Vpx

Mc

Mc=Mpr +Vp(x+dc/2)


Figure 7-4 - Calculation of Demands at Critical Sections

Commentary: Each unique connection configuration may have different criticalsections. The vertical plane that passes through the joint between the beamflanges and column (if such joining occurs) will typically define at least one suchcritical section, used for designing the joint of the beam flanges to the column, aswell as evaluating shear demands on the column panel zone. A second criticalsection occurs at the center line of the column. Moments calculated at this pointare used to check strong column - weak beam conditions. Other critical sectionsshould be selected as appropriate.

7.5.2.5 Check for Strong Column - Weak Beam Condition

When required by the building code, the connection assembly should be checked to determineif strong column - weak beam conditions are satisfied. In lieu of UBC-94 equation 11-3.1{NEHRP-91 equation 10-3}, the following equation should be used:

Z (F f ) M 1.0c yc a c− >∑ ∑ (7-4)

where: Zc is the plastic modulus of the column section above and below the connectionFyc is the minimum specified yield stress for the column above and belowfa is the axial load in the column above and belowMc is the moment calculated at the center of the column in accordance with Section 7.5.2.4

Commentary: The building code provisions for evaluating strong column - weakbeam conditions presume that the flexural stiffness of the columns above andbelow the beam are approximately equal. If non-symmetrical connectionconfigurations are used, such as a haunch on the bottom side of the beam, thiscan result in an uneven distribution of stiffness between the two column segments.



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7.5.2.6 Check Column Panel Zone

The adequacy of the shear strength of the column panel zone should be checked. For thispurpose, the term 0.8ΣMf should be substituted for the term 0.8ΣMs in UBC-94 Section2211.7.2.1 {0.9ΣφbMp in NEHRP-91 Section 10.10.3.1}, repeated below for convenience ofreference. Mf is the calculated moment at the face of the column, when the beam mechanismforms, calculated as indicated in Section 7.5.2.4 above.

2211.7.2.1 Strength. The panel zone of the joint shall be capable of resisting the shear inducedby beam bending moments due to gravity loads plus 1.85 times the prescribed seismic forces, butthe shear strength need not exceed that required to develop 0.8ΣMs of the girders framing intothe column flanges at the joint. The joint panel zone shear strength may be obtained from thefollowing formula:

V 0.55F d t3b td d ty c

c c f2

b c

= +

1 (11-1)

where: bc = width of column flangedb = the depth of the beam (including any haunches or cover plates)dc = the depth of the columnt = the total thickness of the panel zone including doubler platestcf = the thickness of the column flange

7.6 Metallurgy and Welding

For Guidelines on Metallurgy and Welding for New Structures, see Chapter 8 of these InterimGuidelines. The recommendation for welding electrodes capable of depositing weld metal withspecified notch toughness, as described therein, should apply to the critical beam flange to columnflange field welded joints. It need not apply to shop welds of continuity plates, etc.

Commentary: This is an area of continuing controversy in the community,requiring additional research for resolution. Some professionals and researchersknowledgeable in fracture mechanics believe it is essential that all weld metal inthe beam column connection, including both field and shop welds, welds ofcontinuity plates, doubler plates, etc., as well as the welds of beam to columnflanges, should have minimum specified notch toughness. Some of these sameprofessionals believe that the notch toughness requirement should apply to thecombined metal, consisting of deposited electrode metal and fused base metal. The current recommendations, which are less restrictive than this position, arebased on recommendations of members of the AWS D1.1 committee. Theserecommendations are consistent with the observation of damage in the NorthridgeEarthquake, in which most fractures initiated at the root of the beam flange tocolumn flange weld. It is of course possible that if notch tough material is used atthis joint and not at others, fractures will initiate in future events at the nextcritical section, which may be a welded joint using material with low toughness atcontinuity plates, or other locations.


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While there is a lack of agreement as to the extent to which notch toughnessspecifications should apply to welded joints in the moment connection, there isgeneral agreement that previously acceptable electrodes that had no reportednotch toughness values should no longer be used for the critical beam flange tocolumn flange field welded joints. Most of the electrodes that are currentlycommercially available and have specified notch toughness requirements willmeet the notch toughness recommendations contained in Chapter 8 of theseInterim Guidelines. Additional research may indicate that alternate criteria areappropriate.

A similar level of disagreement exists with regard to the need for specifyingnotch toughness in base metals. Most of the fractures which have beeninvestigated have initiated in the weld metal rather than in the base metal. Oncethese fractures extend into the base metal, they have already reached significantsize and material toughness alone may not be able to arrest them. Additionalresearch into the benefits of tough material, both in welds and base metals isclearly called for.

7.7 Quality Control/Quality Assurance

Refer to Chapters 9, 10 and 11 of these Interim Guidelines.

7.8 Guidelines on Other Connection Design Issues

The emphasis thus far in testing of connection assemblages has been on the beamflange/column flange joint. The other components of the connection such as panel zones, webconnections and continuity plates have not been studied significantly as independent parameters inthe available testing programs to date. It is assumed that the variation of these components willhave effects on the performance of the connection and thus on the flange joints, and that an as yetundetermined balance of the sizes and details of these significant components will result in theoptimum performance of a particular connection and its various joints. Interim Guidelines forthese other critical portions of the connection assembly are presented below.

7.8.1 Design of Panel Zones

No current recommendations are made to supplement or modify the UBC-1994 {NEHRP-91}provisions for the design of panel zones, other than as indicated in Section 7.5.2.6, above. Panelzone demands should be calculated in accordance with Section 7.5.2.6. As with other elements ofthe connection, available panel zone strength should be computed using minimum specified yieldstress for the material, except when the panel zone strength is used as a limit on the requiredconnection strength, in which case Fym should be used.

Where connection design for two-sided connection assemblies is relying on test data for one-sided connection assemblies, consideration should be given to maintaining the level of panel zone



7-24

deformation in the design to a level consistent with that of the test, or at least assume that thepanel zone must remain elastic, under the maximum expected shear demands.

Commentary: At present, no changes are recommended to the code requirementsgoverning the design of panel zones, other than in the calculation of the demand. There is evidence that panel zone yielding may contribute to the plastic rotationcapability of a connection. However, there is also concern and some evidencethat if the deformation is excessive, a kink will develop in the column flange at thejoint with the beam flange and, if the local curvature induced in the beam andcolumn flanges is significant, can contribute to failure of the joint. This wouldsuggest that greater conservatism in column panel zone design may be warranted.

In addition to the influence of the deformation of the panel zone on theconnection performance, it should be recognized that the use of doubler platesand especially the welding associated with them is likely to be detrimental to theconnection performance. It is recommended that the Engineer consider use ofcolumn sizes which will not require addition of doubler plates, where practical.

7.8.2 Design of Web Connections to Column Flanges

Specific modifications to the code requirements for design of shear connections are not madeat this time. It should be noted that the emergency code change to the UBC-94 {NEHRP-94}deleted the former requirements for supplemental web welds on shear connections. This is felt tobe appropriate since these welds can apparently contribute to the potential for shear tab failure atlarge induced rotations.

When designing shear connections for moment-resisting assemblies, the designer shouldcalculate shear demands on the web connection in accordance with Section 7.5.2.4, above.

Commentary: Some engineers consider that it is desirable to develop as muchbending strength in the web as possible. Additionally, it has been observed insome laboratory testing that pre-mature slip of the bolted web connection canresult in large secondary flexural stresses in the beam flanges and the weldedjoints to the column flange. However, there is some evidence to suggest that ifflange connections should fail, welding of shear tabs to the beam web maypromote tearing of the tab weld to the column flange or the tab itself through thebolt holes, and some have suggested that welding be avoided and that webconnections should incorporate horizontally slotted holes to limit the momentwhich can be developed in the shear tab, thereby protecting its ability to resistgravity loads on the beam in the event of flexural connection failure.

7.8.3 Design of Continuity Plates

Contrary to current code requirements, it is recommended that continuity plates be providedin all cases and that the thickness be at least equal to the thickness of the beam flange (not


7-25

including cover plates) or one half the total effective flange thickness (flange plus cover plate). Welds of the continuity plate to the column should develop the strength of the continuity plate.

Where two-sided connection assemblies are designed based on one-sided connection assemblytest data, consideration should be given to the effect of the greater distortion of the continuityplates expected in the two sided case.

For reinforced connections using vertical ribs or other configurations of reinforcement,continuity plate sizing should be based on engineering principles and consideration of stresspatterns which may occur due to column flange distortion.

For connections incorporating haunches, continuity plates should be provided opposite thejoint of the haunch flange with the column flange.

Commentary: The determination of continuity plate thickness requires, inaddition to code conformance requirements, engineering judgment based onrecognition of two competing factors:

a) Overly thick continuity plates and their welding will contribute to restraint andconsequent residual stresses in the column, as well as to the other usualdetrimental effects of large welds. Conditions of high restraint tend to beconducive to the initiation of fracture.

b) Omission of continuity plates or the use of overly thin continuity plates willpermit column flange distortions which will, in turn, lead to higher stressconcentrations in the beam flange joint opposite the column web.

Testing to date has not firmly established an appropriate design criteria forcontinuity plates, or even that these are definitely needed to obtain goodconnection performance in all cases. However, tests of specimens reinforced withcover plates to date, have been most successful when continuity plates werepresent (Engelhardt & Sabol - 1994). Tests using otherwise similar designs butwith different continuity plate thicknesses have not been performed. This is anarea where further research would be beneficial.

7.8.4 Design of Weak Column and Weak Way Connections

The code permits the use of strong beam/weak column designs under certain circumstances. There is some question as to what should be required for the connections at such conditions. While testing has demonstrated little capability of the pre-Northridge prescriptive connection todevelop significant beam yielding without failure, it should be recognized that if the beam isstronger than the column, considering conservative estimates of the column strength includingstrain hardening, then the beam and its connection can be expected to remain below even this lowfailure threshold, and it would appear to be unnecessary to provide strengthened connections.



7-26

When beam connections are made to the web of columns (weak way) which are stronger thanthe beams, then connection design should be treated similarly to that of strong directionconnections with additional consideration for the unique features of weak direction connections(see Tsai and Popov - 1988). Note that the question of column flange through-thickness strengthis not a consideration for this type of connection, but that development of the strength of a coverplated flange through welds in shear to the inside face of the column flange may be difficult. Unless the members so connected represent a very small part of lateral resistance of the structure,testing of such connections should be considered as mandatory. Extrapolation of results fromstrong way connection testing should not be done. The effect of weak way connection action onthe strength and behavior of companion strong way connections, for columns participating inorthogonal lateral-force-resisting frames, has not been tested.

Commentary: Since 1985, the strong column/weak beam principle has beenrequired, but exceptions have been provided which permit weak columns in someinstances. These exceptions have not been revoked, and, in fact, the interest inredundancy generated by the Northridge failures has actually increased interestin their use, to the extent permitted, in moment frame systems, where all beam-column connections in the structure are connected for moment resistance andmade part of the lateral-force-resisting system. Considering the fact that columnsresisting flexural demands about their minor axes will not generally be capable ofdeveloping the beam flexural yield strength should permit consideration of thepre-Northridge connections for this use. On the other hand, where specific codeexceptions permit use of weak column systems for all or a large part of the lateralresistance a more conservative approach is merited. Use of weak column systemsas the primary lateral resistance is strongly discouraged and should not beconsidered as a desirable or acceptable method of avoiding beam flangeconnection concerns and reinforcing requirements.

Further, although logic would indicate that the strength demand on connectionsin weak column structures would be limited by column hinging, and that thereforethe beam-column connection should be protected, evidence suggests that this maynot be the case. It has been reported that a hospital structure affected by theNorthridge Earthquake experienced failure of almost all of its beam-columnconnections, despite having all or many weak column conditions.

7.9 Moment Frame Connections for Consideration in New Construction

The moment frame connection formerly prescribed by the code was configured to requiredevelopment of a plastic hinge in the beam adjacent to the beam-to-column connection. TheNorthridge experience and subsequent testing have shown that as the possible result of a number offactors, it is not reasonable to expect reliable development of plastic hinges at this location, at leastwithin the range of design parameters explored to date. Therefore, connections should be configuredto encourage plastic hinging action to other locations.


7-27

The types of connections described in the following subsections are felt to offer some promise ofproviding more reliable inelastic action in WSMFs, consistent with that assumed in the design of suchframes. It is of course assumed that the required joints, both welded and bolted, have been installedwith appropriate quality control as described previously.

Reference to laboratory testing is provided for those connection configurations for which researchhas been reported. However, it should be noted that none of these connections has been testedsufficiently at this time to permit unqualified use of the connection.

The figures provided in the following sections are schematic, indicating the general type ofconnection configuration being described. When designing connections patterned after the reportedtest data, the test specimen details included in the references should be reviewed to determine specificdetails not shown.

The SAC Joint Venture does not endorse or specifically recommend any of the connection detailsshown in this Section. These are presented only to acquaint the reader with available information onrepresentative testing of different connection configurations that have been performed by variousparties.

Commentary: With the large interest and availability of funding for research onsteel moment frame connections, any lists of connection concepts, such as theabove will necessarily become at least partially obsolete by the time they arepublished. With this in mind, it is very important that there be a publiclyaccessible center to accumulate testing results as they become available. It is therecommendation of this guideline that as efforts in this area progress, SACbecome the repository and distribution group for such information. It is hopedthat all engineers, researchers, and contractors responsible for tested connectionswill willingly share all information on the tests and designs with SAC, with thestructural engineering profession, and with the building construction industry.

The various connections suggested in this section were all nominally fullyrestrained (FR) connections. It has been suggested that partially restrained (PR)connections may be a cost-effective and reliable alternative to these connections. AISC and NSF are currently conducting research into the use of this system andit may become an attractive alternative in the future.

7.9.1 Cover Plate Connections

Figure 7-5 illustrates the basic configuration of cover plated connections. Short cover platesare added to the top and bottom flanges of the beam with fillet welds adequate to transfer the coverplate forces to the beam flanges. The bottom flange cover plate is shop welded to the column flangeand the beam bottom flange is field welded to the column flange and to the cover plate. The top flangeand the top flange cover plate are both field welded to the column flange with a common weld. Theweb connection may be either welded or high strength (slip critical) bolted. Limited testing of theseconnections (Engelhardt & Sabol - 1994), (Tsai & Popov -1988) has been performed.



7-28

A variation of this concept which has been tested successfully very recently (Forrel/ElsesserEngineers -1995), uses cover plates sized to take the full flange force, without direct welding of thebeam flanges themselves to the column. In this version of the detail, the cover plate provides a crosssectional area at the column face about 1.7 times that of the beam flange area. In the detail which hasbeen tested, a welded shear tab is used, and is designed to resist a significant portion of the plasticbending strength of the beam web.

T&B

Figure 7-5 - Cover Plate Connection

Design Issues: Approximately eight connections similar to that shown in Figure 7-5 have beenrecently tested (Engelhardt & Sabol - 1994), and they have demonstrated the ability to achieveacceptable levels of plastic rotation provided that the beam flange to column flange welding iscorrectly executed and through-thickness problems in the column flange are avoided. Thisconfiguration is relatively economical, compared to some other reinforced configurations, andhas limited architectural impact.

Six of eight connections tested by the University of Texas at Austin were able to achieveplastic rotations of at least 0.025 radians, or better. Strength loss at the extreme levels of plasticrotation did not reduce the flexural capacity to less than the plastic moment capacity of thesection based on minimum specified yield strength. One specimen achieved plastic rotations of0.015 radians when a brittle fracture of the CJP weld (type W2 failure) occurred. This maypartially be the result of a weld that was not executed in conformance with the specified weldingprocedure specification. The second unsuccessful test specimen achieved plastic rotations of0.005 radian when a section of the column flange (type C2 failure) occurred. A similar failureoccurred in recent testing by Popov of a specimen with cover plates having a somewhat modifiedplan shape.


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6 Specimens : >0.025 radian1 Specimen: 0.015 radian1 Specimen: 0.005 radian

Although apparently more reliable than the former prescriptive connection, thisconfiguration is subject to some of the same flaws including dependence on properly executedbeam flange to column flange welds, and through-thickness behavior of the column flange. Further these effects are somewhat exacerbated as the added effective thickness of the beamflange results in a much larger groove weld at the joint, and therefore potentially more severeproblems with brittle heat affected zones and lamellar defects in the column. Indeed, asignificant percentage of connections of this configuration have failed to produce the desiredamount of plastic rotation.

7.9.2 Flange Rib Connections

Figure 7-6 demonstrates the basic configuration for connections with flange ribs. The intentof the rib plates is to reduce the demand on the weld at the column flange and to shift the plastichinge from the column face.

21

Typ.

Typ.

Figure 7-6 - Flange Rib Connection

Design Issues: There is a limited body of testing of connections similar to these (Engelhardt &Sabol - 1994), (Tsai & Popov - 1988), and they have demonstrated the ability to achieveacceptable levels of plastic rotation provided that the girder flange welding is correctlyexecuted.



7-30


2 Specimens : >0.025 radian

Performance is dependent on properly executed girder flange welds. The joint can be subjectto through-thickness failures in the column flange, although it should be somewhat moreresistant to such failures than connections reinforced with cover plates, as the weld size isreduced. The size of the specimens tested required the use of two upstanding ribs per flange. This increased the costs significantly above those designs that use only one rib per flange,located above the girder center line. However, limited testing of the design with one rib at thegirder centerline (Tsai & Popov) indicated the potential for premature failure of the weld of therib to the girder at the outstanding edge. It should also be noted that the specimens tested byEngelhardt & Sabol, and reported above, incorporated columns with particularly heavy flanges. The ribs have the potential to cause high local stresses in the column flanges and thisconfiguration may not behave acceptably when used with lighter section. Preliminary reportsfrom fabricators and erectors indicate that the cost of this connection is quite high, relative toother configurations.

7.9.3 Bottom Haunch Connections

Figure 7-7 indicates several potential configurations for single, haunched beam-columnconnections. As with the cover plated and ribbed connections, the intent is to shift the plastichinge away from the column face and to reduce the demand on the CJP weld by increasing thedepth of the section. To date, the configuration incorporating the triangular haunch has beensubjected to limited testing. Testing of configurations incorporating the straight haunch arecurrently planned, but have not yet been performed.

WTWT

12

dd/

3

oror


Two tests have been performed to date, both successfully. Both tests were conducted in arepair/modification configuration. In one test, a portion of the girder top flange, adjacent to the


7-31

column, was replaced with a thicker plate. In addition, the bottom flange and haunch were bothwelded to the column. This specimen developed a plastic hinge within the beam span, outside thehaunched area and behaved acceptably. A second specimen did not have a thickened top flangeand the bottom girder flange was not welded to the column. Plastic behavior in this specimenoccurred outside the haunch at the bottom flange and adjacent to the column face at the topflange. Failure initiated in the girder at the juncture between the top flange and web, possiblycontributed to by buckling of the flange as well as lateral torsional buckling of the section. Fracture progressed slowly along the top fillet of the girder and eventually, traveled into theflange itself.

Design Issues: The haunch can be attached to the girder in the shop, reducing field erectioncosts. Weld sizes are smaller than in cover plated connections. The top flange is free ofobstructions.


Specimen 1:0.04 radian (w/o bottom flange weld and reinforced top flange)

Specimen 2:0.05 radian (with bottom flange weld and reinforced top flange)

Performance is dependent on properly executed complete joint penetration welds at thecolumn face. The joint can be subject to through-thickness flaws in the column flange; however,this connection may not be as sensitive to this potential problem because of the significantincrease in the effective depth of the beam section which can be achieved. Welding of the bottomhaunch requires overhead welding when relatively shallow haunches are used. The skewedgroove welds of the haunch flanges to the girder and column flanges may be difficult to execute. The increased depth of the beam, resulting from the haunch may have undesirable impact onarchitectural design. Unless the top flange is prevented from buckling at the face of the column,performance may not be adequate. For configurations incorporating straight haunches, thehaunch must be long, in order to adequately develop stress into the haunch, through the web. This tends to increase demands at the column face. Additional testing of all these configurationsis recommended.

7.9.4 Top and Bottom Haunch Connections

Figure 7-8 illustrates this connection configuration. Haunches are placed on both the top andbottom flanges. Two tests have been performed on connections utilizing this configuration; bothwere highly successful.



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WT

12

dd/

3

or

Figure 7-8 - Top and Bottom Haunch Connection

Design Issues: In two tests of this connection configuration performed to date, it has exhibitedextremely ductile behavior. Plastic rotations as large as 0.07 radians were obtained. Inaddition to having very good plastic capacity, the connection is highly redundant. If failureshould occur at one of the complete joint penetration welds of the haunch plate, significantresidual strength would be available from the remaining girder flange welds.

This is one of the more costly connection configurations. Some of this cost could be reduced byeliminating the welds between the girder flanges and columns, however, the performance of theconnection in that configuration has not been tested. The presence of the haunch at the top ofthe girder could be an architectural problem.

Quantitative Results: No. of Specimens Tested: 2Girder Size: W30 x 99Column Size W14 x 176Plastic Rotation achieved - 0.07 radians

7.9.5 Side-Plate Connections

This approach eliminates loading the column in the through-thickness direction by removingthe CJP welds at the girder flange and by shifting the plastic hinge from the column face. Thetension and compression forces are transferred from the girder flanges into the column throughfillet welds. A mechanism to provide a direct connection between the column panel zone and thebeam flanges is required; the difficulty appears to be equalizing the width of the beam and columnflanges.

At least two configurations of side-plated connections have been tested. One set, shown inFigure 7-9, utilized flat bars at the top and bottom girder flanges, to transfer flange forces to thecolumn (Engelhardt & Sabol - 1994). The girder was widened to the width of the column with


7-33

the use of filler plates. The specimens achieved plastic rotations of 0.015 radians, however,fractures developed within the welds connecting the beam flange to the transfer plates. Failure ofthe shear tab, and finally the side plates themselves followed the initiation of these fractures. It isbelieved that the unsuccessful behavior of this particular specimen was related to the method usedto increase the width of the beam flange to equal that of the column flange, using a combinationof a filler bar and welding. Other approaches, such as providing a full width cover plate for thegirder flanges, may provide better performance.

Tested Configuration Possible Alternative

Figure 7-9 Side Plate Connection

Design Issues: This connection avoids both the large complete joint penetration welds of thebeam flange to the column and the potential for through-thickness failure of the column flange.Much of the additional fabrication can be performed in the shop.

This connection did not demonstrate adequate plastic rotation capacity in the configurationstested to date. Additional testing is required to determine if modified configurations willperform in a more acceptable manner.

Quantitative Results:Separate Top & Bottom Side PlatesNo. of specimens tested: 2Girder Size: W36 x 150Column Size: W14 x 455Plastic Rotation achieved-

2 Specimens :0.015 radian

A second, proprietary configuration, is shown in Figure 7-10. Three specimens haveundergone full-scale testing to date and achieved large plastic rotations. Loss of strength at largeplastic rotation demands was comparable to that of other successful connections. The developer



7-34

of this connection has applied for US and foreign patents. Further information on technical datafor this configuration, may be obtained from the developer.

NOTICE OF CONFIDENTIAL INFORMATION:WARNING: The information presented in this figure is PROPRIETARY. US and Foreign Patents havebeen applied for. Use of this information is strictly prohibited except as authorized in writing by thedeveloper. Violators shall be prosecuted in accordance with US and Foreign Patent Intellectual PropertyLaws.

Figure 7-10 - Proprietary Side Plate Connection

Design Issues: Testing of three prototype specimens (Uang & Latham - 1995) indicates that thisproprietary connection has the ability to achieve very satisfactory levels of plastic rotationwithout relying on sensitive CJP welds between the column and girder flanges or specifying weldmaterial with notch toughness. The elimination of the through-thickness loading of the flangemay result in higher levels of connection reliability. Due to the exclusive use of fillet welds,special inspection requirements for welding and bolting can be reduced significantly with thisconnection.

This connection is proprietary and license fees are associated with its use. The cost of theconnection may be greater than some of the other modification methods discussed above;however, this cost differential may not be as great on double-sided connections because much ofthe cost is associated with the side plates which are similar for both single-sided and double-sided connections. However, double sided connections will require doubling the sizes of thewelds which deliver the forces to the columns, and potentially increasing plate thickness as well. The connection of beams framing into the minor axis of the column are made more difficult bythis connection, particularly if they must be connected for moment resistance. Publicly bidprojects will have to develop performance specifications to permit other connections to beconsidered for use unless a strong case for sole-sourcing the connection can be made.

Quantitative Results:No. of specimens tested: 3Girder Size: W36 x 150


7-35

Column Size: W14 x 426Plastic Rotation achieved-

3 Specimens : >0.042 radian to 0.06 radian

7.9.6 Reduced Beam Section Connections

In this connection, the cross section of the beam is intentionally reduced within a segment, toproduce an intended plastic hinge zone or fuse, located within the beam span, away from thecolumn face. Several ways of performing this cross section reduction have been proposed. Onemethod includes removal of a portion of the flanges, symmetrical about the beam centerline, in aso-called “dog bone” profile. Care should be taken with this approach to provide for smoothlycontoured transitions to avoid the creation of stress risers which could initiate fracture. It has alsobeen proposed to create the reduced section of beam by drilling a series of holes in the beamflanges. Figure 7-11 illustrates both concepts. The most successful configurations taper thereduced section, through the use of unsymmetrical cut-outs, or variable size holes, to balance thecross section and the flexural demand.

Testing of this concept was first performed by a private party, and US patents were appliedfor and granted. These patents have now been released. Limited testing of both “dog-bone” anddrilled hole configurations have been performed in Taiwan (Chen and Yeh - 1995). The AmericanInstitute of Steel Construction is currently performing additional tests of this configuration(Smith-Emery - 1995), however the full results of this testing are not yet available.

There is a concern that the presence of a concrete slab at the beam top flange would tend tolimit the effectiveness of the reduced section of that flange, particularly when loading places thetop flange into compression. It may be possible to mitigate this effect with proper detailing of theslab.

WeakenedSegment

Symmetrical Unsymmetrical

Figure 7-11 - Reduced Beam Section Connection

Design Issues: This connection type is potentially the most economical of the several types whichhave been suggested. The reliability of this connection type is dependent on the quality of thecomplete joint penetration weld of the beam to column flange, and the through-thickness



7-36

behavior of the column flange. If the slab is not appropriately detailed, it may inhibit theintended “fuse” behavior of the reduced section beam segment. It is not clear at this timewhether it would be necessary to use larger beams with this detail to attain the same overallsystem strength and stiffness obtained with other configurations. In limited testing conducted todate of the unsymmetrical “dog-bone” configuration (Smith-Emery - 1995), the plastic hingingwhich occurred at the reduced section was less prone to buckling of the flanges than in some ofthe other configurations which have been tested, due to the very compact nature of the flange inthe region of the plastic hinge.

Quantitative Results: No. of specimens tested: 2Girder Size: W30 x 99Column Size: W14 x 176Plastic Rotation achieved- 0.03 radian

7.9.7 Slip - Friction Energy Dissipating Connection

This connection uses high strength bolts and slotted holes to develop the flange forces into thecolumns. A brass shim in the shear transfer plane provides for controlled friction force. Inconcept, slip along this bolted connection limits the amount of force which can be transferred tothe column and allows plastic deformation to occur in a benign manner. Two alternativeconfigurations have been suggested for attachment of the flanges to the column. Oneincorporates bolted “T” sections and the other welded plates. Figure 7-12 shows the bolted “T”configuration.

To date, two tests have been performed on the bolted “T” configuration (Popov & Yang-1995). Results were excellent with large inelastic displacements obtained without strength orstiffness degradation.

Type “X” bolts

SteelShims

Tee Section w/ slotted holes

Brass Shims

Figure 7-12 - Slip Friction Energy Dissipation Connection


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Design Issues: In the limited testing performed, this connection was able to accommodate largeinelastic displacements without damage to the connection or beams. This connection can beassembled in the field without welding and can accommodate large plastic rotations withoutpermanent damage to the structure.

The connection is sensitive to fit-up, cleanliness of the faying surfaces and tension in the high-strength bolts, and therefore will require careful field quality control. As with the “reducedsection beam” connections, this connection may be sensitive to the presence of a slab andcareful detailing of the slab to permit the expected connection rotations to occur may berequired. The strength that can be developed by this connection is limited by the number of boltsthat can be practically placed. It may not be suitable for use with larger members with highstrength demands. The brass shims, used at the slip plane interface are quite costly. Metalparts kept in contact under pressure over a period of years may tend to become partially weldedtogether, potentially reducing the effectiveness of the connection with time. Additional researchis required on this effect.

7.9.8 Column-Tree Connection

This concept has been widely used in Japan, with mixed success. Short stubs of girders arefabricated and shop welded to the column. Field connection to the balance of the girder is madewith bolted connections. The girder stubs can be intentionally fabricated stronger than thebalance of the girder, to force yielding and formation of a plastic hinge away from the column. Figure 7-13 demonstrates the basic concept.

Extensive testing of this connection has not been performed in the US Some variations of thisconnection, in common use in Japan, performed very poorly in the recent Kobe Earthquake(Watabe-1995). In at least one version of this connection, the beam stub ran continuouslythrough the connection and the columns were shop welded to the top and bottom of this stub. Anumber of these connections experienced fracture of the shop weld of the column to the beamstub. However, it is reasonable to expect that configurations for this concept can be developedthat would permit more favorable behavior.

Figure 7-13 - Column Tree Connection



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Design Issues: The basic advantage to this connection is that critical welding can be performedin the fabrication shop where it should be possible to attain better quality control. In addition,field erection costs are reduced through the use of bolted field connections.

Testing of this connection in configurations similar to US construction practice has not beenperformed. Some configurations utilized in Japan performed poorly in the Kobe Earthquake. The connection is dependent on the quality of beam flange to column flange welding and thethrough-thickness behavior of the column flange. Transportation and handling of tree columnsis probably somewhat more difficult and expensive than for standard columns.

7.9.9 Slotted Web Connections

In the former prescriptive connection, in which the beam flanges were welded directly to thecolumn flanges, beam flexural stress was transferred into the column web through the combinedaction of direct tension across the column flange, opposite the column web, and through flexureof the column flange. This stress transfer mechanism results in a large stress concentration at thecenter of the beam flange, opposite the column web. Recent research (Allen, et. al. - 1995)indicates that the provision of continuity plates within the column panel zone reduces this stressconcentration somewhat, but not completely. The intent of slotted web connections is to furtherreduce this stress concentration and to achieve a uniform distribution of flexural stress across thebeam flange at the connection. Figure 7-14 indicates one configuration for this connection typethat has been successfully tested. In this configuration, vertical plates are placed between thecolumn flanges, opposite the edges of the top and bottom beam flanges to stiffen the outstandingcolumn flanges and draw flexural stress away from the center of the beam flange. Horizontalplates are placed between these vertical plates and the column web to transfer shear stresses to thepanel zone. The web itself is softened with the cutting of a vertical slot in the column web,opposite the beam flange. High fidelity finite element models were utilized to confirm that anearly uniform distribution of stress occurs across the beam flange.

Typical

PP

1/4” Slot

3/4” Hole

Figure 7-14 - Slotted Web Connection


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Design Issues: This detail is potentially quite economical, entailing somewhat more shopfabrication than the former prescriptive connection, but similar levels of field erection work. Contrary to the recommendations contained in these Interim Guidelines, this connection doesnot shift the location of plastic hinging away from the column face. However, two connectionssimilar to that shown in Figure 7-14 have recently been tested succesfully (Allen. - 1995). Theconnection detail is sensitive to the quality of welding employed in the critical welds, includingthose between the beam and column flanges, and between the vertical and horizontal plates andthe column elements. It has been reported that one specimen, with a known defect in the beamflange to column flange weld was informally tested and failed at low levels of loading.

The detail is also sensitive to the balance in stiffness of the various plates and flanges. Forconfigurations other than those tested, detailed finite element analyses may be necessary toconfirm that the desired uniform stress distribution is achieved. The developer of this detailindicates that for certain column profiles, it may be possible to omit the vertical slots in thecolumn web and still achieve the desired uniform beam flange stress distribution.

This detail may also be sensitive to the toughness of the column base metal at the region ofthe fillet between the web and flanges. In heavy shapes produced by some rolling processes themetal in this region may have substantially reduced toughness properties relative to the balanceof the section. This condition, coupled with local stress concentrations induced by the slot in theweb may have the potential to initiate premature fracture. The developer believes that it isessential to perform detailed analyses of the connection configuration, in order to avoid suchproblems. Popov tested one specimen incorporating a locally softened web, but without thevertical and horizontal stiffener plates contained in the detail shown in Figure 7-14. Thatspecimen failed by brittle fracture through the column flange which progressed into the holes cutinto the web. The stress patterns induced in that specimen, however, were significantly differentthan those which occur in the detail shown in the figure.

Quantitative Results: Number of specimens tested: 2Girder Size: W 27x94Column Size: W 14x176Plastic Rotation Achieved:

Specimen 1: 0.025 radianSpecimen 2: 0.030 radian

7.10 Other Types of Welded Connection Structures

These Interim Guidelines have focused on the design of the moment and shear resisting FRconnections in moment frame systems in which the lateral forces are resisted by bending in beamsand columns. In addition to moment frame systems, there are a number of other system typeswhich conceivably could exhibit connection or joint distress similar to that seen in moment framesin Northridge when deformed under high intensity earthquake motions. Except for one detail of awelded base plate at the base of a braced frame, there has been no reported damage to thefollowing systems from the Northridge earthquake. The response to earthquake motions,however, presents similar potential conditions to those found in moment frame connections.



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Therefore when designing new construction, close attention should be given to these structuralsystems and details so that damage similar to that observed on WSMF systems can be avoided infuture earthquakes.

7.10.1 Eccentrically Braced Frames (EBF)

EBF provisions in the code require the use of "link beam" elements. Link beams are usuallydesigned to yield in shear in the web, but can be designed to yield in flexure. In someconfigurations, the link beam is connected to the column flange in a manner nearly identical to theconnection of the WSMF. The connection of the brace element to the beam also connects to thebeam flange, but the connection has additional design requirements which modify the type ofconnection. In addition to connection concerns for EBFs, the currently recognized variability ofsteel strengths should be considered in designing the components of the EBF. It is recommendedthat connections in EBFs intended to resist plastic rotation demands be designed the same asWSMF connections, as previously defined in these Interim Guidelines, with due considerationgiven to the additional shear forces which may be induced in the link beams.

Commentary: Although the code provisions are intended to cause the link beamsconnected to columns to yield in shear (length of link limited to 1.6MS/VS), thelink may be at or near its bending strength when shear yielding would occur. Thus the connection to the column flanges may be vulnerable to the same orsimilar problems as those exhibited by WSMF connections during the Northridgeearthquake.

Recognition of the probable strength of steel in the link beam could be criticalto the performance of these structures. The inelastic behavior of EBF structuresis intended to be controlled through yielding of these links. If link beams arefabricated from excessively strong material, they may not yield before other partsof the frame become damaged.

7.10.2 Dual Systems

The provisions for Dual Systems in the code require that the system include a SpecialMoment-Resisting Frame (SMRF) designed according to the same provisions as if it were theprimary system but capable of resisting at least 25% of the required lateral forces. In addition, itis required to have a primary system consisting of either concrete shear walls, SpecialConcentrically Braced Frames (SCBF), Concentrically Braced Frames (CBF), or EccentricallyBraced Frames (EBFs). Connection design for moment frames used in Dual Systems shouldconform to the recommendations of these Interim Guidelines for SMRF systems.

Commentary: Prior to the 1967 UBC, Dual Systems design required a primarysystem (shear walls or CBFs) capable of resisting 100% of the required lateralforces in conjunction with a "back-up" SMRF capable of resisting at least 25% ofthe total forces. The assumption for this type of system was that the SMRF wouldtake over and prevent collapse of the structure in the event of failure of the stiffer,


7-41

but ostensibly less ductile, primary system. In this concept of design, the SMRFwas there solely for redundancy. In the 1967 UBC, an additional provision wasadded in which the primary system and the SMRF were required to be designed toshare the total required lateral force according to their elastic stiffness. In the1988 UBC, the requirement that the primary system (shear walls or bracedframes) be designed to resist 100% of the required total force was eliminated, butthe other two requirements remain. This potentially makes it even more importantthat the SMRF portion of a dual system have adequate ductility to survive a majorevent.

In general, dual systems have been a somewhat controversial system. Someengineers believe that the added redundancy provided by the backup system isquite beneficial while others do not believe that the relatively weak and oftenflexible back-up system improves building performance significantly. Littleanalytical research of these systems has been performed. Such research would bebeneficial, however, in providing guidance as to the amount of ductility requiredof the backup frame system.

7.10.3 Welded Base Plate Details

The detail of concern is in any system of steel framing where a column, which is subject tohigh axial tension or flexure, or both, is directly welded to its baseplate in a manner similar to thatused for beam-to-column moment connections. Additional concerns occur when anchor bolts arefastened to the base plate in close proximity to the bottom of the column.

Commentary: When a column is welded directly to the base plate and has thepotential for being loaded with significant tension or tension in combination withflexure, CJP welds and the through-thickness strength of the base plate arerequired to resist the tensile forces. The combination of uncertainty of thethrough-thickness strength and the uncertainty of the axial loading suggests thatanother type of connection detail should be chosen. Frequently, the anchor boltsare placed close to the face of the column flanges. If the anchor bolts are strongenough so that the mechanism of failure is flexure in the base plate, the shortflexural span makes it impossible for flexural yielding to occur and may result ina brittle fracture of the plate or of the CJP welds.

7.10.4 Vierendeel Truss Systems

A Vierendeel Truss (VT) is a type of truss without diagonals in which shear forces are resistedby the vertical members and chords, acting together as moment-resisting frames. VT's may havediagonals in some bays in some designs, but may also be designed to rely totally on the verticals. Where both chords and verticals of VT's are wide flange shapes, the connections of the verticalsto the chords and the chords to the columns are often detailed in the same manner as the beam-to-column flange connection of WSMFs. A variation on the conventional horizontal VierendeelTruss which deserves similar attention is a system where vertical loads in a discontinuous column



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are supported by moment connected beams at several floors, rather than by a single transfer girderabove the location of the column discontinuity.

Commentary: Considering the brittle nature of the damage to steel structures dueto the Northridge earthquake, Engineers should have some concerns about VTsystems as described above even when they are designed to carry vertical loadsonly, particularly if the loadings are variable and could significantly exceeddesign loads in extreme cases. Where VT's are a part of the lateral system, eitherserving simply as a moment frame girder or as a transfer girder, seismicdeformations potentially could lead to yielding at the connections of the trussverticals to the chords. Such connections should be designed in the same manneras the beam-to-column connections of WSMFs. If such yielding is possible, theeffect of such yielding on the vertical load capacity and deformation should beinvestigated.

7.10.5 Moment Frame Tubular Systems

This type of lateral-force-resisting system is common in very tall buildings. The momentframe is arranged with relatively short spacing of columns around the perimeter of the structure. The system is actually a special type of WSMF which has very stiff beams so that the chord forcesat the end of a moment frame can be distributed to adjacent columns perpendicular to the plane oraround the corner of the moment frame. The system is defined as a three-dimensional spaceframe structure composed of three or more frames connected at the corners (or intersections) toform a vertical tube-like structure (or a structure composed of several adjacent tubes). Ofparticular concern is the short beam span which renders some of the solutions for localstrengthening of beams difficult to achieve. On the other hand, plastic rotational demands due tohigh seismic forces may be shown to be very low in some designs.

Commentary: Moment frame tube structures are normally very redundantsystems with many moment-resisting members and connections. A thoroughanalysis of the structural system should be made to determine what potentialplastic rotation demand would be required on the connections. With very tallbuildings, seismic response becomes more heavily influenced by the higher modesof vibration, and design of members and connections might be controlled by windforces.

7.10.6 Welded Connections of Collectors, Ties and Diaphragm Chords

These members are part of a building’s lateral-force-resisting system. They are usuallyhorizontal members which, in addition to supporting adjacent gravity load, are also required totransmit large axial tensile and compressive forces. If development of the tensile and flexuralforces at the connections to the columns requires welding of the member flanges to the column,all of the recommendations for WSMF connections should be followed.


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Commentary: Where chord, collector and time members are nearly pure axialmembers, such as would occur in a building with shear walls or braced framesthat is laterally very stiff, the former prescriptive connection may be found to besufficient, depending on the size of the member..

7.10.7 Welded Column Splices

Even though no column splice damage has been reported from the Northridge earthquake,column splices incorporating partial penetration flange welds should be used cautiously,particularly if the potential for large tensile and/or flexural forces are present. Partial penetrationwelds result in a crack-like feature, which can initiate fracture under conditions of high stress. Anumber of structures experienced failure of column splices in the 1995 Kobe Earthquake, withsome such failures leading to structural collapse.

Commentary: Of particular concern would be the use of partial penetration buttwelds on the column flanges. The configuration of partial penetration weldsprovides a notch on the inner edge of the weld. Thus other methods of effecting acolumn splice should be used if significant or unpredictable tensile or flexuralforces are possible. When considering bending in column splices of momentframes, it has been shown by inelastic time-history analyses that reliance shouldnot be placed on the inflection point occurring at the mid height of the column. The studies show that the location of the hinge can change significantly as thestructure deforms, both due to higher mode effects and due to the inelasticresponse of the members.

7.10.8 Built-up Moment Frame Members

Built-up beams and columns used in moment frame systems have the same concerns in thedesign of connections as the rolled shape systems previously discussed. The welds connecting thevarious component parts of the built-up members should be designed to be capable of resisting theeffects of potential plastic behavior and connections of built-up members should be designed topreclude reliance on yielding of steel in areas of confined or restrained joints.

Commentary: In beams, the effect of the shape of the components, including therelative thickness of flanges and web can be significant in determining the forcesrequired to be developed in the various joints in the connection of the beam to thecolumn. Also the joint between the web and flanges of the built-up beams,particularly in the areas of potential plastic hinges, should be designed to becapable of permitting flange buckling without weld failure.

In H shaped built-up columns, the welded joints between the web and flangesshould be designed to develop the panel zone shears based on the probablelocation of the plastic hinges. In tubular or box shaped columns, the placing ofthe plates and the selection of the type of weld connecting webs to flanges is



7-44

important in providing adequate joints to resist the forces in the beam-to-columnconnection zone. Some testing has been performed Taiwan (Tsai - 1995).

Interim Guidelines:Evaluation, Repair, Modification andDesign of Steel Moment Frames Chapter 8 - Metallurgy & Welding

8-1

8. METALLURGY & WELDING

Standard industry specifications for construction materials and processes permit wide variation instrength, toughness and other properties that can be critical to structural performance. ThisChapter provides basic information on the variations in properties that occur, practical steps anengineer can take to control critical properties to acceptable levels of tolerance, and the specificinstances when such measures may be appropriate.

8.1 Parent Materials

8.1.1 Steels

Designers should specify materials which are readily available for building construction and whichwill provide suitable ductility and weldability for seismic applications. Structural steels which may beused in the lateral-force-resisting systems for structures designed for seismic resistance without specialqualification include those contained in Table 8-1. Refer to the applicable ASTM reference standardfor detailed information.

Table 8-1 - Structural Steel Prequalified for Use in Seismic Lateral-Force-Resisting Systems

ASTM Specification DescriptionASTM A36 Carbon Structural SteelASTM A283Grade D

Low and Intermediate Tensile Strength Carbon Steel Plates

ASTM A500(Grades B & C)

Cold-Formed Welded & Seamless Carbon Steel Structural Tubing in Rounds &Shapes

ASTM A501 Hot-Formed Welded & Seamless Carbon Steel Structural TubingASTM A572(Grades 42 & 50)

High-Strength Low-Alloy Columbium-Vanadium Steels of Structural Quality

ASTM A588 High-Strength Low-Alloy Structural Steel (weathering steel)

Structural steels which may be used in the lateral-force-resisting systems of structures designed forseismic resistance with special permission of the building official are those listed in Table 8-2. Steelmeeting these specifications has not been demonstrated to have adequate weldability or ductility forgeneral purpose application in seismic-force-resisting systems, although it may well possess suchcharacteristics. In order to demonstrate the acceptability of these materials for such use in WSMFconstruction it is recommended that connections be qualified by test, in accordance with the guidelinesof Chapter 7. The test specimens should be fabricated out of the steel using those welding proceduresproposed for use in the actual work.


Chapter 8 - Metallurgy & Welding Design of Steel Moment Frames

8-2

Table 8-2 - Non-prequalified Structural Steel

ASTM Specification DescriptionASTM A242 High-Strength Low-Alloy Structural SteelASTM A709 Structural Steel for BridgesASTM A913 High-Strength Low-Alloy Steel Shapes of Structural Quality, Produced by

Quenching & Self-Tempering Process

Commentary: Many WSMF structures designed in the last 10 years incorporatedASTM A36 steel for the beams and ASTM A572 grade 50 steel for the columns. This provided an economical way to design structures for the strong column -weak beam provisions contained in the building code. Recent studies conductedby the Structural Shape Producers Council (SSPC), however, indicate thatmaterial produced to the A36 specification has wide variation in strengthproperties with actual yield strengths that often exceed 50 ksi. This widevariation makes prediction of connection and frame behavior difficult. Somehave postulated that one of the contributing causes to damage experienced in theNorthridge Earthquake was inadvertent pairing of overly strong beams withaverage strength columns.

The AISC and SSPC have been working for several years to develop a newspecification for structural steel that would have both minimum and maximumyield values defined and provide for a margin between maximum yield andminimum ultimate tensile stress. AISC recently submitted such a specification,for a material with 50 ksi specified yield strength, to ASTM for development intoa standard specification. It is anticipated that domestic mills will beginproducing structural shapes to this specification within a few years and thateventually, this new material will replace A36 as the standard structural materialfor incorporation into lateral-force-resisting systems.

Under certain circumstances it may be desirable to specify steels that are notrecognized under the UBC for use in lateral-force-resisting systems. Forinstance, ASTM A709 might be specified if the designer wanted to place limits ontoughness for fracture-critical applications. In addition, designers may wish tobegin incorporating ASTM A913, Grade 65 steel, as well as other higher strengthmaterials, into projects, in order to again be able to economically design forstrong column - weak beam conditions. Designers should be aware, however, thatthese alternative steel materials may not be readily available. It is alsoimportant when using such non-prequalified steel materials, that precautions betaken to ensure adequate weldability of the material and that it has sufficientductility to perform under the severe loadings produced by earthquakes. Thecyclic test program recommended by these Interim Guidelines for qualification ofconnection designs, by test, is believed to be an adequate approach to qualifyalternative steel material for such use as well.


8-3

Note that ASTM A709 steel, although not listed in the building code asprequalified for use in lateral-force-resisting systems, actually meets all of therequirements for ASTM A36 and ASTM A572. Consequently, specialqualification of the use of this steel should not be required.

8.1.2 Chemistry

ASTM specifications define chemical requirements for each steel. A chemical analysis is performedby the producer on each heat of steel. End product analyses can also be specified on certain products. A certified mill test report is furnished to the customer with the material. The designer should specifythat copies of the mill test reports be submitted for his/her conformance review. In general, ASTMspecifications for structural steels include maximum limits on carbon, manganese, silicon, phosphorousand sulfur. Ranges and minimums are also limited on other elements in certain steels. Chromium,columbium, copper, molybdenum, nickel and vanadium may be added to enhance strength, toughness,weldability and corrosion resistance. These chemical requirements may vary with the specific productand shape within any given specification.

Commentary: Some concern has been expressed with respect to the movement inthe steel producing industry of utilizing more recycled steel in its processes. Thisresults in added trace elements not limited by current specifications. Althoughthese have not been shown quantitatively to be detrimental to the performance ofwelding on the above steels, a new specification for structural steel proposed byAISC does place more control on these trace elements. Mill test reports nowinclude elements not limited in some or all of the specifications. They includecopper, columbium, chromium, nickel, molybdenum, silicon and vanadium. Theanalysis and reporting of an expanded set of elements should be possible, andcould be beneficial in the preparation of welding procedure specifications (WPSs)by the welding engineer if critical welding parameters are required. Modernspectrographs used by the mills are capable of automated analyses. Whenrequired by the engineer, a request for special supplemental requests should benoted in the contract documents.

8.1.3 Tensile/Elongation Properties

Mechanical property test specimens are taken from rolled shapes or plates at the rolling mill in themanner and location prescribed by ASTM A6 and ASTM A370. Table 8-3 gives the basic mechanicalrequirements for commonly used structural steels. Properties specified, and controlled by the mills, incurrent practice include minimum yield strength, ultimate tensile strength and minimum elongation. However, there can be considerable variability in the actual properties of steel meeting thesespecifications.

SSPC, in cooperation with SEAOC, has collected statistical data on the strength characteristics oftwo grades (ASTM A36 and ASTM A572 Grade 50) of structural steels, based on mill test reportsfrom selected domestic producers for the 1992 production year. Data were also collected for "Dual



8-4

Grade" material that was certified by the producers as complying with both ASTM A36 and ASTMA572 Grade 50. Table 8-4 summarizes these results as well as data provided by a single producer forASTM A913 material.

Table 8-3 - Typical Tensile Requirements for Structural Shapes

ASTMMinimum Yield

Strength, KsiUltimate Tensile

Strength, KsiMinimum Elongation

%in 2 inches

Minimum Elongation%

in 8 inchesA36 36 58-801 212 20A242 424 63 MIN. 213 18

A572, GR50 50 65 MIN. 212 18A588 50 70 MIN. 213 18

A709, GR36 36 58-80 212 20A709, GR50 50 65 MIN. 21 18A913, GR50 50 65 MIN. 21 18A913, GR65 65 80 MIN. 17 15

Notes: 1- No maximum for shapes greater than 426 lb./ft.2- Minimum is 19% for shapes greater than 426 lb. /ft.3- Minimum is 18% for shapes greater than 426 lb./ft.4. Minimum is 50 ksi for Shape Groups 1 and 2, 46 ksi for Shape Group 3

Unless special precautions are taken to limit the actual strength of material incorporated into thework to defined levels, new material specified as ASTM A36 should be assumed to be the dual gradefor connection demand calculations, whenever the assumption of a higher strength will result in a moreconservative design condition.

Commentary: The data given in Table 8-4 for A36 and A572 Grade 50 issomewhat weighted by the lighter, Group 1 shapes that will not ordinarily be usedin WSMF applications. Excluding Group 1 shapes and combining the DualGrade and A572 Grade 50 data results in a mean yield strength of 48 ksi for A36and 57 ksi for A572 Grade 50 steel. It should also be noted that 50% of thematerial actually incorporated in a project will have yield strengths that exceedthese mean values. For the design of facilities with stringent requirements forlimiting post-earthquake damage, consideration of more conservative estimates ofthe actual yield strength may be warranted.

In wide flange sections the tensile test coupons are currently taken from theweb. The amount of reduction rolling, finish rolling temperatures and coolingconditions affect the tensile and impact properties in different areas of themember. Typically, the web exhibits about five percent higher strength than theflanges due to faster cooling.

Table 8-4 - Statistics for Structural Shapes1

Statistic A 36 DualGrade

GRADE

A572GR50

A913GR65


8-5

Statistic A 36 DualGrade

GRADE

A572GR50

A913GR65

Yield Point (ksi) Mean 49.2 55.2 57.6 75.3 Minimum 36.0 50.0 50.0 68.2 Maximum 72.4 71.1 79.5 84.1 Standard Deviation [ s ] 4.9 3.7 5.1 4.0 Mean + 1 s 54.1 58.9 62.7 79.3

Tensile Strength (ksi) Mean 68.5 73.2 75.6 89.7 Minimum 58.0 65.0 65.0 83.4 Maximum 88.5 80.0 104.0 99.6 Standard Deviation [ s ] 4.6 3.3 6.2 3.5 Mean + 1 s 73.1 76.5 81.8 93.2

Yield/Tensile Ratio Mean 0.72 0.75 0.76 0.84 Minimum 0.51 0.65 0.62 0.75 Maximum 0.93 0.92 0.95 0.90 Standard Deviation [ s ] 0.06 0.04 0.05 0.03 Mean + 1 s 0.78 0.79 0.81 0.87 Mean - 1 s 0.66 0.71 0.71 0.81

1: The data presented for ASTM A36, “Dual Grade” and ASTM A572 Grade 50 were included aspart of the SSPC study (SSPC-1994). The data for ASTM A913 were derived from a singleproducer and may not be available from all producers.

Design professionals should be aware of the variation in actual propertiespermitted by the ASTM specifications. This is especially important for yieldstrength. Yield strengths for ASTM A36 material have consistently increased overthe last 15 years so that several grades of steel may have the same properties orreversed properties, with respect to beams and columns, from those the designerintended. Investigations of structures damaged by the Northridge earthquakefound some WSMF connections in which beam yield strength exceeded columnyield strength despite the opposite intent of the designer.

As an example of the variations which can be found, Table 8-5 presents thevariation in material properties found within a single building affected by theNorthridge earthquake. Properties shown include measured yield strength (Fya,),measured tensile strength (Fua ) and Charpy V-Notch energy rating (CVN).



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Table 8-5 - Sample Steel Properties from a Building Affected by the Northridge Earthquake

Shape Fya1 ksi Fua, ksi CVN, ft-lb.

W36 X 182 38.0 69.3 18

W36 X 230 49.3 71.7 195Note 1 - ASTM A36 material was specified for both structures.

The practice of dual certification of A36 and A572, Grade 50 can result inmean yield strengths that are fifty percent higher than the specified yield of A36. Since there is no practical way to discern whether dual grade steel will besupplied, unless direct purchase of steel from specific suppliers is made, in theabsence of such procurement practices, the prudent action for determiningconnection requirements, where higher strengths could be detrimental to thedesign, would be to assume the dual grade material whenever A36 or A572 Grade50 is specified.

8.1.4 Toughness Properties

For critical connections, non-redundant components and unusual or difficult geometries involvingGroup 3 (with flanges 11/2 inches or thicker) 4 and 5 shapes and plates and built-up sections over twoinches thick with welded connections, the designer should consider specifying toughness requirementson the parent materials. A Charpy V-Notch (CVN) value of 20 ft.-lb. at 70 degrees F. should bespecified when toughness is deemed necessary for an application. Refer to Figure 8-1 for typical CVNtest specimen locations. The impact test should be conducted in accordance with ASTM A673,frequency H, with the following exceptions:

a) The center longitudinal axis of the specimens should be located as near as practicable tomidway between the inner flange surface and the center of the flange thickness at theintersection of the web mid-thickness. Refer to AISC LRFD specification, Section A3-1c,Heavy Shapes (American Institute of Steel Construction - 1993)

b) Tests should be conducted by the producer on material selected from a locationrepresenting the top of each ingot or part of an ingot used to produce the productrepresented by these tests. For the continuous casting process, the sample may be taken atrandom.


8-7

TypicalCVNSpecimenASTM A673

tf/2

CLbf/2bf/3

CVNSpecimenAISC- LRFDA3-1c

Figure 8-1 - Standard Locations for Charpy V-Notch Specimen Extraction, Longitudinal Only

Commentary: Many variables are recognized in analyzing the metallurgy ofWSMF members. Until more research is available on the through-thicknessproperties of members thicker than two inches, a conservative approach isindicated. Specifying toughness properties in critical, unusual or non-redundantconnections should be considered.

As temperature decreases or strain rate increases, toughness propertiesdecrease. Charpy V-notch impact (CVN) tests, pre-cracked CVN tests and otherfracture toughness tests can identify the nil ductility temperature (NDT) - thetemperature below which a material loses all ductility and fractures in a brittlemanner. On a microscopic level, this equates to a change in the fracturemechanism from shear to cleavage. Fracture that occurs by cleavage at anominal tensile stress below yield is referred to as a brittle fracture. A brittlefracture can occur in structural steel when a particular combination of lowtemperature, tensile stress, high strain rate and a metallurgical or mechanicalnotch is present.

Plastic deformation can only occur through shear stress. Shear stress isgenerated when uniaxial or bi-axial straining occurs. In tri-axial stress states,the maximum shear stress approaches zero as the principal stresses increase. When these stresses approach equality, a cleavage failure can occur. Weldingand other sources of residual stresses can set up a state of tri-axial stress leadingto brittle fractures.

The necessity for minimum toughness requirements is not agreed to by all. There is also disagreement as to how much toughness should be required. The



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AWS Presidential Task Group recommended toughness values of 15 ft-lb. atdifferent temperatures, depending on the anticipated service conditions. Atemperature of 70 degrees F was recommended for enclosed structures and 40degrees F for exposed structures. The 1993 AISC LRFD Specification, SectionA3-1c, Heavy Shapes, requires toughness testing [Charpy V-Notch] under thefollowing conditions for Group 4 and 5 shapes and plates exceeding 2 inches inthickness: a) When spliced using complete joint penetration welds; b) whencomplete joint penetration welds through the thickness are used in connectionssubjected to primary tensile stress due to tension or flexure of such members.” Where toughness is required, the minimum value should be 20 ft-lb. at 70°F.

Plates thicker than two inches and sections with flanges thicker than twoinches can be expected to have significantly variable grain sizes across thesection. The slower cooling rate of the web-flange intersection in thick sectionsproduces a larger grain size which exhibits less ductility and notch toughness.

ANSI/ASTM A673 and A370 establish the procedure for longitudinal CharpyV-notch testing. The impact properties of steel can vary within the same heat andpiece, be it as-rolled, controlled rolled, or heat treated. Normalizing orquenching and tempering will reduce the degree of variation. Three specimensare taken from a single test coupon or location. The average must exceed thespecified minimum, but one value may be less than the specified minimum butmust be greater than the larger of two thirds of the specified minimum or 5 ft-lb. The longitudinal axis of the specimen is parallel to the longitudinal axis of theshape or final rolling direction for plate. For shapes, the specimen is taken fromthe flange 1/3 the distance from the edge of the flange to the web. The frequencyof testing [heat or piece], the test temperature, and the absorbed energy arespecified by the user. [NOTE: heat testing (frequency H) for shapes, means oneCVN test set of samples from at least each 50 tons of the same shape size,excluding length, from each heat in the as-rolled condition. Piece testing(frequency P) for shapes, means one CVN test set of specimens from at least each15 tons or each single length of 15 tons of the same shape size, excluding length,from each heat in the as-rolled condition.] Heat testing is probably adequate inmost circumstances.

The specimen location required by ASTM A673 is not at the least tough partof a W shape. For a W shape, the volume at the flange web intersection has thelowest ratio of surface area to volume and hence cools the slowest. This slowcooling causes grain growth and reduced toughness. The finer the grain, thetougher the material. Also, ASTM A673 does not specify where in the productrun of an ingot to sample. Impurities tend to rise to the upper portion of the ingotduring cooling from molten metal. Impurities reduce the toughness of thefinished metal. Hence, shapes produced from the upper portions of an ingot canbe expected to have lower toughness, and samples should be taken from shapes


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produced from this portion of the ingot. In the continuous casting process,impurities tend to be more evenly distributed; hence, samples taken anywhereshould suffice. The AISC LRFD specification requires testing from the upperportion of the ingot and near the web flange intersection. Even though the AISCLRFD specification does not require toughness testing for the typical WSMFconnection, i.e., a Group 2 beam to a Group 4 column, it appears that there maybe inadequate through thickness toughness in the Group 4 and 5 column flanges.

In response to concerns raised following the Northridge Earthquake, the AISCconducted a statistical survey of the toughness of material produced in structuralshapes, based on data provided by six producers for a production period ofapproximately one year (American Institute of Steel Construction - 1995). Thissurvey showed a mean value of Charpy V notch toughness for all shape groupsthat was well in excess of 20 ft-lb. at 70 degrees F. However, not all of thesamples upon which these data are based were taken from the core area,recommended by these Interim Guidelines. Consequently, this survey does notprovide definitive information on the extent to which standard material producedby the mills participating in this survey will meet the recommended values.

8.1.5 Lamellar Discontinuities

For critical joints (beam to column CJP welds or other tension applications where Z-axis or tri-axialstress states exist), ultrasonic testing (UT) should be specified for the member loaded in the Z axisdirection, in the area of the connection. A distance 3 inches above and below the location to be weldedto the girder flange is recommended. The test procedure and acceptance criteria given in ASTMA898-91, Standard Specification for Straight Beam Ultrasonic Examination of Rolled Steel StructuralShapes, Level I, should be applied. This testing should be done in the mill or fabrication shop for newconstruction. For repair welding, the same procedure should be applied in the field, as access permits.

Commentary: Very little test data exist on the through thickness properties ofstructural shapes nor are there any standard test methods for determining theseproperties. Nevertheless, the typical beam-column joints in WSMFs rely heavilyon the through-thickness properties of column flanges. Some of the proposedstrengthening and reinforcing solutions will transmit even more forces into the Zaxis of the column flanges. Laminations (pre-existing planes of weakness) andlamellar tearing (cracks parallel to the surface) will impair the Z axis strengthand toughness properties. These defects are mainly caused by non-metallicsulfides and oxides which begin as almost spherical in shape, and becomeelongated in the rolling process. When Z axis loading occurs from weldshrinkage strains or external loading, microscopic cracks may form between thediscrete, elongated nonmetallic inclusions. As they link up, lamellar tearingoccurs.



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Longitudinal wave ultrasonic testing is very effective in mapping seriouslamellar discontinuities. Improved quality steel does not eliminate weldshrinkage and, by itself, will not necessarily avoid lamellar tearing in highlyrestrained joints. Ultrasonic testing should not be specified without due regardfor design and fabrication considerations.

In cases where lamellar defects or tearing are discovered in erection or onexisting buildings, the designer must consider the consequences of making repairsto these areas. Gouging and repair welding will add additional cycles of weldshrinkage to the connection and may promote crack extensions or new lamellartearing. When secondary cracking is discovered, a welding engineer should beconsulted to generate a special WPS for the repair.

8.2 Welding

8.2.1 Welding Process

The welding process to be used to execute the joint weld [e.g. shielded metal (SMAW), flux cored(FCAW), submerged (SAW), gas metal arc weld (GMAW), or electroslag (requires qualification of thewelding procedure specification)] should be specified in the Contract Documents for weld repairs. Contract documents for new construction should state any restrictions on weld parameters orprocesses. Most pre-Northridge production welding was executed using FCAW using a self-shieldingprocess (FCAW-SS). Shielded metal arc welding (stick welding) is often used for damage repairs, intight conditions and in some shop applications.

Commentary: At this time there is no clear evidence that one method can produceuniformly superior welds although poor welds can be produced with any of themethods.

8.2.2 Welding Procedures

Welding should be performed within the parameters established by the electrode manufacturer andthe Welding Procedure Specification (WPS), required under AWS D1.1.

Commentary: For example, the position (if applicable), electrode diameter,amperage or wire feed speed range, voltage range, travel speed range andelectrode stickout (e.g. all passes, 0.072 in. diameter, 248 to 302 amps, 19 to 23volts, 6 to 10 inches/minute travel speed, 170 to 245 inches/minute wire feedspeed, 1/2" to 1" electrode stickout) should be established. This information isgenerally submitted by the fabricator as part of the Welding ProcedureSpecification. Its importance in producing a high quality weld is essential. Thefollowing information is presented to help the engineer understand some of theissues surrounding these parameters.


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The amperage, voltage, travel speed, electrical stickout and wire feed speedare functions of each electrode. If prequalified WPSs are utilized, theseparameters must be in compliance with the AWS D1.1 requirements. For FCAWand SMAW, the parameters required for an individual electrode vary frommanufacturer to manufacturer. Therefore, for these processes, it is essential thatthe fabricator/erector utilize parameters that are within the range ofrecommended operation published by the filler metal manufacturer. Alternately,the fabricator/erector could qualify the welding procedure by test in accordancewith the provisions of AWS D1.1 and base the WPS parameters on the test results.For submerged arc welding, the AWS D1.1 code provides specific amperagelimitations since the solid steel electrodes used by this process operate essentiallythe same regardless of manufacture. The filler metal manufacturer’s guidelineshould supply data on amperage or wire feed speed, voltage, polarity, andelectrical stickout. The guidelines will not, however, include information ontravel speed which is a function of the joint detail. The contractor should select abalanced combination of parameters, including travel speed, that will ensure thatthe code mandated weld-bead sizes (width and height) are not exceeded.

8.2.3 Welding Filler Metals

The current AWS D1.1 requirements should be incorporated as written in the Code. The weldingparameters should be clearly specified using a combination of the Project Specifications, the ProjectDrawings, the Shop Drawings and the welding procedure specifications, as required by AWS D1.1. For welding on ASTM A572 steel, the AWS D1.1 code requires the use of low-hydrogen electrodes. With SMAW welding, a variety of non-low hydrogen electrodes are commercially available. Theseelectrodes are not appropriate for welding on the higher strength steels used in building constructiontoday, although they were popular in the past when lower strength steels were employed. All of theelectrodes that are employed for flux cored arc welding (both gas shielded and self-shielded), as well assubmerged arc welding, are considered low hydrogen.

For critical joints (beam to column CJP welds or other tension applications where Z-axis loading ortri-axial stress states exist), toughness requirements for the filler metals should be specified. Aminimum CVN value of 20 ft.-lb. at a temperature of 0 degrees F. should be required, unless morestringent requirements are indicated by the service conditions and/or the Contract Documents. Thefiller metal should be tested in accordance with the AWS A5 filler metal specification to ensure it iscapable of achieving this level of notch toughness. The filler metal manufacturers Typical Certificate ofConformance, or a suitably documented test performed by the contractor, should be used to documentthe suitability of the electrode used. These tests should be performed for each filler metal by AWSclassification, filler metal manufacturer and filler metal manufacturer’s trade name. The sizes asspecified by the AWS A5 document should be tested, although the exact diameter used in productionneed not be specifically tested. This requirement should not be construed to imply lot or heat testing offiller metals.



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Electrode specification sheets should be provided by the Fabricator/Erector prior to commencingfabrication/erection.

Commentary: Currently, there are no notch toughness requirements for weldmetal used in welding ASTM A 36 or A 572, Grade 50, steel in AWS D1.1. Thistopic has been extensively discussed by the Welding Group at the JointSAC/AISC/AISI/NIST Invitational Workshop on September 8 and 9, 1994, and byall participants of the SAC Invitational Workshop on October 28 and 29, 1994.The topic was also considered by the AWS Presidential Task Group, whichdecided that additional research was required to determine the need fortoughness in weld metal. There is general agreement that adding a toughnessrequirement for filler metal would be desirable and easily achievable. Most fillermetals are fairly tough, but some will not achieve even a modest requirement suchas 5 ft-lb. at + 70°F. What is not in unanimous agreement is what level oftoughness should be required. The recommendation from the Joint Workshop was 20 ft-lb. at -20°F per Charpy V-Notch [CVN] testing. The recommendationfrom the SAC Workshop was 20 ft-lb. at 30°F lower than the Lowest AmbientService Temperature (LAST) and not above 0°F. The AWS Presidential TaskGroup provided an interim recommendation for different toughness valuesdepending on the climatic zone, referenced to ASTM A709. Specifically, therecommendation was for 20 ft-lb. at temperatures of 70 degrees F for Zone 1, 40degrees F for Zone 2, and 10 degrees F for Zone 3. The AWS also suggestedtoughness values for base metals used in these applications.

Some fractured surfaces in the Northridge and Kobe Earthquakes revealedevidence of improper use of electrodes and welding procedures. Prominentamong the misuses were high production deposition rates. Pass widths of up to 1-1/2 inches and pass heights of 1/2 inch were common. The kind of heat inputassociated with such large passes promotes grain growth in the HAZ andattendant low notch toughness. Root gap, access capability, electrode diameter,stick-out, pass thickness, pass width, travel speed, wire feed rate, current andvoltage were found to be the significant problems in evaluation of welds inbuildings affected by the Northridge earthquake.

Welding electrodes for common welding processes include:

AWS A5.20: Carbon Steel Electrodes for FCAWAWS A5.29: Low Alloy Steel Electrodes for FCAWAWS A5.1: Carbon Steel Electrodes for SMAWAWS A5.5: Low Alloy Steel Covered Arc Welding Electrodes (for SMAW)AWS A5.17: Carbon Steel Electrodes and Fluxes for SAWAWS A5.23: Low Alloy Steel Electrodes and Fluxes for SAWAWS A5.25: Carbon and Low Alloy Steel Electrodes and Fluxes for Electroslag

Welding


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In flux cored arc welding, one would expect the use of electrodes that meeteither AWS A5.20 or AWS A5.29 provided they meet the toughness requirementsspecified below.

Except to the extent that one requires Charpy V-Notch toughness andminimum yield strength, the filler metal classification is typically selected by theFabricator. Compatibility between different filler metals must be confirmed bythe Fabricator, particularly when SMAW and FCAW-SS processes are mixed. Generally speaking, SMAW-type filler metals may not be applied to FCAW-SStype filler metals (e.g. when a weld has been partially removed) while FCAW-typefiller metals may be applied to SMAW-type filler metals. This recommendationconsiders the use of aluminum as a killing agent in FCAW-SS electrodes that canbe incorporated into the SMAW filler metal with a reduction in impact toughnessproperties.

As an aid to the engineer, the following interpretation of filler metalclassifications is provided below:

E1X2X3T4X5 For electrodes specified under AWS A5.20E1X2X3T4X5X6 For electrodes specified under AWS A5.29E1XX7X8X9X10 For electrodes specified under AWS A5.1 or AWS A5.5.

NOTES:

1. Indicates an electrode.

2. Indicates minimum tensile strength of deposited weld metal (in tens of ksi, e.g., 7 = 70ksi).

3. Indicates primary welding position for which the electrode is designed (0 = flat andhorizontal and 1 = all positions).

4. Indicates a flux cored electrode. Absence of a letter indicates a "stick" electrode forSMAW.

5. Describes usability and performance capabilities. For our purposes, it conveys whetheror not Charpy V-Notch toughness is required (1, 5, 6 and 8 have impact strengthrequirements while 2, 4, 7, 10 and 11 do not). A "G" signifies that the properties are notdefined by AWS and are to be agreed upon between the manufacturer and the specifier. Impact strength is specified in terms of the number of foot-pounds at a given temperature(e.g., 20 ft-lb. at 0 degrees F). Note that for electrodes specified under AWS A5.20, theformat for usage is "T-X".



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6. Designates the chemical composition of deposited metal for electrodes specified underAWS A5.29. Note that there is no equivalent format for chemical composition forelectrodes specified under AWS A5.20.

7. The first two digits (or three digits in a five digit number) designate the minimum tensilestrength in ksi.

8. The third digit (or fourth digit in a five digit number) indicates the primary weldingposition for which the electrode is designed (1 = all positions, 2 = flat position and filletwelds in the horizontal position, 4 = vertical welding with downward progression and forother positions.)

9. The last two digits, taken together, indicate the type of current with which the electrodecan be used and the type of covering on the electrode.

10. Indicates a suffix (e.g., A1, A2, B1, etc.) designating the chemical composition of thedeposited metal.

Electrode Diameter: (See AWS D1.1 Section 4.14.1.2) The issue of maximumelectrode diameter has not been studied sufficiently to determine whether or notelectrode diameter is a critical variable. Recent tests have produced modifiedframe joints with acceptable test results using the previous standard-of-practice 0.120 in. diameter wire. The use of smaller diameter electrodes will slow the rateof deposition (as measured by volume) but will not, in and of itself, produce anacceptable weld. The following lists the maximum allowable electrode diametersfor prequalified FCAW WPS’s according to D1.1:

• Horizontal, complete or partial penetration welds: 1/8 inch (0.125")*• Vertical, complete or partial penetration welds: 5/64 inch (0.078")• Horizontal, fillet welds: 1/8 inch (0.125")• Vertical, fillet welds: 5/64 inch (0.078")• Overhead, reinforcing fillet welds: 5/64 inch (0.078")

* This value is not part of D1.1-94, but will be part of D1.1-96.

For a given electrode diameter, there is an optimum range of weld bead sizesthat may be deposited. Weld bead sizes that are outside the acceptable size range(either too large or too small) may result in unacceptable weld quality. The D1.1code controls both maximum electrode diameters and maximum bead sizes (widthand thickness). Prequalified WPS’s are required to meet these coderequirements. Further restrictions on suitable electrode diameters are notrecommended.


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8.2.4 Preheat and Interpass Temperatures

The preheat temperatures and conditions given in AWS D1.1, Chapter 4 should be strictlyobserved with special attention given to Section 4.2, for the thickness of metal to be welded. Forrepair welding of earthquake damage, the AASHTO/AWS D1.5 Bridge Welding Code preheatrequirements for fracture-critical, non-redundant applications should be considered.

Cracking of welds and heat affected zones should be avoided. One type of weld cracking ishydrogen induced cracking (HIC). For a given steel, variables that reduce HIC tendencies areprioritized as follows:

1. Lower levels of hydrogen.

2. Higher preheat and interpass temperatures.

3. Postheat.

4. Retarded cooling (insulating blankets).

Only low hydrogen electrodes should be used for fabrication and/or repair of seismically loadedstructures. Proper preheat and interpass temperatures should be maintained. AWS D1.1 requirementsare generally adequate for new construction. Highly restrained repair welds may require higher preheatlevels.

Control of hydrogen and proper preheat and interpass temperature is much more powerful forovercoming HIC than postheat or retarded cooling methods. Retarded cooling has limited benefit ifthe entire piece is not preheated - obviously impractical for structural applications.

The engineer is encouraged to emphasize proper preheat and the use of low hydrogen electrodesand practice. If these measures are insufficient to prevent cracking, a welding engineer should beconsulted to determine appropriate measures that should be incorporated to eliminate cracking. Thesemeasures may or may not call for additional preheat, postheat, or retarded cooling.

While low hydrogen electrodes and proper preheat is essential, postheat and retarded cooing is notgenerally required and should not be used for routine construction or repair.

Commentary: There are two primary purposes for preheating and interpasstemperature requirements:

(1) To drive off any surface moisture or condensation which may be presenton the steel so as to lessen the possibility of hydrogen being introduced into theweld metal and HAZ, and

(2) To prevent the steel mass surrounding the weld from quenching the HAZas cooling occurs after welding.



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Virtually all weld repairs are made under conditions of high restraint. Consequently, higher preheat/interpass temperatures may be required for repairapplications. As steel is cooled from the austenitic range (above about 1330degrees F), it goes through a critical transition temperature. If it goes throughthat temperature range too fast, a hard, brittle phase called martensite forms(quenching). If it passes through that temperature range at a slower rate, ductile,tougher phases called bainite or ferrite/pearlite form. Preheating of thesurrounding mass provides a slower cooling rate for the weld metal and HAZ.

The American Association of State Highway and Transportation Officials(AASHTO) recognizes repair welding as more critical in its guidelines for therepair of fracture-critical bridge members. The purpose, in part, is to allow moreplastic flow and yielding, at welding temperatures, in the area near the weld. Therequirements are given in Table 8-6:

Table 8-6 - AASHTO Preheat Requirements for Fracture Critical Repairs1

Steel Thickness, in. Minimum Preheat/InterpassTemp., °F

A36/A572 to 1-1/2 325A36/A572 >1-1/2 375

1- Reference AASHTO/AWS D1.5-95 Bridge Welding Code

Preheat temperatures should be measured at a distance from the weld equalto the thickness of the part being welded, but not less than three inches, in anydirection including the through thickness of the piece. Where plates are ofdifferent thicknesses, the pre-heat requirement for the thicker plate shouldgovern. Maintenance of these temperatures through the execution of the weld(i.e. the interpass temperature) is essential. Maximum interpass temperaturesshould be limited to 550 degrees F for prequalified WPSs, for fracture-criticalapplications. Higher interpass temperatures could be employed if those highertemperature limits are qualified by test.

8.2.5 Postheat

Postheat is the application of heat in the 400 degrees F to 600 degrees F range after completion ofwelding. It may be helpful in mitigating some cracking tendencies.

Commentary: A postheat specification might require that complete jointpenetration groove welds in existing buildings be postheated at 450 degrees F fortwo hours. The purpose of this postheat is to accelerate the removal of hydrogenfrom the weld metal and HAZ and reduce the probability of cracking due tohydrogen embrittlement. Hydrogen will migrate within the weld metal atapproximately 1 inch per hour at 450 degrees F, and at about 1 inch per month at70 degrees F. To the extent that hydrogen embrittlement is of concern, postheat


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is one method of mitigating cracking. The use of low hydrogen electrodes, properwelding procedures, and uniformly applied and maintained preheat mayrepresent a cost-effective method of addressing the problem of hydrogenembrittlement in lieu of postheat.

When postheat is required, AASHTO/AWS D1.5-95 specifications require thisto be done immediately upon completion of welding. The postheat is between 400to 500 degrees F for one hour minimum, for each inch of the thickest member orfor two hours, whichever is less.

8.2.6 Controlled Cooling

Most of the weldment cooling is effected by conductance within the steel rather than radiation. Retarded cooling should only be specified in cases where large weldments subject to significant residualstresses due to restraint (e.g. multiple members framing into one connection with Z axis loading) orambient temperatures that would result in rapid cooling of large weldments. The length of time to cooldown the weld and the level of insulation required are a function of weldment temperature, thickness ofbase metal and ambient temperature.

Commentary: Active systems of ramp-down cooling are generally not required;however, in highly restrained conditions they may offer an added advantage.

8.2.7 Metallurgical Stress Risers

Metallurgical discontinuities such as tack welds, air-arc gouging and flame cutting withoutpreheating or incorporation into the final weld should not be permitted. Inadvertent damage of thistype should be repaired by methods approved by the engineer, following the AWS D1.1 criteria and aspecific WPS covering repairs of this type.

Commentary: Metallurgical stress risers may result from tack welds, air-arcgouging and flame cutting performed without adequate preheat. However,preheating is not necessarily required for air arc gouging or flame cutting used inthe preparation of a surface to receive later welding. The subsequent heat inputduring the welding process should adequately anneal the affected area. The AWSD1.1 code requires the same preheating for tack welding operations as normalwelding, with the exception of tack welds that are incorporated into subsequentsubmerged arc weld deposits.

Arc strikes can also be a source of metallurgical stress risers and should notbe indiscriminately made. AWS D1.1 Section 3.10 indicates that “arc strikesoutside the area of permanent welds should be avoided on any base metal. Cracks or blemishes caused by arc strikes should be ground to a smooth contourand checked to ensure soundness.”



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8.2.8 Welding Preparation & Fit-up

Any cracked columns, welds, or beam flanges should be prepared to receive the weldingcontemplated by the engineer. AWS D1.1 provides guidance on the precise nature of the fit-uprequirements and tolerances.


Interim Guidelines:Evaluation, Repair, Modification andDesign of Steel Moment Frames Chapter 9 - Quality Control/Quality Assurance

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9. QUALITY CONTROL/QUALITY ASSURANCE

Quality control is principally the responsibility of the contractor, while Quality Assurance isperformed at the prerogative of the owner and as mandated by the Building Code. Key parts ofthe Quality Control program include assuring that all parties understand what is to be constructed,and the standards that apply. All workers and inspectors should be adequately qualified toperform the required work, and should have written procedures, approved by the engineer, for thework that is to be performed.

9.1 Quality Control

Fabrication/erection inspection and testing should be the responsibility of the contractor,unless otherwise provided for in the Contract Documents.

9.1.1 General

A pre-job meeting or series of meetings should be held with the owner's representatives, theengineer, the Fabricator/Erector's production and QC personnel to plan and discuss the projectand fabrication procedures. Welders and welding operators should also be involved at some level,either by a meeting or direct dissemination of the information. Fabrication/erection inspection andtesting should be performed prior to assembly, during assembly, fit-up, tacking, welding and afterwelding to ensure that the materials and workmanship meet the requirements of the ContractDocuments. The fitters and welders should have the applicable WPS document and drawings foreach connection and joint at their assembly station.


Inspectors responsible for acceptance or rejection of materials and workmanship should bequalified in accordance with Sections 10 and 11 of these Guidelines. The engineer should havethe authority and duty to verify the qualifications of the inspectors.

9.1.3 Duties

The inspector should ascertain that all materials comply with the Contract Documents, eitherby mill certifications or testing. The inspector should verify that all fabrication and erectionwelding is performed in accordance with the Contract Documents. Detailed duties are furtherdescribed in Section 10 of these Guidelines.

9.1.4 Records

The QC inspector should insure that each welder has a unique identification mark or die stampto identify his or her welds. The inspector should also mark the welds/parts/ joints that have beeninspected and accepted with a distinguishing mark or die stamp, or alternatively, maintain records


Chapter 9 - Quality Control/Quality Assurance Design of Steel Moment Frames

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indicating the specific welds inspected by each person. The NDT technician should use the weldidentification system given in AWS D1.1, Sections 6.19.1 and 6.19.2. The inspector should keepa record of all welders, welding operators and tack welders; all procedure and operatorqualifications; all accepted parts; the status of all rejected joints; NDT test reports; and other suchinformation as may be required.

9.1.5 Engineer Obligations

The structural engineer or designated welding engineer should perform a review of theFabricator/Erector’s Quality Control program, equipment condition, and availability of equipmentand qualified personnel. The review should include the following:

a) Interview with Fabricator/Erector’s designated Quality Control personnel.

b) Review of Fabricator/Erector’s written quality procedure manual.

c) Review of Fabricator/Erector’s Procedure Qualification Records (PQR’s) and WPSapplicable to the specific project.

d) Review of Welder Performance Records.

e) Review of the Fabricator/Erector’s NDT procedures, equipment calibration records,and personnel training records. Alternatively, the Fabricator/Erector may contract withan outside Quality Control company for NDT services; however, this should not takethe place of the owner’s QA responsibility for NDT.

f) Designate any specific NDT requirements which apply to the project and which arebeyond those required by the Code.

g) A meeting with the owner’s representative, fabricator/erector’s Quality Controlpersonnel and the welder, to review the WPS.

9.1.6 Contractor Obligations

The contractor should make available to the inspector and NDT Technician all drawings,project specifications, mill certifications, welder qualifications, WPSs and PQRs applicable to theproject. The contractor should cooperate fully with requests from inspection and testingpersonnel for access to the connections and joints to be inspected or tested. This includes beamand column turning in the shop, weld backing removal and access platforms or scaffolding asrequired to perform the work safely. The contractor should be responsible for all necessarycorrections of deficiencies in materials and workmanship. The contractor should comply with allrequests of the inspector to correct deficiencies. The NDT Technician should be apprised of anyrepairs made by the contractor. Inspections should be performed in a timely manner. Disputesshould be resolved by the structural engineer of record, or by a welding engineer.


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9.1.7 Extent of Testing

Information furnished to the bidding contractors should clearly identify the extent ofinspection and testing to be performed by the contractor. Weld joints requiring testing byContract Documents shall be tested for their full length, unless partial or spot testing is specified. When partial or spot testing is specified, the location and lengths of welds or categories of weld tobe tested should be clearly designated in the Contract Documents. Each spot test should cover atleast 4 inches of the weld length. When spot testing reveals indications of rejectablediscontinuities that require repair, the extent of those discontinuities should be explored. Twoadditional spots in the same segment of weld joint should be taken at locations away from theoriginal spot. When either of the two additional spots show defects that require repair, the entiresegment of weld represented by the original weld should be completely tested.

Where spot testing or percentage sampling is specified on certain welds, the contract drawingsand shop drawings should so state using NDT symbols in conjunction with the welding symbols. On projects where a sliding sampling scale is specified, based on the UT reject level of individualwelders, the inspector should keep records on each welder or welding operator. These recordswill be used as a basis for sampling rate reduction.

Commentary: AWS D1.1 uses the term "Fabrication/Erection Inspection"synonymously with the classical "Quality Control" function of other industries. Abasic premise of Quality Control is to have the production, engineering andQuality Control departments independent of one another.

The contractor should be responsible for establishing the Quality Controlprogram and for in-progress Quality Control of work. Part of this effort is torequire that welders meet established minimum requirements. Execution ofcritical welds requires skilled welders who will follow the project weldingrequirements. An important part of any Quality Control program is assuringthat the workers have the appropriate qualifications to perform the work. Weldsexecuted by welders who do not satisfy the welder performance qualificationsshould be considered rejectable. Important aspects of a QC program shouldinclude as a minimum:

1. Welders shall be qualified for the work they will be doing per AWS D1.1, Section 5,Part C.

2. The qualifications of each welder should be certified by an appropriate authority andverified by the contractor and Special Inspector. The engineer should establishwhether there are certifications from selected jurisdictions that will or will not beaccepted as acceptable substitutions.

The Quality Control function of the contractor should be isolated from theproduction department and the QC Manager should report directly to a high levelcompany officer to avoid conflicts of interest with production.



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9.2 Quality Assurance & Special Inspection

Verification inspection and testing should be the responsibility of the owner and/or theengineer unless otherwise provided for in the Contract Documents. The Quality Assurancedesignate should act for and in behalf of the owner or engineer on all inspection, NDT and qualitymatters that are within the scope of the Contract Documents.

9.2.1 General

Verification inspection and testing are the prerogatives of the owner who may perform thisfunction or, when provided for in the Contract Documents, waive independent verification, orstipulate that both inspection and verification shall be performed by the contractor. Inmunicipalities that have adopted the UBC, verification inspection and testing is mandated forstructural welding, and is designated as “Special Inspection.”

The QA inspector should be included in the pre-job meetings for fabrication and erectiondiscussions referenced in 9.1.1. Fabrication/erection verification inspection and testing should beperformed concurrently with the Quality Control inspection and testing to ensure that thecontractor's QC program is meeting the requirements of the Contract Documents. The QAinspector should ensure that the fitters and welders have the applicable WPS document andrequired information for each connection and joint at their assembly station.


Inspectors responsible for acceptance or rejection of materials and workmanship should bequalified in accordance with Chapters 10 and 11 of these Interim Guidelines. The engineer shouldhave the authority and duty to verify the qualifications of the inspectors. The inspector may useassistants who are formally designated, aware of their assigned responsibility and the acceptancecriteria, and work under the direct supervision and monitoring of a qualified inspector.

9.2.3 Duties

The QA inspector should verify the qualifications of the QC inspectors and the NDTtechnicians. The inspector should verify that the mill certifications for all materials are beingchecked by the QC inspector and that they comply with the Contract Documents. The inspectorshould verify that all fabrication and erection welding is performed in accordance with theContract Documents. Detailed duties are further described in Chapter 10 of these InterimGuidelines.

9.2.4 Records

The inspector should ensure that each welder, NDT technician and QC inspector has a uniqueidentification mark or die stamp to identify his or her welds/weld tests/weld inspections. The QAInspector should ensure that the QA and NDT personnel are keeping the proper records of allwelders, welding operators and tack welders; all procedure and operator qualifications; all


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accepted parts; the status of all rejected joints; NDT test reports; and other such information asmay be required.

9.2.5 Engineer Obligations

The structural engineer or designated welding engineer should perform a complete review ofthe QA Agency. This review should encompass personnel qualification, written proceduresmanual, and availability of equipment and qualified personnel. The Agency should employ anAmerican Society for Nondestructive Testing (ASNT) Level III qualified person who overseesequipment calibration and personnel certification and training for the project on a full time basis. Reviews should be performed in a timely manner. Disputes should be resolved by a qualifiedwelding engineer.

9.2.6 Contractor Obligations

The contractor should make available to the QA Inspector and QA NDT Technician (ifapplicable) all drawings, project specifications, mill certifications, welder qualifications, WPSs andPQRs applicable to the project. The contractor should cooperate fully with requests frominspection and testing personnel for access to the connections and joints to be inspected or tested. This includes beam and column turning in the shop, weld backing removal and access platformsor scaffolding as required to perform the work safely. The contractor should be responsible forall necessary corrections of deficiencies in materials and workmanship. The contractor shouldcomply with all requests of the QA Inspector to correct deficiencies. The QA NDT Technicianshould be apprised of any repairs made by the contractor.

9.2.7 Extent of QA Testing

The QA representative may perform independent inspecting and testing to the extentestablished in the contract documents. When conditions exist that make further testing advisable,the QA representative, with the concurrence of the structural engineer of record, may performadditional independent inspection and testing, to the degree his/her judgment suggests asappropriate. Acceptance criteria should be mutually agreeable to the inspector and contractor. Discrepancies between the QC and QA decisions should be resolved by the engineer.

Commentary: AWS D1.1 uses the term "Verification Inspection" synonymouslywith the "Quality Assurance" function of other industries. The purpose of QAprograms is to provide an oversight to the contractor's QC program. This mayrange from simple records/report reviews to a full testing and inspectionprogram, depending on the effectiveness of the Fabricator/Erector's QC program,and the requirements of the building code. Often this cannot be established untilthe contractor is selected.

The owner must ensure that an adequate Quality Control program is in place,and is responsible for the Quality Assurance program. The use of “licensed” or



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“qualified” fabrication shops in lieu of requiring independent Quality Assuranceprovided by the owner is not recommended. However, a fabrication shop that islicensed or qualified by a recognized program, such as the AISC QualityCertification Program, does provide a minimum assurance of capability of goodperformance.

The owner is responsible for establishing the Quality Assurance program. Elements in an acceptable Quality Assurance program should conform to thoserequired by the UBC.

Since most owners have little expertise or knowledge related to construction,this often means that the engineer must advise the owner, and, in many cases,establish the program. Example Quality Assurance requirements might includethe following:

1. The lead welding inspector should be a Certified Welding Inspector (CWI) per AWS-QC-1 Standards, and, where applicable, should be certified by the responsiblejurisdiction as a qualified inspector for structural steel welding. Other weldinginspectors performing visual inspection under the supervision of the lead weldinginspector should hold an active and appropriate certification. Not more than fournon-CWIs should be under the supervision of a CWI.

2. All welding should be inspected visually as required by AWS D1.1 (See AWS D1.1Section 8.15.1).

3. All complete and partial joint penetration welds should be inspected ultrasonically asrequired by AWS D1.1 (See AWS D1.1 Section 8.15.4) after the weld is completed andhas cooled down. The inspector and NDT technician should perform the followingtasks for each weld.

a. Verify material identification per approved shop drawings and specifications.

b. Perform a UT lamination check of the column and beam as required by AWSD1.1 or at least within a 6 in. radius around the weld. As a minimum thischeck should be performed after welding, however, if performed beforewelding as well, this may save some rework effort.

c. Verify that an approved welding procedure specification (WPS) has beenprovided and that the WPS has been reviewed with each welder performingthe weld. A copy of the appropriate WPSs should be at each joint. Welds notexecuted in conformance with the WPS should be considered rejectable (SeeAWS D1.1 Section 6.3.1).

d. Identify welding consumables per approved shop drawings and approvedWPS (See AWS D1.1 Sections 6.2 and 6.5.3).

e. Verify welder identification and certification. Verify that requiredsupplemental qualification tests have been passed (See AWS D1.1 Section 6.4)and mock-ups, if required by the Contract Documents, have been executed.


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f. Verify proper amperage and voltage of the welding process by using a handheld calibrated amp and volt meter. (Similar equipment should also be usedby the fabricator.) Amperage and voltage should be measured at the arc withthis equipment.

g. Visually inspect all required welds in accordance with AWS D1.1. Verify anddocument the fabrication sequence including the following per approved shopdrawings and approved WPS (See AWS D1.1 Section 6.5.4):

1. Fit-up;

2. Preheat and interpass temperatures;

3. Welding machine settings. Voltage should be determined at the arcand amperage on the cables. Welds executed outside of theparameters contained in the approved WPS should be consideredrejectable;

4. Weld sequence;

5. Weld pass sequence and size of weld bead;

6. Peening, if required;

7. Removal of backup and weld (extension) tabs, preparatory grindingand cleaning, and execution of reinforcing fillet weld, as required bythe WPS;

8. Application and maintenance of postheat or insulation to completedweld as required by the WPS.

h. Ultrasonically inspect in accordance with AWS D1.1. Attempt to pass soundthrough the entire weld volume from two crossing directions where possible. In particular, inspect the beam bottom flange from both "A" and "B" faces. This will require adequate staging to be provided by the contractor to permitsafe access by the inspector. This is normally not a problem in existingbuildings; however, it may be more difficult on buildings under construction.



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Interim Guidelines:Evaluation, Repair, Modification andDesign of Steel Moment Frames Chapter 10 - Visual Inspection

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10. VISUAL INSPECTION

Visual inspection is the primary method which should be used to confirm that the procedures,materials and workmanship incorporated in the Work are those that have been specified andapproved for the project. Visual inspection should be conducted by appropriately qualifiedpersonnel, in accordance with a written practice.

10.1 Personnel Qualification

Visual inspection personnel should be qualified under AWS D1.1, Chapter 6. The basis ofqualification should be specified by the Engineer. Acceptable qualification bases are :

a) Current or previous certification as an AWS Certified Welding Inspector (CWI) inaccordance with the provisions of AWS QC1, Standard and Guide forQualification and Certification of Welding Inspectors, or

b) Current or previous qualification by the Canadian Welding Bureau (CWB) to therequirements of the Canadian Standard Association (CSA) Standard W178.2,Certification of Welding Inspectors, or

c) An engineer or technician who, by training or experience, or both, in metals fabrication,inspection and testing, is competent to perform inspection work.

The qualification of an inspector will remain in effect indefinitely, provided the inspectorremains active in the inspection of welded steel fabrication, unless there is a specific reason toquestion the inspector's ability.

The Engineer should have the authority to verify the qualification of inspectors.

10.2 Written Practice

a) The employer (Testing Agency or Fabricator/Erector) should maintain a writtenpractice for the control and administration of inspection personnel training andqualification.

b) The written practice should describe the employer's procedures for visual weldinginspection and material controls for determining the acceptability of materials andweldments in accordance with the applicable codes, standards, specifications andprocedures.

c) The employer's written practice should describe the training and experience andrequirements for qualification.


Chapter 10 - Visual Inspection Design of Steel Moment Frames

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10.3 Duties

a) The inspector should review and understand the applicable portions of theSpecifications, the Contract Drawings and the Shop Drawings for the project.

b) The inspector should verify that all applicable welding Procedure Qualification Records(PQR)s, welder and welding operator qualifications and welding procedurespecifications (WPS) are available, current and accurate.

c) The inspector should require requalification of any welder, welding operator or tackwelder who has, for a period of six months, not used the process for which the personwas qualified.

d) The inspector should check all mill certificates for material compliance with the projectrequirements.

e) The inspector should verify the electrode/wire specification sheets for compliance withthe Contract Documents.

f) The inspector should make certain that all electrodes are used only in the positions andwith the type of welding parameters specified in the WPS.

g) The inspector should, at suitable intervals, observe joint preparation, assembly practice,preheat temperatures, interpass temperatures, welding techniques, welder performanceand post-weld dressing to make certain that the applicable requirements of the WPSand Code are met.

h) The inspector should inspect the work to ensure compliance with AWS D1.1, Sections3 and 8.15. Size and contour of welds should be measured with suitable gauges. Visual inspection may be aided by a strong light, magnifiers, or other devices whichmay be helpful.

i) The QC inspector should be responsible for scheduling the NDT technicians in a timelymanner, after the visual inspection is complete and the assembly has cooled. For repairwelding, the NDT should not be performed sooner than 48 hours after the welding iscomplete and cooled to ambient temperature.

j) Inspectors should identify the inspected and accepted welds, assemblies andconnections with a personal mark or stamp, or maintain adequate records to indicatethe status of inspection work. The accepted and rejected items should be documentedin a written report. The report should be transmitted to the designated recipients in atimely manner.

Commentary: Depending on how the QA and QC functions are structured for anyparticular project, the role of the visual inspector may vary considerably. Ideally, the QC inspector is an employee of the contractor and answers to a QA

Interim Guidelines:Evaluation, Repair, Modification andDesign of Steel Moment Frames Chapter 10 - Visual Inspection

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department head who is not connected with production. If this is not the case, aninherent conflict of interest may be present. The level of involvement of the QAagency is highly dependent on the structure of the contractor's QC program. Ifthe contractor's QC program is well organized, has competent inspection andtesting personnel and is truly independent of production, the QA function canoperate in the classical manner as an overseer wherein random spot inspectionand testing suffice. In the opposite case where the QC department is being run byproduction, the QA agency must take a very active role and perform many of theQC duties.

The definitions of these roles can directly affect the project structure andassociated budgets. The Owner cannot accurately budget for QA testing andinspection until the contractor is selected and the QC program established. Alleviating this dilemma requires the designer to tightly specify the QC and QAprograms.

Although AWS D1.1 allows inspector qualification without the CWIcertification under the QC1 criteria, it is strongly recommended that theinspection personnel be CWI certified (or previously certified), by experience andwritten examination.


Chapter 10 - Visual Inspection Design of Steel Moment Frames

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Interim Guidelines:Evaluation, Repair, Modification andDesign of Steel Moment Frames Chapter 11 - Nondestructive Testing

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11. NONDESTRUCTIVE TESTING

Nondestructive testing includes magnetic particle testing (MT), Liquid Dye Penetrant testing(PT), Radiographic Testing (RT) and Ultrasonic Testing (UT). The purpose of nondestructivetesting is to serve as a backup to Visual Inspection and to detect flaws and defects that are notvisible. Nondestructive examination is not a replacement for an adequate program of VisualInspection, and should not be used as such.

11.1 Personnel

11.1.1 Qualification

Nondestructive testing personnel shall be qualified under The American Society forNondestructive Testing, Inc., Recommended Practice No. SNT-TC-1A, in one of the threefollowing levels:

a) NDT Level I - An NDT Level I individual should be qualified to properly performspecific calibrations, specific NDT, and specific evaluations for acceptance or rejectiondeterminations according to written instructions and to record results. The NDT LevelI should receive the necessary instruction or supervision from a certified NDT Level IIIindividual or designee.

b) NDT Level II - An NDT Level II individual should be qualified to set up and calibrateequipment and to interpret and evaluate results with respect to applicable codes,standards and specifications. The NDT Level II should be thoroughly familiar with thescope and limitations of the methods for which he/she is qualified and should exerciseassigned responsibility for on-the-job training and guidance of trainees and NDT LevelI personnel. The NDT Level II should be able to organize and report the results ofNDT.

c) NDT Level III - An NDT Level III individual should be capable of establishingtechniques and procedures; interpreting codes, standards, specifications andprocedures; and designating the particular NDT methods, techniques, and proceduresto be used. The NDT Level III should be responsible for the NDT operations forwhich he/she is qualified and assigned and should be capable of interpreting andevaluating results in terms of existing codes, standards, and specifications. The NDTLevel III should have sufficient practical background in applicable materials,fabrication, and product technology to establish techniques and to assist in establishingacceptance criteria when none are otherwise available. The NDT Level III should havegeneral familiarity with other appropriate NDT methods, as demonstrated by theASNT Level III Basic examination or other means. The NDT Level III, in themethods in which certified, should be capable of training and examining NDT Level Iand II personnel for certification in those methods.

Interim Guidelines:Evaluation, Repair, Modification andChapter 11 - Nondestructive Testing Design of Steel Moment Frames

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11.1.2 Written Practice

a) The employer (Testing Agency or Fabricator/Erector) should maintain a writtenpractice for the control and administration of NDT personnel training, examination andcertification.

b) The employer's written practice should describe the responsibility of each level ofcertification for determining the acceptability of materials and weldments in accordancewith the applicable codes, standards, specifications and procedures.

c) The employer's written practice should describe the training, experience andexamination requirements for each level of certification.

11.1.3 Certification

a) Certification of all levels of NDT personnel is the responsibility of the employer.

b) Certification of NDT personnel should be based on demonstration of satisfactoryqualification in accordance with Sections 6, 7 and 8 of SNT-TC-1A, as modified by theemployer's written practice.

c) Personnel certifications should be maintained on file by the employer and a copy shouldbe carried by the technician.

11.1.4 Recertification

a) All levels of NDT Personnel should be recertified periodically in accordance with oneof the following criteria:

i) Evidence of continuing satisfactory performance

ii) Reexamination in those portions of the examinations in Section 8 deemednecessary by the employer's NDT Level III

b) Recommended maximum recertification intervals are:

i) Levels I and II - 3 years

ii) Level III - 5 years

c) The employer's written practice should include rules covering the duration ofinterrupted service that requires reexamination and recertification.

11.2 Execution

11.2.1 General

Nondestructive testing should not be used in lieu of visual inspection.

Commentary: Visual inspection and NDT should be used as a complement to oneanother. There are four basic testing methods beyond visual inspection which arecommonly used: magnetic particle (MT), liquid penetrant (PT), radiographic


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testing (RT) and ultrasonic testing (UT). The uses of the methods are describedin detail in AWS B1.0, Guide for Nondestructive Inspection of Welds.

When nondestructive testing other than visual is to be required, it should beso stated in the bid documents. This information should designate the categoriesof welds to be examined, the extent of examination in each category and themethods of testing.

The designer should require that the testing laboratory employing the NDTtechnicians be certified by the National Institute of Standards and Technology,NAVLAP program and that the technicians are qualified under ASTM E543. Additionally, the laboratory should employ a Level III NDT supervisor under therequirements of SNT-TC-1A.

The designer or his/her designated welding engineer should be familiar withthe strengths and limitations of each NDT method. Incorrect selection of themethods has caused false reliance on the results. Each method has its ownstrengths and weaknesses. Magnetic particle and liquid penetrant testing requirethe least amount of training; radiographic and ultrasonic testing require a higherlevel of training and background. NDT technicians are not generally required tobe certified welding inspectors under the QC1 requirements; however, it is highlyrecommended that at least one NDT technician active on the project site be soqualified.

11.2.2 Magnetic Particle Testing (MT)

MT may be used for surface and near-surface linear defect flaw detection. It is essential thatfor linear indications to respond to MT, they must be oriented at an angle between 45o and 90o,with the maximum influence occurring at 90o to the flux field. Therefore each area tested shouldhave the electromagnetic yoke positioned at 0o then at 90o.

Commentary: MT’s depth limitation is less than 1/8 inch for typical flaws. Theinstrument consists of an electro-magnetic yoke which sets up a magnetic fluxfield around a weld. A very fine magnetic powder dust is applied to the areabeing tested. As the flux lines cross a linear defect the field is interrupted and thepowder aligns with the defect. Spurious indications are sometimes encounteredalong areas of poor weld bead contour, undercut or overlap. The use of a whitebackground paint to improve contrast can improve the reliability of this method.

A key use of this method is during air-arc gouging to determine if a crack hasbeen totally removed. Root pass testing is also commonly done with MT. Thesetests, of course, require that the NDT technician be continually present duringwelding.


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11.2.3 Liquid Penetrant Testing (PT)

PT may be used to locate defects which are open to the surface.

Commentary: In PT, a highly fluid, red dye penetrant is sprayed on the surfaceof the joint and allowed to soak into any open surface defect by gravity andcapillary action. The surface is then wiped clean and a white developer with apowder consistency is applied. The red dye bleeds back out of the defecthighlighting the flaw. The method is typically used on completed welds.

Due to the problems associated with additional surface preparations and thetime involved with PT, it is recommended that MT be applied when ever possible. There may be situations where, because of geometrical conditions or restrictedaccess, MT cannot be performed. PT is an allowable option keeping in mind thatadditional surface preparation may be necessary.

11.2.4 Radiographic Testing (RT)

RT may be specified for internal flaw detection.

Commentary: The RT procedure consists of using an X-ray or gamma ray sourceto expose a film similar to that used in medical applications. The most commonshop and field technique uses an iridium 192 source of gamma rays on one sideof the member being inspected and a film cassette on the opposite side. Anexposure is made and the film developed much the way photographic negativesare produced. Areas of different film density relate to flaws in the weldment.

RT is sensitive to cracks, lack of fusion, lack of penetration, slag inclusionsand porosity defects. RT is rather insensitive to lamellar type defectsperpendicular to the path of radiation. It does produce a permanent film record. Due to its two dimensional capability, it gives limited information about thedepth of the defect or the angular orientation of a crack. RT has limitedapplication in WSMFs because groove welds in T-joints and the associatedgeometry of beam-column connections make it impractical. Additionally, thesurrounding area must be cleared of personnel for radiation safety requirements. RT is a very useful tool for inspection of groove welds in butt splices in plateapplications.

11.2.5 Ultrasonic Testing (UT)

UT should be specified as the main form of NDT used in support of VI for the testing ofWSMFs. The bottom beam flange to column flange weld should be inspected in accordance withthe requirements of AWS D1.1. The proper scanning techniques beam angle(s) and transducershould be used as specified in a written ultrasonic test procedure. The acceptance standardshould be that specified in the original contract documents. If these documents are unavailable


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the acceptance criteria of D1.1 Chapter 8 Statically Loaded Structures should be used. The shearwave scan should be preceded by a scan for laminations in the base metal as specified in D1.1. Rejectable discontinuities should be reported on a standard format as recommended by D1.1, i.e.;length, amplitude and classification. Reflections generated from the root and backing bar area ofthe weld may be cause for further exploration when:

1. the operator is unable to determine if the signal is from a crack or the weld backing.

2. a reflection can be detected in the web zone but the received signal is not great enough tocause rejection.

Although different angles, transducer sizes and scanning methods may be used to furtherevaluate the root area, the removal of the backing bar may be just as cost effective and will alwaysyield more positive results. After the backing has been thoroughly removed, the root should betested with MT to detect any linear indication.

Typically, on existing buildings being inspected for damage, only the inside face of the topflange of the beam to column weld is accessible. This will require the lower portion including theroot to be tested in the second leg of the ultrasonic sound path. This increases the difficulty ofevaluating the root and weld backing which is difficult enough to evaluate in the first leg of soundtravel. As in the bottom flange, all rejectable discontinuities should be recorded. If root defectsare found or discontinuities which are difficult to interpret, it should be the engineers decisionwhether or not to do further exploration by UT and/or remove the steel backing. Access maybecome a problem at perimeter columns where one half of the top beam flange is inaccessible.

Commentary: The UT test involves sending ultrasonic frequency sound wavesinto a weldment. Any reflector within the weld or parent metal sends back areflected signal to the instrument. The sent and received signals are presented onan oscilloscope for interpretation. Unlike RT, MT and PT, the interpretation ofthe received signal is highly dependent on the skill and training of the technician. The location and depth of the flaw can be accurately determined. The shape andtype can also be interpreted to some degree by competent operators. Thescanning surfaces must be clean and free from fireproofing, upset metal and weldspatter for proper transducer contact.

AWS D1.1, Section 6.19, requires that the entire area to be scanned by shearwave for weld flaw detection be first scanned by longitudinal wave to detect anylamellar defects. These defects can mask indications from the weld areas, ifpresent, and are not favorably oriented for shear wave testing.

UT is highly sensitive to planar defects if they are favorably oriented to thesound beam. The primary testing is done by utilizing a shear wave transducerfrom the flange faces of the beams. The key to detection is to select the propertesting angle which will intercept the flaw perpendicular to its orientation. Theamplitude of the received signal is directly related to the flaw orientation and,


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hence, the rejection criteria. In the typical T-joint configuration of WSMFconnections, defects in the HAZ of the prepared bevel and root area are favorablyoriented to the sound path. This is not the case for the column face HAZ which isnot optimally oriented. Sometimes this area can be inspected by using alongitudinal wave transducer from the back side of the column face if nocontinuity plates are present; however, AWS has no rejection criteria for thismethod.

UT technicians are prone to skipping the lamination check when pressed forproduction. Recalibration of the instrument is required each time the transduceris changed.

The intent of D1.1, 6.19.6.2 is to achieve shear wave testing from both the topof the beam flange (A surface) and from the bottom of the beam flange (Bsurface). High production pressures sometimes force premature movement ofscaffolding, allowing the UT technician access to only the top of the bottomflange (A surface). This precludes proper testing of the weld area below the beamweb.

Another area of concern is back-up bar removal. Removal of backing is leftas an option in D1.1 which defers to the Contract Documents. It is stronglyrecommended that back-up bar removal be required in the Contract Documentsto enhance visual and UT inspection.

A common problem with rejects identified by UT technicians occurs duringthe air-arc gouging of the defect area. If too large of a carbon arc electrode isused or if too large a pass is taken, the defect can easily be gouged out withoutever being observed by the welder or the UT inspector. For typical WSMF welds,a 1/4 or 3/16 inch maximum size electrode should be used and light skim passestaken. The UT technician should observe the process through a welding shield. A technician can be falsely lured into reducing his/her rejection criteria if nodefect is found during gouging.

Interim Guidelines:Evaluation, Repair, Modification andDesign of Steel Moment Frames References

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12. REFERENCES

ATLSS, Fractorgrahpic Analysis of Speciments from Failed Moment Connections, (publicationpending, title not exact), SAC, 1995.

ATLSS, Testing of Welded “T” Specimens, (publication pending, title not exact), SAC, 1995.

Allen J., Personal Correspondence, Test Reports for New Detail, July 30, 1995.

Allen J., Partridge, J.E., and Richard, R.M., Stress Distribution in Welded/Bolted Beam toColumn Moment Connections. The Allen Company, March, 1995.

American Association of State Highway and Transportaion Officials, Bridge Welding CodeAASHTO/AWS D1.5, 1995.

American Institute of Steel Construction, Statistical Analysis of Charpy V-notch Toughness ForSteel Wide Flange Structural Shapes, July, 1995.

American Institute of Steel Construction, Manual of Steel Construction, ASD, Ninth Edition,1989.

American Institute of Steel Construction, Manual of Steel Construction, LRFD, Second Edition,1994.

American Institute of Steel Construction, Load and Resistance Factor Design Specification forStructural Steel Buildings, December 1, 1993.

American Institute of Steel Construction, Specification for Structural Joints using ASTM A325or A490 Bolts. 1985.

American Institute of Steel Construction, AISC Northridge Steel Update I, October, 1994.

American Welding Society, Guide for Nondestructive Inspection of Welds, AWS B1.10-86, 1986.

American Welding Society, Guide for Visual Inspection of Welds, AWS B1.11-88, 1988.

American Welding Society, Structural Welding Code - Steel AWS, D1.1-94, 1994.

Anderson, J.C., Johnson, R.G., Partridge, J.E., “Post Earthquake Studies of A Damaged LowRise Office Building” Technical Report: Case Studies of Steel Moment Frame BuildingPerformance in the Northridge Earthquake of January 17, 1994 SAC 95-07. SAC, December,1995.

Anderson, J.C., Filippou, F.C., Dynamic Response Analysis of the 18 Story Canoga Building, SAC, March, 1995.


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Anderson, J.C., Test Results for Repaired Specimen NSF#1, Report to AISC Steel AdvisoryCommittee, June, 1995.

Applied Technology Council, Earthquake Damage Evaluation Data for California ATC-13,Redwood City, CA 1985.

Applied Technology Counicl, Procedures for Post Earthquake Safety Evaluations of BuildingsATC-20, Redwood City, CA, 1989.

Applied Technology Council, Guidelines for Cyclic Seismic Testing of Components of SteelStructures, ATC-24, Redwood City, CA, 1992.

Astaneh-Asl, A. Post-Earthquake Stability of Steel Moment rames with Damage Connections. Proceedings of the Third International Workshop on Connections in Steel Structures, Universityof Trento, Trento, Italy, 1995.

Avent, R., “Designing Heat-Straightening Repairs,” National Steel Construction ConferenceProceedings, Las Vegas, NV, AISC, 1992.

Avent, R., “Engineered Heat Straightening,” National Steel Construction ConferenceProceedings, San Antonio, TX, AISC, 1995.

Beck, J.L., May, B.S., Polidori, D.C., Vanik, M.W., “Ambient Vibration Surveys of Three Steel-Frame Buildings Strongly Shaken by the 1994 Northridge Earthquake”, Analytical and FieldInvestigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC, December, 1995.

Bertero, V.V., and Whittaker, A., Testing of Repaired Welded Beam Column Assemblies, SAC,publication pending (title not exact), 1995.

Blodgett, O., “Evaluation of Beam to Column Connections”, SAC Steel Moment FrameConnection Advisory No. 3, Feb. 1995.

Bonowitz, D, and Youssef, N. “SAC Survey of Steel-Moment Frames Affected by the 1994Northridge Earthquake”, Surveys and Assessment of Damage to Buildings Affected by theNorthridge Earthquake of January 17, 1994 SAC 95-06, SAC, 1995.

Building Seismic Safety Council. NEHRP Recommended Provisions for Seismic Regulations forNew Buildings -1991 Edition FEMA 222, FEMA 223, Washington D.C., January, 1992.

Building Seismic Safety Council. NEHRP Recommended Provisions for Seismic Regulations forNew Buildings -1994 Edition FEMA 222A, FEMA223A, Washington D.C., July, 1995.

Campbell, K.W. and Bazorgnia, Y., “Near Source Attentuation of Peak Horizontal Accelerationfrom World Wide Accelerogram Records from 1957 - 1993,” Proceedings of the Fifth NationalConference on Earthquake Engineering, Chicago, Ill, 1994.


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Campbell, S., “Modeling of Weld Fractures Using the Drain Programs”, Technical Report:Parametric Analytical Investigations of Ground Motion and Structural Response, NorthridgeEarthquake of January 17, 1994 SAC95-05. SAC, 1995.

Chen, S.J. and Yeh, C.H., Enhancement of Ductility of Steel Beam-to-Column Connections forSeismic Resistance, Department of Construction Engineering, National Taiwan University, May,1995.

Diererlein, G. “Summary of Building Analysis Studies” Analytical and Field Investigations ofBuildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC,December, 1995

Durkin, M. E., “Inspection, Damage, and Repair of Steel Frame Buildings Following theNorthridge Earthquake”, Technical Report: Surveys and Assessment of Damage to BuildingsAffected by the Northridge Earthquake of January 17, 1994 SAC 95-06, SAC, December, 1995.

Engelhardt, M.D., and Sabol, T.A. Testing of Welded Steel Moment Connections In Response tothe Northridge Earthquake, Progress Report to the AISC Advisory Subcommittee on SpecialMoment Resisting Frame Research, October, 1994.

Engelhardt, M. D., Keedong, K.M. Sabol T. A., Ho, L., Kim, H. Uzarski, J. and Abunnasar, H. “Analysis of a Six Story Steel Moment Frame Building in Santa Monica”, Analytical and FieldInvestigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1 SAC, December, 1995.

Engelhardt, M. D., Keedong, K.M., Uzarski, J., Abunassar, H., Sabol, T.A., Ho, L., and Kim, H.“Parametric Studies on Inelastic Modeling of Steel Moment Frames”, Technical Report:Parametric Analytical Investigations of Ground Motion and Structural Response, NorthridgeEarthquake of January 17, 1994 SAC95-05. SAC, 1995.

Engelhardt, M.D., et. al. Testing of Repaired Welded Beam Column Assemblies, SAC,publication pending (title not exact), 1995.

Englehardt, M.D., et. al. Accoustic Emission Recordings for Welded Beam Column AssemblyTests, SAC, publication pending (title not exact), 1995.

Frank, K.H. “The Physical and Metallurgical Properties Of Structural Steels” State of Art Papers:Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System BehaviorSAC 95-09. SAC, September, 1996

Fillippou, F.C. “Nonlinear Static and Dynamic Analysis of Canoga Park Towers with FEAP-STRUC”, Analytical and Field Investigations of Buildings Affected by the NorthridgeEarthquake of January 17, 1994, SAC 95-04 Part 2, SAC., December, 1995.



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Fisher, J.W., Dexter, R.J., and Kauffman, E.J., “Fracture Mechanics of Welded Structural SteelConnections.” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, MomentConnections and Frame System Behavior SAC 95-09. SAC, September, 1996

Forrel/Elsesser Engineers, Inc., Lawrence Berkeley National Labs Steel Joint Test - TechnicalBrief, San Francisco, CA, July 17, 1995.

Gates, W.E., and Morden, M., “Lessons from Inspection, Evaluation, Repair and Construction ofWelded Steel Moment Frames Following the Northridge Earthquake”, Surveys and Assessment ofDamage to Buildings Affected by the Northridge Earthquake of January 17, 1994 SAC 95-06SAC, December, 1995.

Gates, W.E. “Interpretation of SAC Survey Data on Damaged Welded Steel Moment FramesFollowing the Northridge Earthquake”, Surveys and Assessment of Damage to Buildings Affectedby the Northridge Earthquake of January 17, 1994 SAC 95-06, SAC, December, 1995.

Green, M. “Santa Clarita City Hall; Northridge Earthquake Damage” Technical Report: CaseStudies of Steel Moment Frame Building Performance in the Northridge Earthquake of January17, 1994 SAC 95-07. SAC, December, 1995.

Hall, J.F., “Parameter Study of the Response of Moment-Resisting Steel Frame Buildings toNear-Source Ground Motions”, Technical Report: Parametric Analytical Investigations ofGround Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC95-05. SAC, 1995.

Hajjar, J.F., O’Sullivan D.P., Leon, R. T., Gourley, B.C. “Evaluation of the Damage to the BoraxCorporate Headquarters Building As A Result of the Northridge Earthquake”, Technical Report:Case Studies of Steel Moment Frame Building Performance in the Northridge Earthquake ofJanuary 17, 1994 SAC 95-07. SAC, December, 1995.

Harrison, P.L. and Webster, S.E., Examination of Two Moment Resisting Frame ConnectorsUtilizing a Cover-Plate Design, Brittish Steel Technical, Swinden Laboratories, Moorgate,Rotherham, 1995.

Hart, G.C., Huang, S.C., Lobo, R.F., Van Winkle, M., Jain, A., “Earthquake Response ofStrengthened Steel Special moment Resisting Frames” Analytical and Field Investigations ofBuildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC.,December, 1995

Hart, G.C., Huang, S., Lobo, R., and Stewart, J., “Elastic and Inelastic Analysis for Weld FailurePrediction of Two Adjacent Steel Buildings”, ” Analytical and Field Investigations of BuildingsAffected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC, December,1995.


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Hart, G.C., Huang, S., Lobo, R., and Stewart, J., “Influence of Vertical Ground Motion onSpecial Moment-Resisting Frames”, Technical Report: Parametric Analytical Investigations ofGround Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC95-05. SAC, 1995.

Heaton, T.H., Hall, J.F., Wald, D.J., and Halling, M.W. “Response of High-Rise and Base-Isolated Buildings to a Hypothetical Mw 7.0 Blind Thrust Earthquake” Science Vol. 26, pp 206-211, January, 1995.

International Conference of Building Officials, Uniform Building Code UBC-94. Whittier, CA,1994.

Iwan, W.D., “Drift Demand Spectra for Selected Northridge Sites”, Technical Report:Parametric Analytical Investigations of Ground Motion and Structural Response, NorthridgeEarthquake of January 17, 1994 SAC95-05. SAC, 1995.

Joyner, W.B., and Boore, D.M., “Ground Motion Parameters for Seismic Design,”Bulletin of theSesimological Society of America, 1994.

Kariotis, J. and Eimani, T.J., “Analysis of a Sixteen Story Steel Frame Building at Site 5, for theNorthridge Earthquake”, Analytical and Field Investigations of Buildings Affected by theNorthridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC, December, 1995.

Krawinkler, H.K., “Systems Behavior of Structural Steel Frames Subjected to EarthquakeGround Motions” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, MomentConnections and Frame System Behavior SAC 95-09. SAC, September, 1996

Krawinkler, H.K., Ali, A.A., Thiel, C.C., Dunlea, J.M., “Analysis of a Damaged 4-Story Buildingand an Undamaged 2- Story Building”, Analytical and Field Investigations of Buildings Affectedby the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC, December, 1995.

Leon, R. T., “Seismic Performance of Bolted and Riveted Connections” State of Art Papers:Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System BehaviorSAC 95-09. SAC, September, 1996

Miller, D.K. “Welding of Seismically Resistant Steel Structures” State of Art Papers: Metallurgy,Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09.SAC, September, 1996

Naeim F., DiJulio, R., Benuska, K., Reinhorn, A. M., and Chen, L. “Evaluation of SeismicPerformance of an 11 Story Steel Moment Frame Building During the 1994 NorthridgeEarthquake”, ” Analytical and Field Investigations of Buildings Affected by the NorthridgeEarthquake of January 17, 1994, SAC 95-04 Part 2 SAC, December, 1995.



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Newmark, N.M. and Hall W.J., Earthquake Spectra and Design. Earthquake EngineeringResearch Institute, 1982.

Paret, T.F., Sasaki, K.K., “Analysis of a 17 Story Steel Moment Frame Building Damaged by theNorthridge Earthquake”, Analytical and Field Investigations of Buildings Affected by theNorthridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC, December, 1995.

Popov, E.P. and Yang, T.S. Steel Seismic Moment Resisting Connections. University ofCalifornia at Berkeley, May, 1995.

Popov, E.P., et. al. Testing of Repaired Welded Beam Column Assemblies, SAC, publicationpending (title not exact), 1995.

SAC, Proceedings of the International Workshop on Steel Moment Frames, October 23-24, 1994SAC-94-01. Sacramento, CA, December, 1994.

SAC . Steel Moment Frame Advisory No. 1. September, Sacramento, CA, 1994.

SAC . Steel Moment Frame Advisory No. 2. October, Sacramento, CA, 1994.

SAC . Steel Moment Frame Advisory No. 3 SAC-95-01, February, Sacramento, CA, 1995.

Shonafelt, G.O., and Horn, W.B.. Guidelines for Evaluation and Repair of Damaged SteelBridge Members, NCHRP Report 271, Transportation Research Board, 1984.

Skiles, J.L. and Campbell, H.H., “Why Steel Fractured in the Northridge Earthquake” SACAdvisory No. 1, October, 1994.

Seismic Safety Commission, Northridge Earthquake Turning Loss to Gain, Report to theGovernor, Sacramento, CA, 1995.

Smith-Emry Company. Report of Test, July, 1995.

Sommerville, P, Graves, R., Chandan, S. Technical Report: Characterization of Ground MotionDuring the Northridge Earthquake of January 17, 1994, SAC 95-03, SAC, December, 1995.

State of California. Division of the State Architect (DSA) and Office of Statewide HealthPlanning and Development (OSHPD). Interpretation of Regulations Steel Moment ResistingFrames, Sacramento, CA, 1994.

Structural Engineers Association of California (SEAOC), Seismology Committee, RecommendedLateral Force Requirements and Commentary, Sacramento, CA. 1990.

Structural Engineers Association of California (SEAOC), Seismology Committee, InterimRecommendations for Design of Steel Moment Resisting Connection,. Sacramento, CA, January,1995.


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Structural Engineers Association of California (SEAOC), Vision 2000: A Framework forPerformance Based Engineering of Buildings, Sacramento, CA, April, 1995.

Structural Shape Producers Council, Statistical Analysis of Tensile Data for Wide FlangeStructural Shapes, 1994.

Thiel, C.C., and Zsutty, T.C., “Earthquake Characteristics and Damage Statistics,” EarthquakeSpectra, Volume 3, No. 4., Earthquake Engineering Research Institute, Oakland, Ca. 1987.

Tsai, K.C. and Popov, E. P. “Seismic Steel Beam-Column Moment Connections” State of ArtPapers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame SystemBehavior SAC 95-09. SAC, September, 1996

Uang, C.M. and Latham, C.T. Cyclic Testing of Full-Scale MNH-SMRF Moment Connections,Structural Systems Research, University of California, San Diego, March, 1995.

Tsai, K.C. and Popov, E.P., Steel Beam - Column Joints In Seismic Moment Resisting Frames,Report No. UCB/EERC-88/19, Earthquake Engineering Research Center, University ofCalifornia, Berkeley, Nov., 1988.

Uang, C.M., Yu, Q.S., Sadre, A., Bonowitz, D., Youssef, N. “Performance of a 13 Story SteelMoment-Resisting Frame Damaged in the 1994 Northridge Earthquake”, ” Analytical and FieldInvestigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2 SAC, December, 1995.

Uang, C.M. and Bondad, D. Progress Report on Cyclic Testing of Three Repaired UCSDSpecimens, SAC, 1995.

Uang, C.M. and Lee, C.H. “Seismic Response of Haunch Repaired Steel SMRFs: AnalyticalModelling and Case Studies” ” Analytical and Field Investigations of Buildings Affected by theNorthridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC., December, 1995

Wald, D.J., Heaton, T.H., and Hudnut, K.W., The Slip History of the 1994 Northridge,California, Earthquake Determined from Strong-Motion, Teleseismic, GPS, and Leveling Data,United Sates Geologic Survey, 1995.

Watabe, M. Peformance of Wooden Houses and Steel Buildings during the Great HanshinEarthquake, Architectural Institute of Japan, May, 1995.

Youssef, N.F.G, Bonowitz, D., and Gross, J.L., A Survey of Steel Moment-Resisting FrameBuildings Affected by the 1994 Northridge Earthquake, NISTR 5625, Gaithersburg Md, April,1995.

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FEDERAL EMERGENCY MANAGEMENT AGENCY FEMA 267b / June, 1999

Interim GuidelinesAdvisory No. 2

Supplement to FEMA-267

i

INTERIM GUIDELINES ADVISORY NO. 2Supplement to FEMA-267 Interim Guidelines:

Evaluation, Repair, Modification and Design ofWelded Steel Moment Frame Structures

Report No. SAC-99-01

SAC Joint Venturea partnership of:

Structural Engineers Association of California (SEAOC)Applied Technology Council (ATC)

California Universities for Research in Earthquake Engineering (CUREe)

Prepared for SAC Joint Venture Partnership byGuidelines Development Committee

Ronald O. Hamburger, Chair

John D. HooperRobert E. Shaw

Lawrence D. Reaveley

Thomas SabolC. Mark Saunders

Raymond H.R. Tide

Project Oversight CommitteeWilliam J. Hall, Chair

John N. BarsomShirin Ader

John BarsomRoger Ferch

Theodore V. GalambosJohn Gross

James R. Harris

Richard HolguinNestor IwankiwRoy G. Johnston

Len Joseph Duane K. Miller

John TheissJohn H. Wiggins

SAC Project Management CommitteeSEAOC: William T. HolmesATC: Christoper RojahnCUREe: Robin Shepherd

Program Manager: Stephen A. MahinInvestigations Director: James O. MalleyProduct Director: Ronald O. Hamburger

Federal Emergency Management AgencyProject Officer: Michael Mahoney Technical Advisor: Robert D. Hanson

SAC Joint Venture555 University Avenue, Suite 126

Sacramento, California 95825916-427-3647

June, 1999

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THE SAC JOINT VENTURE

SAC is a joint venture of the Structural Engineers Association of California (SEAOC), the Applied Technology Council(ATC), and California Universities for Research in Earthquake Engineering (CUREe,) formed specifically to address bothimmediate and long-term needs related to solving problems of the Welded Steel Moment Frame (WSMF) connection thatbecame apparent as a result of the 1994 Northridge earthquake. SEAOC is a professional organization composed ofmore than 3,000 practicing structural engineers in California. The volunteer efforts of SEAOC’s members on varioustechnical committees have been instrumental in the development of the earthquake design provisions contained in theUniform Building Code as well as the National Earthquake Hazards Reduction Program (NEHRP) Provisions for SeismicRegulations for New Buildings. The Applied Technology Council is a non-profit organization founded specifically toperform problem-focused research related to structural engineering and to bridge the gap between civil engineeringresearch and engineering practice. It has developed a number of publications of national significance including ATC 3-06, which serves as the basis for the NEHRP Recommended Provisions. CUREe is a nonprofit organization formed topromote and conduct research and educational activities related to earthquake hazard mitigation. CUREe’s eightinstitutional members are: the California Institute of Technology, Stanford University, the University of California atBerkeley, the University of California at Davis, the University of California at Irvine, the University of California at LosAngeles, the University of California at San Diego, and the University of Southern California. This collection ofuniversity earthquake research laboratory, library, computer and faculty resources is among the most extensive in theUnited States. The SAC Joint Venture allows these three organizations to combine their extensive and unique resources,augmented by subcontractor universities and organizations from around the nation, into an integrated team ofpractitioners and researchers, uniquely qualified to solve problems related to the seismic performance of WSMFstructures.

DISCLAIMER

The purpose of this document is to serve as a supplement to the FEMA-267 publication Interim Guidelines: Evaluation,Repair, Modification and Design of Welded Steel Moment Frame Structures. This Advisory, which is intended to be usedin conjunction with FEMA-267, supercedes and entirely replaces Interim Guidelines Advisory No. 1 (FEMA 267a). FEMA-267 was published to provide engineers and building officials with guidance on engineering procedures forevaluation, repair, modification and design of welded steel moment frame structures, to reduce the risks associated withearthquake-induced damage. The recommendations were developed by practicing engineers based on professionaljudgment and experience and a preliminary program of laboratory, field and analytical research. This preliminaryresearch, known as the SAC Phase 1 program, commenced in November, 1994 and continued through the publication ofthe Interim Guidelines document. This Interim Guidelines Advisory No. 2, which updates and replaces InterimGuidelines Advisory No. 1, is based on supplementary data developed under a program of continuing research, known asthe SAC Phase 2 program, as well as findings developed by other, independent researchers. Final designrecommendations, superceding both FEMA-267 and this document are scheduled for publication in early 2000. Independent review and guidance in the production of both the FEMA-267, Interim Guidelines and the advisories wasprovided by a project oversight panel comprised of experts from industry, practice and academia. Users are cautioned thatresearch into the behavior of these structures is continuing. Interpretation of the results of this research may invalidate orsuggest the need for modification of recommendations contained herein. No warranty is offered with regard to therecommendations contained herein, either by the Federal Emergency Management Agency, the SAC JointVenture, the individual joint venture partners, their directors, members or employees. These organizations andtheir employees do not assume any legal liability or responsibility for the accuracy, completeness, or usefulness ofany of the information, products or processes included in this publication. The reader is cautioned to carefullyreview the material presented herein. Such information must be used together with sound engineering judgment whenapplied to specific engineering projects. This Interim Guidelines Advisory has been prepared by the SAC Joint Venturewith funding provided by the Federal Emergency Management Agency, under contract number EMW-95-C-4770. TheSAC Joint Venture gratefully acknowledges the support of FEMA and the leadership of Michael Mahoney and RobertHanson, Project Officer and Technical Advisor, respectively. The SAC Joint Venture also wishes to express its gratitudeto the large numbers of engineers, building officials, organizations and firms that provided substantial efforts, materials,and advice and who have contributed significantly to the progress of the Phase 2 effort.

Interim Guidelines Advisory No. 2 SAC 99-01

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PREFACE

Purpose

The purpose of the Interim Guidelines Advisory series is to provide engineers and buildingofficials with timely information and guidance resulting from ongoing problem-focused studies ofthe seismic behavior of moment-resisting steel frame structures. These advisories are intended tobe supplements to FEMA-267 Interim Guidelines: Evaluation, Repair, Modification and Designof Welded Steel Moment Frame Structures first published in August 1995.

The first Interim Guidelines Advisory, FEMA-267a, was published in January 1997. Thespecific revisions and updates to the Interim Guidelines contained in FEMA-267a were developedbased on input obtained from a group of engineers and building officials actively engaged in theuse of the FEMA-267 document, in the period since its initial publication in August 1995. Thatinput was obtained during a workshop held in August 1996, in Los Angeles, California.

This second Interim Guidelines Advisory has been prepared as a series of updates andrevisions both to the FEMA-267, Interim Guidelines which it supplements and to the FEMA-267a, Interim Guidelines Advisory publication, which it supercedes. The material contained inthis Interim Guidelines Advisory No. 2 is based on the extensive analytical and laboratoryresearch that has been conducted by the SAC Joint Venture and other researchers during theintervening period, along with recent developments in the steel construction industry. Thematerial contained in this Advisory has been formatted to match that contained in the originalInterim Guidelines, to permit the user to insert this material directly into appropriate sections ofthat document. This Advisory is not intended to serve as a self-contained text and should not beused as such. It does, however, completely replace the material contained in FEMA-267a.

A new set of recommendations for the design, analysis, evaluation repair, retrofit andconstruction of moment-resisting steel frames is currently being prepared as part of the Phase 2Program to Reduce Earthquake Hazards in Steel Moment Frame Structures. These new SeismicDesign Criteria, which are anticipated to be completed early in the year 2000, will replace in theirentirety the FEMA-267 Interim Guidelines and this Interim Guidelines Advisory No. 2.

Background

The Northridge earthquake of January 17, 1994, dramatically demonstrated that theprequalified, welded beam-to-column moment connection commonly used in the construction ofwelded steel moment resisting frames (WSMFs) in the period 1965-1994 was much moresusceptible to damage than previously thought. The stability of moment frame structures inearthquakes is dependent on the capacity of the beam-column connection to remain intact and toresist tendencies of the beams and columns to rotate with respect to each other under theinfluence of lateral deflection of the structure. The prequalified connections were believed to beductile and capable of withstanding the repeated cycles of large inelastic deformation explicitlyrelied upon in the building code provisions for the design of these structures. Although manyaffected connections were not damaged, a wide spectrum of unexpected brittle connection


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fractures did occur, ranging from isolated fractures through or adjacent to the welds of beamflanges to columns, to large fractures extending across the full depth of the columns. At the timethis damage was discovered, the structural steel industry and engineering profession had littleunderstanding of the specific causes of this damage, the implications of this damage for buildingsafety, or even if reliable methods existed to repair the damage which had been discovered. Although the connection failures did not result in any casualties or collapses, and many WSMFbuildings were not damaged, the incidence of damage was sufficiently pervasive in regions ofstrong ground motion to cause wide-spread concern by structural engineers and building officialswith regard to the safety of these structures in future earthquakes.

In response to these concerns, the Federal Emergency Management Agency (FEMA) enteredinto a cooperative agreement with the SAC Joint Venture to perform problem-focused study ofthe seismic performance of welded steel moment connections and to develop interimrecommendations for professional practice. Specifically, these recommendations were intended toaddress the inspection of earthquake affected buildings to determine if they had sustainedsignificant damage; the repair of damaged buildings; the upgrade of existing buildings to improvetheir probable future performance; and the design of new structures to provide more reliableseismic performance. Within weeks of receipt of notification of FEMA’s intent to enter into thisagreement, the SAC Joint Venture published a series of two design advisories (SAC, 1994a; SAC,1994b). These design advisories presented a series of papers, prepared by engineers andresearchers engaged in the investigation of the damaged structures and presenting individualopinions as to the causes of the damage, potential methods of repair, and possible designs formore reliable connections in the future. In February 1995, Design Advisory No. 3 (SAC, 1995a)was published. This third advisory presented a synthesis of the data presented in the earlierpublications, together with the preliminary recommendations developed in an industry workshop,attended by more than 50 practicing engineers, industry representatives and researchers, onmethods of inspecting, repairing and designing WSMF structures. At the time this third advisorywas published, significant disagreement remained within the industry and the profession as to thespecific causes of the damage observed and appropriate methods of repair given that the damagehad occurred. Consequently, the preliminary recommendations were presented as a series of issuestatements, followed by the consensus opinions of the workshop attendees, where consensusexisted, and by majority and dissenting opinions where such consensus could not be formed.

During the first half of 1995, an intensive program of research was conducted to moredefinitively explore the pertinent issues. This research included literature surveys, data collectionon affected structures, statistical evaluation of the collected data, analytical studies of damagedand undamaged buildings and laboratory testing of a series of full-scale beam-column assembliesrepresenting typical pre-Northridge design and construction practice as well as various repair,upgrade and alternative design details. The findings of this research (SAC 1995c, SAC 1995d,SAC 1995e, SAC 1995f, SAC 1995g, SAC 1996) formed the basis for the development of FEMA267 - Interim Guidelines: Evaluation, Repair, Modification, and Design of Welded Steel MomentFrame Structures (SAC, 1995b), which was published in August, 1995. FEMA 267 provided thefirst definite, albeit interim, recommendations for practice, following the discovery of connectiondamage in the Northridge earthquake.


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As a result of these and supplemental studies conducted by the SAC Joint Venture, as well asindependent research conducted by others, it is now known that a large number of factorscontributed to the damage sustained by steel frame buildings in the Northridge earthquake. Theseincluded:

• design practice that favored the use of relatively few frame bays to resist lateralseismic demands, resulting in much larger member and connection geometries than hadpreviously been tested;

• standard detailing practice which resulted in the development of large inelasticdemands at the beam to column connections;

• detailing practice that often resulted in large stress concentrations in the beam-columnconnection, as well as inherent stress risers and notches in zones of high stress;

• the common use of welding procedures that resulted in deposition of low toughnessweld metal in the critical beam flange to column flange joints;

• relatively poor levels of quality control and assurance in the construction process,resulting in welded joints that did not conform to the applicable quality standards;

• excessively weak and flexible column panel zones that resulted in large secondarystresses in the beam flange to column flange joints;

• large variations in the strengths of rolled shape members relative to specified values;

• an inherent inability of material to yield under conditions of high tri-axial restraint suchas exist at the center of the beam flange to column flange joints.

With the identification of these factors it was possible for FEMA 267 to present arecommended methodology for the design and construction of moment-resisting steel frames toprovide connections capable of more reliable seismic performance. This methodology includedthe following recommendations:

• proportion the beam-column connection such that inelastic behavior occurs at adistance remote from the column face, minimizing demands on the highly restrainedcolumn material and the welded joints;

• specify weld filler metals with rated toughness values for critical welded joints;

• detail connections to incorporate beam flange continuity plates, to minimize stressconcentrations;

• remove backing bars and weld tabs from critical joints to minimize the potential forstress risers and notch effects and also to improve the reliability with which flaws atthe weld root can be observed and repaired;


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• qualify connection configurations through a program of full-scale inelastic testing ofrepresentative beam-column assemblies, fabricated in the same manner as is proposedfor use in the structure;

• increased participation of the design professional in the specification and surveillanceof welding procedures and the quality assurance process for welded joints.

In the time since the publication of FEMA-267, SAC has continued, under funding providedby FEMA, to perform problem-focused study of the performance of moment resisting connectionsof various configurations. This work, which is generally referred to as the SAC Phase II program,includes detailed analytical evaluations of buildings and connections, parametric studies into theeffects on connection performance of connection configuration, base and weld metal strength,toughness and ductility, as well as additional large scale testing of connection assemblies. Theintent of this study is to support development of final guidelines that will present more reliable andeconomical performance-based methods for:

• identification of damaged structures following an earthquake and determination of theextent, severity and consequences of such damage;

• design of effective repairs for damaged structures;

• identification of existing structures that are vulnerable to unacceptable levels ofdamage in future earthquakes;

• design of structural upgrades for existing vulnerable structures;

• design of new structures that are suitably resistant to earthquake induced damage;

• procedures for construction quality assurance that are consistent with the levels ofreliability intended by the design criteria.

This Phase II program of research, which is being conducted by the SAC Joint Venture inparallel and coordination with work by other researchers, is anticipated to be complete in late1999. It is the intent of FEMA and the SAC Joint Venture to ensure that pertinent informationand findings from this program are made available to the user community in a timely mannerthrough the publication of this series of design advisory documents. This Interim GuidelinesAdvisory No. 2 is the second such publication.

Format

This Advisory has been prepared as a series of updates and revisions to the FEMA-267,Interim Guidelines publication. It has been formatted in a manner intended to facilitate theidentification of changes to the original FEMA-267 text. Only those sections of FEMA-267 thatare being revised at this time are included. Other sections of FEMA-267 remain in effect as thecurrent best recommendations of the SAC Joint Venture. This Advisory replaces the earlierInterim Guidelines Advisory, FEMA-267a, in its entirety.


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To facilitate coordination of this Advisory with FEMA-267, the existing system of chapter andsection numbering has been retained. The Table of Contents lists all sections of the chaptersbeing revised, including those sections for which no revisions are included. Within the body ofthis document, a section heading is provided for each section of the chapter; however, if norevision to the section is currently being made, this is indicated immediately beneath the sectionheading.

To facilitate reading of this document, where a revision is made to a section in FEMA 267, theentire text of that section is included herein. Where existing text from FEMA-267 is reproducedin this document, without edit, it is shown in normal face type for guidelines, and in italicized typefor commentary. Where existing text is being deleted, this is shown in strike through format. Asingle strikethrough indicates text deleted in the first advisory, FEMA-267a. A doublestrikethrough indicates text deleted in this current advisory. New text is shown in underlineformat. A single underline identifies text added in the first advisory, FEMA-267a. A doubleunderline identifies text added in this current advisory. When a modification has been made to aportion of text, relative to FEMA-267, this will also be noted by the presence of a vertical line atthe outside margin of the page. The following two paragraphs illustrate these conventions forguideline and commentary text, respectively.

This sentence is representative of typical guideline text, that has been reprintedfrom FEMA-267 without change.This sentence, is representative of the way inwhich text being deleted from FEMA-267 in this Interim Guidelines Advisory isidentified. This sentence illustrates the way in which text deleted from FEMA-267in the previous Interim Guidelines Advisory is identified. This sentence illustratesthe way in which text being added to FEMA-267 in this Interim GuidelinesAdvisory is identified.This sentence illustrates the way in which text added toFEMA-267 in the previous Interim Guidelines Advisory is identified.

Commentary: This sentence is representative of typical commentary text, that hasbeen reprinted from FEMA-267 without change. This sentence is representative ofthe way in which commentary text being deleted from FEMA-267 in this InterimGuidelines Advisory is identified. However, this sentence, is representative of theway in which text being deleted from FEMA-267 commentary in the previousAdvisory is identified. This sentence indicates the way in which text added to theFEMA-267 commentary in this Advisory is shown.This final sentence illustratesthe way in which text added in previous advisory, FEMA-267a, is identified.

Intent

This Interim Guidelines Advisory, together with the Interim Guidelines they modify, are primarilyintended for two different groups of potential users:

a) Engineers engaged in evaluation, repair, and upgrade of existing WSMF buildings and inthe design of new WSMF buildings incorporating either Special Moment-Resisting Framesor Ordinary Moment-Resisting Frames utilizing welded beam-column connections. The


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recommendations for new construction are applicable to all WSMF construction expectedto resist earthquake demands through plastic behavior.

b) Regulators and building departments responsible for control of the evaluation, repair, andoccupancy of WSMF buildings that have been subjected to strong ground motion and forregulation of the design, construction, and inspection of new WSMF buildings.

The fundamental goal of the information presented in the Interim Guidelines as modified by thisAdvisory is to help identify and reduce the risks associated with earthquake-induced fractures inWSMF buildings through provision of timely information on how to inspect existing buildings fordamage, repair damage if found, upgrade existing buildings and design new buildings. The informationpresented here primarily addresses the issue of beam-to-column connection integrity under the severeinelastic demands that can be produced by building response to strong ground motion. Users arereferred to the applicable provisions of the locally prevailing building code for information with regardto other aspects of building construction and earthquake damage control.

Limitations

The information presented in this Interim Guidelines Advisory, together with that contained in theInterim Guidelines it modifies, is based on limited research conducted since the NorthridgeEarthquake, review of past research and the considerable experience and judgment of the professionalsengaged by SAC to prepare and review this document. Additional research on such topics as the effectof floor slabs on frame behavior, the effect of weld metal and base metal toughness, the efficacy ofvarious beam-column connection details and the validity of current standard testing protocols forprediction of earthquake performance of structures is continuing as part of the Phase 2 program and isexpected to provide important information not available at the time this Advisory was formulated. Therefore, many of the recommendations cited herein may change as a result of forthcoming researchresults.

The recommendations presented herein represent the group consensus of the committee ofGuideline Writers retained by SAC following independent review by the Project OversightCommittee. They may not reflect the individual opinions of any single participant. They do notnecessarily represent the opinions of the SAC Joint Venture, the Joint Venture partners, or thesponsoring agencies. Users are cautioned that available information on the nature of the WSMFproblem is in a rapid stage of development and any information presented herein must be usedwith caution and sound engineering judgment.


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TABLE OF CONTENTSTHE SAC JOINT VENTURE iiDISCLAIMER iiPREFACE iii

Purpose iiiBackground iiiFormat viIntent viiLimitations viii

1 INTRODUCTION1.1 Purpose 1-11.2 Scope 1-11.3 Background 1-11.4 The SAC Joint Venture 1-81.5 Sponsors 1-81.6 Summary of Phase I Research 1-81.7 Intent 1-81.8 Limitations 1-91.9 Use of the Guidelines 1-9

3 CLASSIFICATIONS AND IMPLICATIONS OF DAMAGE3.1 Summary of Earthquake Damage 3-13.2 Damage Types 3-1

3.2.1 Girder Damage 3-13.2.2 Column Flange Damage 3-13.2.3 Weld Damage, Defects and Discontinuities 3-13.2.4 Shear Tab Damage 3-43.2.5 Panel Zone Damage 3-43.2.6 Other Damage 3-4

3.3 Safety Implications 3-53.4 Economic Implications 3-7

4 POST-EARTHQUAKE EVALUATION4.1 Scope 4-14.2 Preliminary Evaluation 4-1

4.2.1 Evaluation Process 4-14.2.1.1 Ground Motion 4-14.2.1.2 Additional Indicators 4-1

4.2.2 Evaluation Schedule 4-14.2.3 Connection Inspections 4-2

4.2.3.1 Analytical Evaluation 4-24.2.3.2 Buildings with Enhanced Connections 4-3

4.2.4 Previous Evaluations and Inspections 4-3


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4.3 Detailed Evaluation Procedure 4-34.3.1 Eight Step Inspection and Evaluation Procedure 4-34.3.2 Step 1 - Categorize Connections By Group 4-44.3.3 Step 2 - Select Samples of Connections for Inspection 4-4

4.3.3.1 Method A - Random Selection 4-54.3.3.2 Method B - Deterministic Selection 4-54.3.3.3 Method C - Analytical Selection 4-5

4.3.4 Step 3- Inspect the Selected Samples of Connections 4-54.3.4.1 Damage Characterization 4-5

4.3.5 Step 4 - Inspect Connections Adjacent to Damaged Connections 4-84.3.6 Step 5 - Determine Average Damage Index for the Group 4-84.3.7 Step 6 - Determine the Probability that the Connections in a

Group at a Floor Level Sustained Excessive Damage 4-94.3.7.1 Some Connections In Group Not Inspected 4-94.3.7.2 All Connections in Group Inspected 4-9

4.3.8 Step 7 - Determine Recommended RecoveryStrategies for the Building 4-9

4.3.9 Step 8 - Evaluation Report 4-94.4 Alternative Group Selection for Torsional Response 4-94.5 Qualified Independent Engineering Review 4-9

4.5.1 Timing of Independent Review 4-94.5.2 Qualifications and Terms of Employment 4-94.5.3 Scope of Review 4-94.5.4 Reports 4-94.5.5 Responses and Corrective Actions 4-104.5.6 Distribution of Reports 4-104.5.7 Engineer of Record 4-104.5.8 Resolution of Differences 4-10

5 POST-EARTHQUAKE INSPECTION5.1 Connection Types Requiring Inspection 5-1

5.1.1 Welded Steel Moment Frame (WSMF) Connections 5-15.1.2 Gravity Connections 5-35.1.3 Other Connection Types 5-3

5.2 Preparation 5-45.2.1 Preliminary Document Review and Evaluation 5-4

5.2.1.1 Document Collection and Review 5-45.2.1.2 Preliminary Building Walk-Through 5-45.2.1.3 Structural Analysis 5-45.2.1.4 Vertical Plumbness Check 5-4

5.2.2 Connection Exposure 5-45.3 Inspection Program 5-6

5.3.1 Visual Inspection (VI) 5-65.3.1.1 Top Flange 5-65.3.1.2 Bottom Flange 5-6


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5.3.1.3 Column and Continuity Plates 5-65.3.1.4 Beam Web Shear Connection 5-7

5.3.2 Nondestructive Testing (NDT) 5-75.3.3 Inspector Qualification 5-95.3.4 Post-Earthquake Field Inspection Report 5-95.3.5 Written Report 5-9

6 POST-EARTHQUAKE REPAIR AND MODIFICATION6.1 Scope 6-16.2 Shoring 6-16.3 Repair Details 6-16.4 Preparation 6-16.5 Execution 6-16.6 Structural Modification 6-1

6.6.1 Definition of Modification 6-16.6.2 Damaged vs. Undamaged Connections 6-16.6.3 Criteria 6-16.6.4 Strength and Stiffness 6-4

6.6.4.1 Strength 6-46.6.4.2 Stiffness 6-6

6.6.5 Plastic Rotation Capacity 6-76.6.6 Connection Qualification and Design 6-10

6.6.6.1 Qualification Test Protocol 6-116.6.6.2 Acceptance Criteria 6-116.6.6.3 Calculations 6-12

6.6.6.3.1 Material Strength Properties 6-136.6.6.3.2 Determine Plastic Hinge Location 6-166.6.6.3.3 Determine Probable Plastic Moment at Hinges 6-186.6.6.3.4 Determine Beam Shear 6-196.6.6.3.5 Determine Strength Demands on Connection 6-206.6.6.3.6 Check Strong Column - Weak Beam Conditions 6-216.6.6.3.7 Check Column Panel Zone 6-23

6.6.7 Modification Details 6-246.6.7.1 Haunch at Bottom Flange 6-246.6.7.2 Top and Bottom Haunch 6-266.6.7.3 Cover Plate Sections 6-266.6.7.4 Upstanding Ribs 6-286.6.7.5 Side-Plate Connections 6-296.6.7.6 Bolted Brackets 6-29

7 NEW CONSTRUCTION7.1 Scope 7-17.2 General - Welded Steel Frame Design Criteria 7-3

7.2.1 Criteria 7-37.2.2 Strength and Stiffness 7-4


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7.2.2.1 Strength 7-47.2.2.2 Stiffness 7-5

7.2.3 Configuration 7-67.2.4 Plastic Rotation Capacity 7-97.2.5 Redundancy 7-137.2.6 System Performance 7-157.2.7 Special Systems 7-15

7.3 Connection Design and Qualification Procedures - General 7-157.3.1 Connection Performance Intent 7-157.3.2 Qualification by Testing 7-167.3.3 Design by Calculation 7-16

7.4 Guidelines for Connection Qualification by Testing 7-167.4.1 Testing Protocol 7-167.4.2 Acceptance Criteria 7-16

7.5 Guidelines for Connection Design by Calculation 7-187.5.1 Material Strength Properties 7-187.5.2 Design Procedure - Strengthened Connections 7-23

7.5.2.1 Determine Plastic Hinge Locations 7-237.5.2.2 Determine Probable Plastic Moment at Hinge 7-247.5.2.3 Determine Shear at Plastic Hinge 7-267.5.2.4 Determine Strength Demands at Critical Sections 7-267.5.2.5 Check for Strong Column - Weak Beam Condition 7-277.5.2.6 Check Column Panel Zone 7-29

7.5.3 Design Procedure - Reduced Beam Section Connections 7-307.5.3.1 Determine Reduced Section and Plastic Hinge Locations 7-337.5.3.2 Determine Strength and Probable Plastic Moment in RBS 7-337.5.3.3 Strong Column - Weak Beam Condition 7-357.5.3.4 Column Panel Zone 7-367.5.3.5 Lateral Bracing 7-367.5.3.6 Welded Attachments 7-37

7.6 Metallurgy & Welding 7-387.7 Quality Control / Quality Assurance 7-387.8 Guidelines on Other Connection Design Issues 7-38

7.8.1 Design of Panel Zones 7-397.8.2 Design of Web Connections to Column Flanges 7-397.8.3 Design of Continuity Plates 7-407.8.4 Design of Weak Column and Weak Way Connections 7-40

7.9 Moment Frame Connections for Consideration in New Construction 7-407.9.1 Cover Plate Connections 7-407.9.2 Flange Rib Connections 7-437.9.3 Bottom Haunch Connections 7-447.9.4 Top and Bottom Haunch Connections 7-467.9.5 Side-Plate Connections 7-467.9.6 Reduced Beam Section Connections 7-467.9.7 Slip-Friction Energy Dissipating Connections 7-48


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7.9.8 Column Tree Connections 7-487.9.9 Slotted Web Connections 7-487.9.10 Bolted Bracket Connections 7-50

7.10 Other Types of Welded Connection Structures 7-527.10.1 Eccentrically Braced Frames (EBF) 7-527.10.2 Dual Systems 7-527.10.3 Welded Base Plate Details 7-527.10.4 Vierendeel Truss Systems 7-527.10.5 Moment Frame Tubular Systems 7-527.10.6 Welded Connections of Collectors, Ties and Diaphragm Chords 7-537.10.7 Welded Column Splices 7-537.10.8 Built-up Moment Frame Members 7-53

8 METALLURGY & WELDING

8.1 Parent Materials 8-18.1.1 Steels 8-18.1.2 Chemistry 8-38.1.3 Tensile/Elongation Properties 8-38.1.4 Toughness Properties 8-108.1.5 Lamellar Discontinuities 8-108.1.6 K-Area Fractures 8-10

8.2 Welding 8-118.2.1 Welding Process 8-118.2.2 Welding Procedures 8-128.2.3 Welding Filler Metals 8-138.2.4 Preheat and Interpass Temperatures 8-178.2.5 Postheat 8-178.2.6 Controlled Cooling 8-178.2.7 Metallurgical Stress Risers 8-178.2.8 Welding Preparation & Fit-up 8-17

12. REFERENCES 12-1


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Introduction

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1. INTRODUCTION

1.1 Purpose

There are no modifications to the Guidelines or Commentary of Section 1.1 at this time.

1.2 Scope


1.3 Background

Following the January 17, 1994 Northridge, California Earthquake, more than 100 steel buildingswith welded moment-resisting frames were found to have experienced beam-to-column connectionfractures. The damaged structures cover a wide range of heights ranging from one story to 26 stories;and a wide range of ages spanning from buildings as old as 30 years of age to structures just beingerected at the time of the earthquake. The damaged structures are were spread over a largegeographical area, including sites that experienced only moderate levels of ground shaking. Althoughrelatively few such buildings were located on sites that experienced the strongest ground shaking,damage to these buildings was quite severe. Discovery of these extensive connection fractures, oftenwith little associated architectural damage to the buildings, was has been alarming. The discovery hasalso caused some concern that similar, but undiscovered damage may have occurred in other buildingsaffected by past earthquakes. Indeed, there are now confirmed isolated reports of such damage. Inparticular, a publicly owned building at Big Bear Lake is known to have been was damaged by theLanders-Big Bear, California sequence of earthquakes, and at least one building, under construction inOakland, California at the time fo the several buildings were damaged during the 1989 Loma PrietaEarthquake, was reported to have experienced such damage in the San Francisco Bay Area.

WSMF construction is used commonly throughout the United States and the world, particularlyfor mid- and high-rise construction. Prior to the Northridge Earthquake, this type of construction wasconsidered one of the most seismic-resistant structural systems, due to the fact that severe damage tosuch structures had rarely been reported in past earthquakes and there was no record of earthquake-induced collapse of such buildings, constructed in accordance with contemporary US practice.However, the widespread severe structural damage which occurred to such structures in theNorthridge Earthquake calleds for re-examination of this premise.

The basic intent of the earthquake resistive design provisions contained in the building codes is toprotect the public safety, however, there is also an intent to control damage. The developers of thebuilding code provisions have explicitly set forth three specific performance goals for buildingsdesigned and constructed to the code provisions (SEAOC - 1990). These are to provide buildings withthe capacity to

• resist minor earthquake ground motion without damage;


Introduction

1-2

• resist moderate earthquake ground motion without structural damage but possibly somenonstructural damage; and

• resist major levels of earthquake ground motion, having an intensity equal to the strongesteither experienced or forecast for the building site, without collapse, but possibly with somestructural as well as nonstructural damage.

In general, WSMF buildings in the Northridge Earthquake met the basic intent of the buildingcodes, to protect life safety. However, the ground shaking intensity experienced by most of thesebuildings was significantly less than that anticipated by the building codes. Many buildings thatexperienced moderate intensity ground shaking experienced significant damage that could be viewed asfailing to meet the intended performance goals with respect to damage control. Further, somemembers of the engineering profession (SEAOC - 1995b) and government agencies (Seismic SafetyCommission - 1995) have stated that even these performance goals are inadequate for society’s currentneeds.

WSMF buildings are designed to resist earthquake ground shaking based on the assumption thatthey are capable of extensive yielding and plastic deformation, without loss of strength. The intendedplastic deformation is intended to be developed through a combination of consists of plastic rotationsdeveloping within the beams, at their connections to the columns, and plastic shear yielding of thecolumn panel zones,. and is tTheoretically these mechanisms should be capable of resulting in benigndissipation of the earthquake energy delivered to the building. Damage is expected to consist ofmoderate yielding and localized buckling of the steel elements, not brittle fractures. Based on thispresumed behavior, building codes require a minimum lateral design strength for WSMF structures thatis approximately 1/8 that which would be required for the structure to remain fully elastic. Supplemental provisions within the building code, intended to control the amount of interstory driftsustained by these flexible frame buildings, typically result in structures which are substantially strongerthan this minimum requirement and in zones of moderate seismicity, substantial overstrength may bepresent to accommodate wind and gravity load design conditions. In zones of high seismicity, mostsuch structures designed to minimum code criteria will not start to exhibit plastic behavior until groundmotions are experienced that are 1/3 to 1/2 the severity anticipated as a design basis. This designapproach has been developed based on historical precedent, the observation of steel buildingperformance in past earthquakes, and limited research that has included laboratory testing of beam-column models, albeit with mixed results, and non-linear analytical studies.

Observation of damage sustained by buildings in the Northridge Earthquake indicates that contraryto the intended behavior, in some many cases brittle fractures initiated within the connections at verylow levels of plastic demand, and in some cases, while the structures remained essentially elastic. Typically, but not always, fractures initiated at, or near, the complete joint penetration (CJP) weldbetween the beam bottom flange and column flange (Figure 1-1). Once initiated, these fracturesprogressed along a number of different paths, depending on the individual joint and stress conditions. Figure 1-1 indicates just one of these potential fracture growth patterns. Investigators initially identifieda number of factors which may have contributed to the initiation of fractures at the weld root including:notch effects created by the backing bar which was commonly left in place following joint completion;sub-standard welding that included excessive porosity and slag inclusions as well as incomplete fusion;


Introduction

1-3

and potentially, pre-earthquake fractures resulting from initial shrinkage of the highly restrained weldduring cool-down. Such problems could be minimized in future construction, with the application ofappropriate welding procedures and more careful exercise of quality control during the constructionprocess. However, it is now known that these were not the only causes of the fractures whichoccurred.

Backing bar

Column flange

Beam flange

Fused zone

Fracture

Figure 1-1 - Common Zone of Fracture Initiation in Beam -Column Connection

Current production processes for structural steel shapes result in inconsistent strength anddeformation capacities for the material in the through-thickness direction. Non-metallic inclusions inthe material, together with anisotropic properties introduced by the rolling process can lead to lamellarweakness in the material. Further, the distribution of stress across the girder flange, at the connectionto the column is not uniform. Even in connections stiffened by continuity plates across the panel zone,significantly higher stresses tend to occur at the center of the flange, where the column web produces alocal stiffness concentration. Large secondary stresses are also induced into the girder flange tocolumn flange joint by kinking of the column flanges resulting from shear deformation of the columnpanel zone.

The dynamic loading experienced by the moment-resisting connections in earthquakes ischaracterized by high strain tension-compression cycling. Bridge engineers have long recognized thatthe dynamic loading associated with bridges necessitates different connection details in order to provideimproved fatigue resistance, as compared to traditional building design that is subject to “static”loading due to gravity and wind loads. While the nature of the dynamic loads resulting fromearthquakes is somewhat different than the high cycle dynamic loads for which fatigue-prone structuresare designed, similar detailing may be desirable for buildings subject to seismic loading.

In design and construction practice for welded steel bridges, mechanical and metallurgical notchesshould be avoided because they may be the initiators of fatigue cracking. As fatigue cracks grow underrepetitive loading, a critical crack size may be reached whereupon the material toughness (which is afunction of temperature) may be unable to resist the onset of brittle (unstable) crack growth. Thebeam-to-column connections in WSMF buildings are comparable to category C or D bridge details thathave a reduced allowable stress range as opposed to category B details for which special metallurgical,inspection and testing requirements are applied. The rapid rate of loading imposed by seismic events,and the complete inelastic range of tension-compression-tension loading applied to these connections is


Introduction

1-4

much more severe than typical bridge loading applications. The mechanical and metallurgical notchesor stress risers created by the beam-column weld joints are a logical point for fracture problems toinitiate. This, coupled with the tri-axial restraint provided by the beam web and the column flange, is arecipe for brittle fracture.

During the Northridge Earthquake, oOnce fractures initiated in beam-column joints, theyprogressed in a number of different ways. In some cases, the fractures initiated but did not grow, andcould not be detected by visual observation. In other cases, In many cases, the fractures progressedcompletely directly through the thickness of the weld, and if fireproofing was removed, the fractureswere evident as a crack through exposed faces of the weld, or the metal just behind the weld (Figure 1-2a). Other fracture patterns also developed. In some cases, the fracture developed into a surface thatresembled a through-thickness failure of the column flange material behind the CJP weld (Figure 1-2b). In these cases, a portion of the column flange remained bonded to the beam flange, but pulled freefrom the remainder of the column. This fracture pattern has sometimes been termed a “divot” or“nugget” failure.

A number of fractures progressed completely through the column flange, along a near horizontalplane that aligns approximately with the beam lower flange (Figure 1-3a). In some cases, thesefractures extended into the column web and progressed across the panel zone Figure (1-3b). Investigators have reported some instances where columns fractured entirely across the section.

a. Fracture at Fused Zone b. Column Flange “Divot” Fracture

Figure 1-2 - Fractures of Beam to Column Joints

a. Fractures through Column Flange b. Fracture Progresses into Column Web

Figure 1-3 - Column Fractures


Introduction

1-5

Once these fractures have occurred, the beam - column connection has experienced a significantloss of flexural rigidity and capacity. Residual flexural strength and rigidity must be developed througha couple consisting of forces transmitted through the remaining top flange connection and the webbolts. Initial rResearch suggests that residual stiffness is approximately 20% of that of the undamagedconnection and that residual strength varies from 10% to 40% of the undamaged capacity, whenloading results in tensile stress normal to the fracture plane. When loading produces compressionacross the fracture plane, much of the original strength and stiffness remain. However, in providingthis residual strength and stiffness, the beam shear connections can themselves be subject to failures,consisting of fracturing of the welds of the shear plate to the column, fracturing of supplemental weldsto the beam web or fracturing through the weak section of shear plate aligning with the bolt holes(Figure 1-4).

Figure 1-4 - Vertical Fracture through Beam Shear Plate Connection

It is now known that these fractures were the result of a number of complex factors that were notwell understood either when these connections were first adopted as a standard design approach, orwhen the damage was discovered immediately following the Northridge earthquake. Engineers hadcommonly assumed that when these connections were loaded to yield levels, flexural stresses in thebeam would be transferred to the column through a force couple comprised of nearly uniform yieldlevel tensile and compressive stresses in the beam flanges. It was similarly assumed that nearly all ofthe shear stress in the beam was transferred to the column through the shear tab connection to thebeam web. In fact, the actual behavior is quite different from this. As a result of local deformationsthat occur in the column at the location of the beam connection, a significant portion of the shear stressin the beam is actually transferred to the column through the beam flanges. This causes large localizedsecondary stresses in the beam flanges, both at the toe of the weld access hole and also in the completejoint penetration weld at the face of the column. The presence of the column web behind the columnflange tends to locally stiffen the joint of the beam flange to the column flange, further concentratingthe distribution of connection stresses and strains. Finally, the presence of the heavy beam and columnflange plates, arranged in a “+” shaped pattern at the beam flange to column flange joint produces acondition of very high restraint, which retards the onset of yielding, by raising the effective yieldstrength of the material, and allowing the development of very large stresses.


Introduction

1-6

The most severe stresses typically occur at the root of the complete joint penetration weld of thebeam bottom flange to the column flange. This is precisely the region of this welded joint that is mostdifficult for the welder to properly complete, as the access to the weld is restricted by the presence ofthe beam web and the welder often performs this weld while seated on the top flange, in the so-called“wildcat” position. The welder must therefore work from both sides of the beam web, starting andterminating the weld near the center of the joint, a practice that often results in poor fusion and thepresence of slag inclusions at this location. These conditions, which are very difficult to detect whenthe weld backing is left in place, as was the typical practice, are ready-made crack initiators. When thisregion of the welded joints is subjected to the large concentrated tensile stresses, the weld defects beginto grow into cracks and these cracks can quickly become unstable and propagate as brittle fractures. Once these brittle fractures initiate, they can grow in a variety of patterns, as described above, underthe influence of the stress field and the properties of the base and weld metals present at the zone of thefracture.

Despite the obvious local strength impairment resulting from these fractures, many damagedbuildings did not display overt signs of structural damage, such as permanent drifts or extreme damageto architectural elements. Until news of the discovery of connection fractures in some buildings beganto spread through the engineering community, it was relatively common for engineers to performcursory post-earthquake evaluations of WSMF buildings and declare that they were undamaged. Inorder to reliably determine if a building has sustained connection damage, it is necessary to removearchitectural finishes and fireproofing and perform nondestructive examination including visualinspection and ultrasonic testing careful visual inspection of the welded joints supplemented, in somecases, by nondestructive testing. Even if no damage is found, this is a costly process. Repair ofdamaged connections is even more costly. A few WSMF buildings have sustained so much connectiondamage that it has been deemed more practical to demolish the structures rather than to repair them. In the case of one WSMF building, damaged by the Northridge earthquake, repair costs weresufficiently large that the owner elected to demolish rather than replace than building.

Immediately following the Northridge Earthquake, a series of tests of beam-column subassemblieswere performed at the University of Texas at Austin, under funding provided by the AISC as well asprivate sources. The test specimens used heavy W14 column sections and deep (W36) beam sectionscommonly employed in some California construction. Initial specimens were fabricated using thestandard prequalified connection specified by the Uniform Building Code (UBC). Section 2211.7.1.2of UBC-94 {NEHRP-91 Section 10.10.2.3} specified this prequalified connection as follows:

“2211.7.1.2 Connection strength. The girder top column connection may be considered to be adequateto develop the flexural strength of the girder if it conforms to the following:

1. the flanges have full penetration butt welds to the columns.

2. the girder web to column connection shall be capable of resisting the girder shear determined for thecombination of gravity loads and the seismic shear forces which result from compliance with Section2211.7.2.1. This connection strength need not exceed that required to develop gravity loads plus3(Rw/8) times the girder shear resulting from the prescribed seismic forces.


Introduction

1-7

Where the flexural strength of the girder flanges is greater than 70 percent of the flexural strength ofthe entire section, (i.e. btf/(d-tf)Fy>0.7ZxFy) the web connection may be made by means of welding orhigh-strength bolting.

For girders not meeting the criteria in the paragraph above, the girder web-to-column connection shallbe made by means of welding the web directly or through shear tabs to the column. That welding shallhave a strength capable of developing at least 20 percent of the flexural strength of the girder web. Thegirder shear shall be resisted by means of additional welds or friction-type slip-critical high strength boltsor both.

and:

2211.7.2.1 Strength. The panel zone of the joint shall be capable of resisting the shear induced by beambending moments due to gravity loads plus 1.85 times the prescribed seismic forces, but the shearstrength need not exceed that required to develop 0.8ΣMs of the girders framing into the column flangesat the joint...”

In order to investigate the effects that backing bars and weld tabs had on connection performance,these were removed from the specimens prior to testing. Despite these precautions, the test specimensfailed at very low levels of plastic loading. Following these tests at the University of Texas at Austin,reviews of literature on historic tests of these connection types indicated a significant failure rate in pasttests as well, although these had often been ascribed to poor quality in the specimen fabrication. It wasconcluded that the prequalified connection, specified by the building code, was fundamentally flawedand should not be used for new construction in the future.

In retrospect, this conclusion may have been somewhat premature. More recent testing ofconnections having configurations similar to those of the prequalified connection, but incorporatingtougher weld metals, having backing bars removed from the bottom flange joint, and fabricated withgreater care to avoid the defects that can result in crack initiation, have performed better than thoseinitially tested at the University of Texas. However, as a class, when fabricated using currentlyprevailing construction practice, these connections still do not appear to be capable of consistentlydeveloping the levels of ductility presumed by the building codes for service in moment-resisting framesthat are subjected to large inelastic demands.When the first test specimens for that series werefabricated, the welder failed to follow the intended welding procedures. Further, no special precautionswere taken to assure that the materials incorporated in the work had specified toughness. Someengineers, with knowledge of fracture mechanics, have suggested that if materials with adequatetoughness are used, and welding procedures are carefully specified and followed, adequate reliabilitycan be obtained from the traditional connection details. Others believe that the conditions of high tri-axial restraint present in the beam flange to column flange joint (Blodgett - 1995) would prevent ductilebehavior of these joints regardless of the procedure used to make the welds. Further they point to theimportant influence of the relative yield and tensile strengths of beam and column materials, and othervariables, that can affect connection behavior. To date, there has not been sufficient researchconducted to resolve this issue.


Introduction

1-8

In reaction to the University of Texas tests as well as the widespread damage discovered followingthe Northridge Earthquake, and the urging of the California Seismic Safety Commission, in September,1994 the International Conference of Building Officials (ICBO) adopted an emergency code change tothe 1994 edition of the Uniform Building Code (UBC-94) {1994 NEHRP Recommended ProvisionsSection 5.2}. This code change, jointly developed by the Structural Engineers Association ofCalifornia, AISI and ICBO staff, deleted the prequalified connection and substituted the following in itsplace:

“2211.7.1.2 Connection Strength. Connection configurations utilizing welds or high-strengthbolts shall demonstrate, by approved cyclic test results or calculation, the ability to sustaininelastic rotation and develop the strength criteria in Section 2211.7.1.1 considering the effect ofsteel overstrength and strain hardening.”

“2211.7.1.1 Required strength. The girder-to-column connection shall be adequate to develop thelesser of the following:


2. The moment corresponding to development of the panel zone shear strength as determined fromformula 11-1.”

Unfortunately, neither the required “inelastic rotation”, or calculation and test procedures are welldefined by these code provisions. Design Advisory No. 3 (SAC-1995) included an InterimRecommendation (SEAOC-1995) that attempted to clarify the intent of this code change, and thepreferred methods of design in the interim period until additional research could be performed andreliable acceptance criteria for designs re-established. The State of California similarly published a jointInterpretation of Regulations (DSA-OSHPD - 1994) indicating the interpretation of the current coderequirements which would be enforced by the state for construction under its control. This appliedonly to the construction of schools and hospitals in the State of California. The intent of these InterimGuidelines is to supplement these previously published documents and to provide updatedrecommendations based on the results of the limited directed research performed to date.

1.4 The SAC Joint Venture


1.5 Sponsors


1.6 Summary of Phase 1 Research


1.7 Intent


Introduction

1-9


1.8 Limitations


1.9 Use of the Guidelines



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References

12-1

12. REFERENCES

ATLSS, Fractographic Analysis of Specimens from Failed Moment Connections, (publicationpending, title not exact)Fracture Analysis of Failed Moment Frame Weld Joints Produced in Full-Scale Laboratory Tests and Buildings Damaged in the Northridge Earthquake, SAC95-08, 1995.

ATLSS, Testing of Welded “T” Specimens, (publication pending, title not exact), SAC, 1995 AStudy of the Effects of Material and Welding Factors on Moment-Frame Weld JointPerformance Using a Small-Scale Tension Specimen. Kauffman, E.J., and Fisher, J.W., SAC95-08 1995.

Allen J., Personal Correspondence, Test Reports for New Detail, July 30, 1995.

Allen J., Partridge, J.E., and Richard, R.M., Stress Distribution in Welded/Bolted Beam toColumn Moment Connections. The Allen Company, March, 1995.

American Association of State Highway and Transportation Officials, Bridge Welding CodeAASHTO/AWS D1.5, 1995.

American Institute of Steel Construction, Seismic Provisions for Structural Steel Buildings, April,1997

American Institute of Steel Construction, Statistical Analysis of Charpy V-notch Toughness ForSteel Wide Flange Structural Shapes, July, 1995.

American Institute of Steel Construction, Manual of Steel Construction, ASD, Ninth Edition,1989.

American Institute of Steel Construction, Manual of Steel Construction, LRFD, Second Edition,1998.

American Institute of Steel Construction, Load and Resistance Factor Design Specification forStructural Steel Buildings, December 1, 1993.

American Institute of Steel Construction, Specification for Structural Joints using ASTM A325or A490 Bolts. 1985.

American Institute of Steel Construction, AISC Northridge Steel Update I, October, 1994.

American Welding Society, Guide for Nondestructive Inspection of Welds, AWS B1.10-86, 1986.

American Welding Society, Guide for Visual Inspection of Welds, AWS B1.11-88, 1988.

American Welding Society, Surface Roughness Guide for Oxygen Cutting, AWS C4.1-77, 1977.

American Welding Society, Structural Welding Code - Steel AWS D1.1-94, 1994.


References

12-2

American Welding Society, Structural Welding Code – Steel AWS D1.1-98, 1998

Anderson, J.C., Johnson, R.G., Partridge, J.E., “Post Earthquake Studies of A Damaged LowRise Office Building” Technical Report: Case Studies of Steel Moment Frame BuildingPerformance in the Northridge Earthquake of January 17, 1994 SAC 95-07. SAC, December,1995.

Anderson, J.C., Filippou, F.C., Dynamic Response Analysis of the 18 Story Canoga Building, SAC, March, 1995.

Anderson, J.C., Test Results for Repaired Specimen NSF#1, Report to AISC Steel AdvisoryCommittee, June, 1995.

Applied Technology Council, Earthquake Damage Evaluation Data for California ATC-13,Redwood City, CA 1985.

Applied Technology Counicl, Procedures for Post Earthquake Safety Evaluations of BuildingsATC-20, Redwood City, CA, 1989.

Applied Technology Council, Guidelines for Cyclic Seismic Testing of Components of SteelStructures, ATC-24, Redwood City, CA, 1992.

Astaneh-Asl, A. Post-Earthquake Stability of Steel Moment Frames with Damaged Connections. Proceedings of the Third International Workshop on Connections in Steel Structures, Universityof Trento, Trento, Italy, 1995.

Avent, R., “Designing Heat-Straightening Repairs,” National Steel Construction ConferenceProceedings, Las Vegas, NV, AISC, 1992.

Avent, R., “Engineered Heat Straightening,” National Steel Construction ConferenceProceedings, San Antonio, TX, AISC, 1995.

Barsom, J. M. and Korvink, S. A. “Through-thickness Properties of Structural Steels”,manuscript submitted to ASCE Journal of Structural Engineering, 1997.

Beck, J.L., May, B.S., Polidori, D.C., Vanik, M.W., “Ambient Vibration Surveys of Three Steel-Frame Buildings Strongly Shaken by the 1994 Northridge Earthquake”, Analytical and FieldInvestigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC, December, 1995.

Bertero, V.V., and Whittaker, A. and Gilani, A., Testing of Repaired Welded Beam ColumnAssembliesSeismic Tesing of Full-Scale Steel Beam-Column Assemblies, SAC96-01, publicationpending (title not exact), 1995X1996.

Blodgett, O., “Evaluation of Beam to Column Connections”, SAC Steel Moment FrameConnection Advisory No. 3, Feb. 1995.


References

12-3

Bonowitz, D, and Youssef, N. “SAC Survey of Steel-Moment Frames Affected by the 1994Northridge Earthquake”, Surveys and Assessment of Damage to Buildings Affected by theNorthridge Earthquake of January 17, 1994 SAC 95-06, SAC, 1995.

Building Seismic Safety Council. NEHRP Recommended Provisions for Seismic Regulations forNew Buildings -1991 Edition FEMA 222, (Commentary FEMA 223), Washington D.C., January,1992.

Building Seismic Safety Council. NEHRP Recommended Provisions for Seismic Regulations forNew Buildings -1994 Edition FEMA 222A, (Commentary FEMA223A), Washington D.C., July,1995.

Building Seismic Safety Council. NEHRP Recommended Provisions for Seismic Regulations forNew Buildings and Other Structures. – 1997 Edition, FEMA 302, (Commentary FEMA303),Washington, D.C., February, 1998

Campbell, K.W. and Bazorgnia, Y., “Near Source Attentuation of Peak Horizontal Accelerationfrom World Wide Accelerogram Records from 1957 - 1993,” Proceedings of the Fifth NationalConference on Earthquake Engineering, Chicago, Ill, 1994.

Campbell, S., “Modeling of Weld Fractures Using the Drain Programs”, Technical Report:Parametric Analytical Investigations of Ground Motion and Structural Response, NorthridgeEarthquake of January 17, 1994 SAC95-05. SAC, 1995.

Chen, S.J. and Yeh, C.H., Enhancement of Ductility of Steel Beam-to-Column Connections forSeismic Resistance, Department of Construction Engineering, National Taiwan University, May,1995.

Diererlein, G. “Summary of Building Analysis Studies” Analytical and Field Investigations ofBuildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC,December, 1995

Durkin, M. E., “Inspection, Damage, and Repair of Steel Frame Buildings Following theNorthridge Earthquake”, Technical Report: Surveys and Assessment of Damage to BuildingsAffected by the Northridge Earthquake of January 17, 1994 SAC 95-06, SAC, December, 1995.

Engelhardt, M.D., and Sabol, T.A. Testing of Welded Steel Moment Connections In Response tothe Northridge Earthquake, Progress Report to the AISC Advisory Subcommittee on SpecialMoment Resisting Frame Research, October, 1994.

Engelhardt, M. D., Keedong, K.M. Sabol T. A., Ho, L., Kim, H. Uzarski, J. and Abunnasar, H. “Analysis of a Six Story Steel Moment Frame Building in Santa Monica”, Analytical and FieldInvestigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1 SAC, December, 1995.


References

12-4

Engelhardt, M. D., Keedong, K.M., Uzarski, J., Abunassar, H., Sabol, T.A., Ho, L., and Kim, H.“Parametric Studies on Inelastic Modeling of Steel Moment Frames”, Technical Report:Parametric Analytical Investigations of Ground Motion and Structural Response, NorthridgeEarthquake of January 17, 1994 SAC95-05. SAC, 1995.

Engelhardt, M.D., Sabol, T. A., and Shuey, B.D. Testing of Repair Concepts for Damaged SteelMoment Connections.et. al. Testing of Repaired Welded Beam Column Assemblies, SAC96-01,publication pending (title not exact), 19951996.

Englehardt, M.D. Fowler, T.J., and Barnes, C.A., Acoustic Emission Monitoring of Welded SteelMoment Connection Tests.et. al. Accoustic Emission Recordings for Welded Beam ColumnAssembly Tests, SAC95-08, publication pending (title not exact), 1995.

Frank, K.H. “The Physical and Metallurgical Properties Of Structural Steels” State of Art Papers:Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System BehaviorSAC 95-09. SAC, September, 1996

Fillippou, F.C. “Nonlinear Static and Dynamic Analysis of Canoga Park Towers with FEAP-STRUC”, Analytical and Field Investigations of Buildings Affected by the NorthridgeEarthquake of January 17, 1994, SAC 95-04 Part 2, SAC., December, 1995.

Fisher, J.W., Dexter, R.J., and Kauffman, E.J., “Fracture Mechanics of Welded Structural SteelConnections.” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, MomentConnections and Frame System Behavior SAC 95-09. SAC, September, 1996

Forrel/Elsesser Engineers, Inc., Lawrence Berkeley National Labs Steel Joint Test - TechnicalBrief, San Francisco, CA, July 17, 1995.

Gates, W.E., and Morden, M., “Lessons from Inspection, Evaluation, Repair and Construction ofWelded Steel Moment Frames Following the Northridge Earthquake”, Surveys and Assessment ofDamage to Buildings Affected by the Northridge Earthquake of January 17, 1994 SAC 95-06SAC, December, 1995.

Gates, W.E. “Interpretation of SAC Survey Data on Damaged Welded Steel Moment FramesFollowing the Northridge Earthquake”, Surveys and Assessment of Damage to Buildings Affectedby the Northridge Earthquake of January 17, 1994 SAC 95-06, SAC, December, 1995.

Green, M. “Santa Clarita City Hall; Northridge Earthquake Damage” Technical Report: CaseStudies of Steel Moment Frame Building Performance in the Northridge Earthquake of January17, 1994 SAC 95-07. SAC, December, 1995.

Hall, J.F., “Parameter Study of the Response of Moment-Resisting Steel Frame Buildings toNear-Source Ground Motions”, Technical Report: Parametric Analytical Investigations ofGround Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC95-05. SAC, 1995.


References

12-5

Hajjar, J.F., O’Sullivan D.P., Leon, R. T., Gourley, B.C. “Evaluation of the Damage to the BoraxCorporate Headquarters Building As A Result of the Northridge Earthquake”, Technical Report:Case Studies of Steel Moment Frame Building Performance in the Northridge Earthquake ofJanuary 17, 1994 SAC 95-07. SAC, December, 1995.

Harrison, P.L. and Webster, S.E., Examination of Two Moment Resisting Frame ConnectorsUtilizing a Cover-Plate Design, British Steel Technical, Swinden Laboratories, Moorgate,Rotherham, 1995.

Hart, G.C., Huang, S.C., Lobo, R.F., Van Winkle, M., Jain, A., “Earthquake Response ofStrengthened Steel Special moment Resisting Frames” Analytical and Field Investigations ofBuildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC.,December, 1995

Hart, G.C., Huang, S., Lobo, R., and Stewart, J., “Elastic and Inelastic Analysis for Weld FailurePrediction of Two Adjacent Steel Buildings”, ” Analytical and Field Investigations of BuildingsAffected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC, December,1995.

Hart, G.C., Huang, S., Lobo, R., and Stewart, J., “Influence of Vertical Ground Motion onSpecial Moment-Resisting Frames”, Technical Report: Parametric Analytical Investigations ofGround Motion and Structural Response, Northridge Earthquake of January 17, 1994 SAC95-05. SAC, 1995.

Heaton, T.H., Hall, J.F., Wald, D.J., and Halling, M.W. “Response of High-Rise and Base-Isolated Buildings to a Hypothetical Mw 7.0 Blind Thrust Earthquake” Science Vol. 26, pp 206-211, January, 1995.

International Conference of Building Officials, Uniform Building Code UBC-97, Whittier, CA,1997.

International Conference of Building Officials, Uniform Building Code UBC-94. Whittier, CA,1994.

Iwan, W.D., “Drift Demand Spectra for Selected Northridge Sites”, Technical Report:Parametric Analytical Investigations of Ground Motion and Structural Response, NorthridgeEarthquake of January 17, 1994 SAC95-05. SAC, 1995.

Joyner, W.B., and Boore, D.M., “Ground Motion Parameters for Seismic Design,”Bulletin of theSesimological Society of America, 1994.

Kariotis, J. and Eimani, T.J., “Analysis of a Sixteen Story Steel Frame Building at Site 5, for theNorthridge Earthquake”, Analytical and Field Investigations of Buildings Affected by theNorthridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC, December, 1995.


References

12-6

Krawinkler, H.K., “Systems Behavior of Structural Steel Frames Subjected to EarthquakeGround Motions” State of Art Papers: Metallurgy, Fracture Mechanics, Welding, MomentConnections and Frame System Behavior SAC 95-09. SAC, September, 1996

Krawinkler, H.K., Ali, A.A., Thiel, C.C., Dunlea, J.M., “Analysis of a Damaged 4-Story Buildingand an Undamaged 2- Story Building”, Analytical and Field Investigations of Buildings Affectedby the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC, December, 1995.

Ksai, K. , and Bleiman, D. “Bolted Brackets for Repair of Damaged Steel Moment FrameConnections,” 7th U.S.-Japan Workshop on the Improvement of Structural Design andConstruction Practices: Lessons Learned from Northridge and Kobe, Kobe, Japan, January, 1996

Leon, R. T., “Seismic Performance of Bolted and Riveted Connections” State of Art Papers:Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame System BehaviorSAC 95-09. SAC, September, 1996

Miller, D.K. “Welding of Seismically Resistant Steel Structures” State of Art Papers: Metallurgy,Fracture Mechanics, Welding, Moment Connections and Frame System Behavior SAC 95-09.SAC, September, 1996

Naeim F., DiJulio, R., Benuska, K., Reinhorn, A. M., and Chen, L. “Evaluation of SeismicPerformance of an 11 Story Steel Moment Frame Building During the 1994 NorthridgeEarthquake”, ” Analytical and Field Investigations of Buildings Affected by the NorthridgeEarthquake of January 17, 1994, SAC 95-04 Part 2 SAC, December, 1995.

Newmark, N.M. and Hall W.J., Earthquake Spectra and Design. Earthquake EngineeringResearch Institute, 1982.

NIST and AISC. Modification of Existing Welded Steel Moment Frame Connections for SeismicResistance. National Institute of Standards and Technology and American Institute of SteelConstruction. 1999

Paret, T.F., Sasaki, K.K., “Analysis of a 17 Story Steel Moment Frame Building Damaged by theNorthridge Earthquake”, Analytical and Field Investigations of Buildings Affected by theNorthridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC, December, 1995.

Popov, E.P. and Yang, T.S. Steel Seismic Moment Resisting Connections. University ofCalifornia at Berkeley, May, 1995.

Popov, E.P. Blondet, M., Stepanov, L, and Stodjadinovic, B. Full-Scale Beam-ColumnConnection Tests. et. al. Testing of Repaired Welded Beam Column Assemblies, SAC,publication pending (title not exact), 1995 SAC 96-01. 1996..

SAC, Proceedings of the International Workshop on Steel Moment Frames, October 23-24, 1994SAC-94-01. Sacramento, CA, December, 1994.


References

12-7

SAC . Steel Moment Frame Advisory No. 1. September, Sacramento, CA, 1994.

SAC . Steel Moment Frame Advisory No. 2. October, Sacramento, CA, 1994.

SAC . Steel Moment Frame Advisory No. 3 SAC-95-01, February, Sacramento, CA, 1995.

Shonafelt, G.O., and Horn, W.B.. Guidelines for Evaluation and Repair of Damaged SteelBridge Members, NCHRP Report 271, Transportation Research Board, 1984.

Skiles, J.L. and Campbell, H.H., “Why Steel Fractured in the Northridge Earthquake” SACAdvisory No. 1, October, 1994.

Seismic Safety Commission, Northridge Earthquake Turning Loss to Gain, Report to theGovernor, Sacramento, CA, 1995.

Smith Emery Company. Report of Test, July, 1995.

Sommerville, P, Graves, R., Chandan, S. Technical Report: Characterization of Ground MotionDuring the Northridge Earthquake of January 17, 1994, SAC 95-03, SAC, December, 1995.

State of California. Division of the State Architect (DSA) and Office of Statewide HealthPlanning and Development (OSHPD). Interpretation of Regulations Steel Moment ResistingFrames, Sacramento, CA, 1994.

Structural Engineers Association of California (SEAOC), Seismology Committee, RecommendedLateral Force Requirements and Commentary, Sacramento, CA. 1990.

Structural Engineers Association of California (SEAOC), Seismology Committee, InterimRecommendations for Design of Steel Moment Resisting Connection,. Sacramento, CA, January,1995.

Structural Engineers Association of California (SEAOC), Vision 2000: A Framework forPerformance Based Engineering of Buildings, Sacramento, CA, April, 1995.

Structural Shape Producers Council, Statistical Analysis of Tensile Data for Wide FlangeStructural Shapes, 1994.

Thiel, C.C., and Zsutty, T.C., “Earthquake Characteristics and Damage Statistics,” EarthquakeSpectra, Volume 3, No. 4., Earthquake Engineering Research Institute, Oakland, Ca. 1987.

Tremblay, R., Tchebotarev, N., and Filiatrault, A., “Seismic Performance of RBS Connections forSteel Moment Resisting Frames: Influence of Loading Rate and Floor Slab,” Proceedings of theSecond International Conference on the Behavior of Steel Structures in Seismic Area, Kyoto,Japan, August, 1997


References

12-8

Tsai, K.C. and Popov, E. P. “Seismic Steel Beam-Column Moment Connections” State of ArtPapers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame SystemBehavior SAC 95-09. SAC, September, 1996

Uang, C.M. and Latham, C.T. Cyclic Testing of Full-Scale MNH-SMRF Moment Connections,Structural Systems Research, University of California, San Diego, March, 1995.

Tsai, K.C. and Popov, E.P., Steel Beam - Column Joints In Seismic Moment Resisting Frames,Report No. UCB/EERC-88/19, Earthquake Engineering Research Center, University ofCalifornia, Berkeley, Nov., 1988.

Uang, C.M., Yu, Q.S., Sadre, A., Bonowitz, D., Youssef, N. “Performance of a 13 Story SteelMoment-Resisting Frame Damaged in the 1994 Northridge Earthquake”, ” Analytical and FieldInvestigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2 SAC, December, 1995.

Uang, C.M. and Bondad, D. Progress Report on Cyclic Testing of Three Repaired UCSDSpecimens, SAC, 1995.

Uang, C.M. and Lee, C.H. “Seismic Response of Haunch Repaired Steel SMRFs: AnalyticalModelling and Case Studies” ” Analytical and Field Investigations of Buildings Affected by theNorthridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC., December, 1995

Wald, D.J., Heaton, T.H., and Hudnut, K.W., The Slip History of the 1994 Northridge,California, Earthquake Determined from Strong-Motion, Teleseismic, GPS, and Leveling Data,United Sates Geologic Survey, 1995.

Watabe, M. Peformance of Wooden Houses and Steel Buildings during the Great HanshinEarthquake, Architectural Institute of Japan, May, 1995.

Youssef, N.F.G, Bonowitz, D., and Gross, J.L., A Survey of Steel Moment-Resisting FrameBuildings Affected by the 1994 Northridge Earthquake, NISTR 5625, Gaithersburg Md, April,1995.


Classifications and Implications of Damage

3-1

3. CLASSIFICATION AND IMPLICATIONS OF DAMAGE

3.1 Summary of Earthquake Damage


3.2 Damage Types


3.2.1 Girder Damage

There are no modifications to the Guidelines or Commentary of Section 3.2.1 at this time.

3.2.2 Column Flange Damage


3.2.3 Weld Damage, Defects and Discontinuities

Six types of weld discontinuities, defects and damage are defined in Table 3-3 and illustratedin Figure 3-4. All apply to the complete joint penetration (CJP) welds between the girder flangesand the column flanges. This category of damage was the most commonly reported typefFollowing the Northridge Earthquake, many instances of W1a and W1b conditions were reportedas damage. These conditions, which are detectable only by ultrasonic testing or by removal ofweld backing, are now thought more likely to be construction defects than damage.

Table 3-3 - Types of Weld Damage, Defects and Discontinuities

Type DescriptionW1 Weld root indications

W1a Incipient indications -– depth <3/16” ortf/4; width < bf/4

W1b Root indications larger than that for W1aW2 Crack through weld metal thicknessW3 Fracture at column interfaceW4 Fracture at girder flange interfaceW5 UT detectable indication - non-rejectable



3-2

W1, W5W2

W3W4

Note: See Figure 3-2 for related column damage and Figure 3-3 for girder damageFigure 3-4 - Types of Weld Damage

Commentary: Despite significant controversy, type W1 and W5 discovered inbuildings following the Northridge earthquake, were commonly reported asdamage. These small discontinuities and defects located at the roots of the CJPwelds are detectable only by ultrasonic testing (UT) when the weld backing is leftin place or by visual testing (VT) or magnetic particle testing (MT) when weldbacking is removed. It now seems likely that most such conditions are notdamage at all, but rather, are pre-existing construction defects. A number offactors point to this conclusion. First, statistical surveys of damage sustained bybuildings in the Northridge earthquake show that if type W1 and W5 conditionsare not considered, there was a much greater incidence of damage in framesresisting north-south ground shaking than in frames resisting east-west shaking.This appears to be correlated with the relative strength of the ground shakingexperienced along these two directional axes. However, there is no significantdifference between the incidence rate of reported W1 and W5 conditions in thesetwo directions, suggesting that these conditions are not correlated with shakingintensity.

The discovery of W1 conditions in welds for which original constructionquality assurance documentation is available, indicating that no such defectswere present when the building was originally constructed, tends to contradictthis argument. However, investigations conducted by SAC under the Phase 2project have indicated that as a result of the joint geometry, UT techniques areoften unable to detect W1 conditions at the weld root, when scanning of the jointis conducted from the top surface of the beam bottom flange. It is important tonote that this is the most common method of conducting UT as part ofconstruction quality assurance. When UT scanning of a joint is conducted fromthe bottom surface of the flange, as is commonly done when inspecting for



3-3

earthquake damage, it becomes more likely that such conditions will be detected,since the geometric constraints present for top flange scanning are altered. Thisleads to the conclusion that it is probable that typical construction qualityassurance UT of welded joints would be likely to miss W1 conditions, allowingthem to be discovered in later post-earthquake surveys.

When FEMA-267 was first published, it was recommended that W1 conditionsbe treated as damage and that UT be used as a routine part of the post-earthquake investigation process, in order to discover these conditions. However,more recent investigations conducted by SAC have revealed that even the carefulscanning typically conducted as part of a post-earthquake inspection is not ableto reliably detect these conditions. Given that it is both expensive and difficult tolocate W1 conditions as part of a post-earthquake investigation, and also, thatmost of these conditions are unlikely to be damage at all, it is no longerrecommended that exhaustive investigations for these conditions be conducted aspart of the earthquake damage investigation process.

Type W1 damage, discontinuities and defects and type W5 discontinuities aredetectable only by NDT, unless the backing bar is removed, allowing directdetection by visual inspection or magnetic particle testing. Type W5 consists ofsmall discontinuities and may or may not actually be earthquake damage. AWSD1.1 permits small discontinuities in welds. Larger discontinuities are termeddefects, and are rejectable per criteria given in the Welding Code. It is likelytherefore that some weld indications detected by NDT in a post-earthquakeinspection may be discontinuities which pre-existed the earthquake and do notconstitute a rejectable condition, per the AWS standards. Repair of thesediscontinuities, designated as type W5 is not generally recommended. Some typeW1 indications are small planar defects, which are rejectable per the AWS D1.1criteria, but are not large enough to be classified as one of the types W2 throughW4. Type W1 is the single most commonly reported non-conforming conditionreported in the post-Northridge statistical data survey, and in some structures,represents more than 80 per cent of the total damage reported. The W1classification is split into two types, W1a and W1b, based on their severity. TypeW1a “incipient” root indications are defined as being nominal in extent, less than3/16” deep or 1/4 of the flange thickness, whichever is less, and having a lengthless than 1/4 of the flange width. Some engineers believe that type W1aindications are not earthquake damage at all, but rather, previously undetecteddefects from the original construction process. A W1b indication is one thatexceeds these limits but is not clearly characterized by one of the other types. Itis more likely that W1b indications are a result of the earthquake than theconstruction process.

As previously stated, some engineers believe that both type W1a and sometype W1b conditions are not earthquake related damage at all, but instead, are



3-4

rejectable conditions not detected by the quality control and assurance programsin effect during the original construction. However, in recent large-scale sub-assembly testing of the inelastic rotation capacity of girder-column connectionsconducted in SAC Phase 1 at the University of Texas at Austin and theEarthquake Engineering Research Center of the University of California atBerkeley, it was reported that significantly more indications were detectable inunfailed CJP welds following the testing than were detectable prior to the test.This tends to indicate that type W1 damage may be related to stresses induced inthe structures by their response to the earthquake ground motions. Regardless ofwhether or not type W1 conditions are directly attributable to earthquakeresponse, it is clear that these conditions result in a reduced capacity for the CJPwelds and can act as stress risers, or notches, to initiate fracture in the event offuture strong demands.

Type W2 fractures extend completely through the thickness of the weld metaland can be detected by either MT or VI techniques. Type W3 and W4 fracturesoccur at the zone of fusion between the weld filler metal and base material of thegirder and column flanges, respectively. All three types of damage result in aloss of tensile capacity of the girder flange to column flange joint and should berepaired.

As with girder damage, damage to welds has most commonly been reported atthe bottom girder to column connection, with fewer instances of reported damageat the top flange. Available data indicates that approximately 25 per cent of thetotal damage in this category occurs at the top flange, and most often, top flangedamage occurs in connections which also have bottom flange damage. For thesame reasons previously described for girder damage, less weld damage may beexpected at the top flange. However, it is likely that there is a significant amountof damage to welds at the top girder flange which have never been discovered dueto the difficulty of accessing this joint. Later sections of these Interim Guidelinesprovide recommendations for situations when such inspection should beperformed.

3.2.4 Shear Tab Damage


3.2.5 Panel Zone Damage


3.2.6 Other Damage




3-5

3.3 Safety Implications

The implications of the damage described above with regard to building safety are discussed inthis section. As part of the SAC Phase 2 program, extensive nonlinear analyses have beenconducted of WSMF buildings to determine the effects of connection fractures on buildingperformance and also to develop an understanding of the risk of earthquake-induced buildingcollapse. These studies indicate that risk of collapse of WSMF buildings designed to modernstandards and having connections capable of ductile behavior is quite low. Even in regions ofvery high seismicity, such as those areas of coastal California adjacent to major active faults, theprobability that such a building would experience earthquake-induced collapse appears to be onthe order of one occurrence per building, every 20,000 years. For buildings that have brittleconnections such as those commonly constructed prior to 1994, the probability of collapseincreases somewhat. If only the bottom flange connections of beams to columns is subject tofracture, the risk of global collapse of buildings increases to perhaps one occurrence in 15,000years, presuming that the fractures do not jeopardize column capacity. However, if both flangesof the connections are subject to fracture, or if substantial column damage occurs, the risk ofcollapse increases significantly. Also, it is important to note that severe connection fractures canresult in significant risk of local collapse and life safety endangerment.

While these studies have been helpful in providing an understanding of the level of riskinherent in WSMF structures with brittle connections, they do not provide sufficient informationto There is insufficient knowledge at this time to permit determination of the assess the degree ofrisk with any real confidence. However, based on the historic performance of modern WSMFbuildings, typical of those constructed in the United States, it appears that the risk of collapse inmoderate magnitude earthquakes, ranging up to perhaps M7, is very low for buildings which havebeen properly designed and constructed according to prevailing standards. A possible exceptionto this may be buildings located in the near field (< 10 km from the surface projection of the faultrupture) of such earthquakes (Heaton, et. al. - 1995), however, this is not uniquely a problemassociated with steel buildings. Our current building codes in general, may not be adequate toprovide for reliable performance of buildings within the near field of large earthquakes. As is alsothe case with all other types of construction, buildings with incomplete lateral force resistingsystems, severe configuration irregularities, inadequate strength or stiffness, poor constructionquality, or deteriorated condition are at higher risk than buildings not possessing thesecharacteristics.

No modern WSMF buildings have been sited within the areas of very strong ground motionfrom earthquakes larger than M7, or for that matter, within the very near field for eventsexceeding M6.5. This style of construction has been in wide use only in the past few decades.Consequently, it is not possible to state what level of risk may exist with regard to buildingresponse to such events. This same lack of performance data for large magnitude, long durationevents exists for virtually all forms of contemporary construction. Consequently, there isconsiderable uncertainty in assigning levels of risk to any building designed to minimum coderequirements for these larger events.



3-6

Commentary: Research conducted to date has not been conclusive with regard tothe risk of collapse of WSMF buildings. Some testing of damaged connectionsfrom a building in Santa Clarita, California have been conducted at theUniversity of Southern California (Anderson - 1995). In these tests, connectionassemblies which had experienced type P6 damage were subjected to repeatedcycles of flexural loading, while the column was maintained under axialcompression. Under these conditions, the specimens were capable of resisting asmuch as 40 per cent of the nominal plastic strength of the girder for severalcycles of slowly applied loading, at plastic deformation levels as large as 0.025radians. However, damage did progress in the specimen, as this testing wasperformed. It is not known how these assemblies would have performed if thecolumns were permitted to experience tensile loading. Data from other testssuggests that the residual strength of connections which have experienced typesG1, G4, W2, W3, and W4 damage is on the order of 15 per cent of theundamaged strength. Some analytical research (Hall - 1995) in which nonlineartime history analyses simulating the effects of connection degradation due tofractures were included, indicates that typical ground motions resulting in thenear field of large earthquakes can cause sufficient drift in these structures toinduce instability and collapse. Other researchers (Astaneh - 1995) suggest thatdamaged structures, even if unrepaired, have the ability to survive additionalground motion similar to that of the Northridge Earthquake.

Even though there were no collapses of WSMF buildings in the 1994Northridge Earthquake, it should not be assumed that no risk of such collapseexists. Indeed, a number of WSMF buildings did experience collapse in the 1995Kobe Earthquake. The detailing of these collapsed Japanese buildings wassomewhat different than that found in typical US practice, however, much of thefracture damage that occurred was similar to that discovered following theNorthridge event.

Because of a lack of data and experience with the effects of larger, longerduration earthquakes, there is considerable uncertainty about the performance ofall types of buildings in large magnitude seismic events. It is believed thatseismic risks in such large events are highly dependent on the individual groundmotion at a specific site and the characteristics of the individual buildings.Therefore, generalizations with regard to the probable performance of individualtypes of construction may not be particularly meaningful.

The risks to occupants of WSMF buildings with brittle connections is regardedas less, in most cases, than to occupants of the types of buildings listed below.However, because of the uncertainties involved, the degree of risk in large eventscannot be definitively quantified, nor can it categorically be stated that properlyconstructed WSMF buildings sited in the near field of large events are either



3-7

more or less at risk than many other code designed building systems which do notappear on the following list:

• Concentric braced steel frames with bracing connections that are weaker than thebraces

• Knee braced steel frames

• Unreinforced masonry bearing wall buildings

• Non-ductile reinforced concrete moment frames (infilled or otherwise)

• Reinforced concrete moment frames with gravity load bearing elements that werenot designed to participate in the lateral force resisting system and that do nothave capacity to withstand earthquake-induced deformations

• Tilt-up and reinforced masonry buildings with inadequate anchorage of theirheavy walls to their horizontal wood diaphragms

• Precast concrete structures without adequate interconnection of their structuralelements.

In addition, WSMF structures with brittle connections would appear to havelower inherent seismic risk than structures of any construction type that:

• do not having complete, definable load paths

• have significant weak and/or soft stories

• have major torsional irregularity and insufficient stiffness and strength to resistthe resulting seismic demands

• minimal redundancy and concentrations of lateral stiffness

These are general statements that represent a global view of systemperformance. As with all seismic performance generalizations, there are manysteel moment frame buildings that are more vulnerable to damage than someindividual buildings of the general categories listed, just as there are many thatwill perform better.

3.4 Economic Implications




3-8


Interim Guidelines Advisory No. 2 SAC99-01

Post Earthquake Evaluation

4-1

4. POST-EARTHQUAKE EVALUATION

4.1 Scope


4.2 Preliminary Evaluation


4.2.1 Evaluation Process

Preliminary evaluation is the process of determining if a building should be subjected todetailed post-earthquake evaluations. Detailed evaluations should be performed for all buildingsthought to have experienced strong ground motion, as indicated in Section 4.2.1.1 or for whichthe other indicators of Section 4.2.1.2 apply. Detailed post-earthquake evaluations include theentire process of determining if a building has experienced significant damage and if damage isfound, determining appropriate strategies for occupancy, structural repair and/or modification.Except as indicated in Section 4.2.3, detailed evaluation should, as a minimum, includeinspections of a representative sample of moment-resisting (and other type) connections withinthe building.

4.2.1.1 Ground Motion

There are no modifications to the Guidelines or Commentary of Section 4.2.1.1 at this time.

4.2.1.2 Additional Indicators


4.2.2 Evaluation Schedule

There are no modifications to the Guidelines of Section 4.2.2 at this time.

Commentary: It is important to conduct post-earthquake evaluations as soonfollowing the earthquake as is practical. Aftershock activity in the monthsimmediately following an earthquake is likely to produce additional strongground motion at the site of a damaged building. If there is adequate reason toassume that damage has occurred, then such damage should be expeditiouslyuncovered and repaired. However, since adequate resources for post-earthquakeevaluation may be limited, a staggered schedule is presented, with those buildingshaving a greater likelihood of damage recommended for evaluation first.



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Large magnitude earthquakes are often followed by large magnitudeaftershocks. Therefore, it is particularly urgent that post-earthquake evaluationsbe performed expeditiously following such events. If insufficient resources areavailable in the affected region to perform the NDT tests recommended by theGuidelines of Chapter 5, it is recommended that visual inspection, in accordancewith Section 5.2.2, proceed as soon as possible. If visual inspection revealssubstantial damage, consideration should be given to vacating the building untileither an adequate period of time has passed so as to make the likelihood of verylarge aftershocks relatively low (e.g. 4 weeks for magnitude 7 and lower, and 8weeks for magnitudes above this), complete inspections and repairs are made, ora detailed evaluation indicates that the structure retains adequate structuralstiffness and strength to resist additional strong ground shaking. Preliminaryvisual inspections should not be used as an alternative to complete evaluation.

The table Table 4-1relates the urgency for post-earthquake buildingevaluation to both the magnitude of the earthquake and the estimated peakground acceleration experienced by the building site. This is because largemagnitude events are more likely to have large magnitude aftershocks andbecause buildings that experienced stronger ground accelerations are more likelyto have been damaged. Except in regions with extensive strong motioninstrumentation, estimates of ground motion are quite subjective. Followingmajor damaging earthquakes, government agencies usually produce groundmotion maps showing projected acceleration contours. These maps should beused when available. When such maps are not available, ground motions can beestimated using any of several attenuation relationships that have been published.

4.2.3 Connection Inspections

Except as indicated in Sections 4.2.3.1 and 4.2.3.2, below, Ddetailed evaluations shouldinclude inspection of the building’s moment-resisting connections in order to determine theircondition. As a first pass, inspections may be limited to careful visual inspection of the joint ofthe beam bottom flange to the column. When such inspection reveals the presence of connectiondamage, a more thorough inspection of the damaged connection should be conducted. Sincemoment-resisting frame buildings commonly have many connections, inspections can be quitecostly. Therefore, it shall be permissible to limit inspections toof a representative sample ofWSMF (and other) connections, except as indicated in Sections 4.2.3.1 and 4.2.3.2, below.Section 4.3.3 provides three alternative approaches to selecting an appropriate sample ofconnections for inspection.

4.2.3.1 Analytical Evaluation




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4.2.3.2 Buildings with Enhanced Connections


4.2.4 Previous Evaluations and Inspections


4.3 Detailed Evaluation Procedure

Where detailed evaluation is recommended by Section 4.2, assessment of the post earthquakecondition of a building, its ability to resist additional strong ground motion and other loads, anddetermination of appropriate occupancy, structural repair and/or modification strategies should bebased on the results of a detailed inspection and assessment of the extent to which structuralsystems have been damaged.

In order to obtain complete data on a building’s post-earthquake condition, it is necessary toinspect each of the building’s moment-resisting frame elements and their connections. However,such extensive inspections could be very costly. As an alternative to that approach, this Sectionpresents a series of procedures by which a representative sample of beam-column connections isselected and inspected. This Section presents one approach for making such assessments. In thisapproach, the results of the sample inspections are used to calculate a cumulative damage index,D, for the structure as well as the probability that if all of the building’s connections had beeninspected, the damage index at any floor of the structure has would have been found to exceededa value of 1/3. General occupancy, structural repair and modification recommendations are madebased upon the values calculated for these damage indices. In particular, a calculated damageindex of 1/3 is used to indicate, in the absence of more detailed analyses, that a potentiallyhazardous condition may exist.

The structural engineer may use other procedures consistent with the principles of statisticsand structural mechanics to determine the residual strength and stiffness of the structure in the as-damaged state and the acceptability of such characteristics relative to the criteria contained in thebuilding code, or other rational criteria acceptable to the building official.

There are no modifications to the Commentary of Section 4.3 at this time.

4.3.1 Eight Step Evaluation Procedure

Post-earthquake evaluation should be carried out under the direct supervision of a structuralengineer. The following eight-step procedure may be used to determine the condition of thestructure and to develop occupancy, repair and modification strategies. Note that this procedureis written presuming that inspection is limited to a representative sample of the total number ofconnections present in the building. If all connections in the building are to be inspected, steps 1,2, 4 and 6 may be omitted.



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Step 1: The moment-resisting connections in the building are categorized into two or more“groups” (Section 4.3.2 and 4.4) comprised of connections expected to have similarprobabilities of being damaged.

Complete steps 2 through 7 below, for each group of connections.

Step 2: Determine the minimum number of connections in each group that should be inspectedand select the specific sample of connections to be inspected. (Section 4.3.3)

Step 3: Inspect the selected set of connections using the technical guidelines of Section 5.2.and determine connection damage indices, dj, for each inspected connection (Section4.3.4)

Step 4: If inspected connections are found to be seriously damaged, perform additionalinspections of connections adjacent to the damaged connections. (Section 4.3.5)

Step 5: Determine the average damage index (davg) for connections in each group, and then theaverage damage index at a typical floor. (Section 4.3.6)

Step 6: Given the average damage index for connections in the group, determine theprobability, P, that the connection damage index for any group, at a floor level,exceeds 1/3, and determine the maximum estimated damage index for any floor, Dmax.(Section 4.3.7)

Step 7: Based on the calculated damage indices and statistics, determine appropriateoccupancy, structural repair and modification strategies (Section 4.3.8). If deemedappropriate, the structural engineer may conduct detailed structural analyses of thebuilding in the as-damaged state, to obtain improved understanding of its residualcondition and to confirm that the recommended strategies are appropriate or tosuggest alternative strategies.

Step 8: Report the results of the inspection and evaluation process to the building official andbuilding owner. (Section 4.3.9)

Sections 4.3.2 through 4.3.9 indicate how these steps should be performed.

There are no modifications to the Commentary of Section 4.3.1 at this time.

4.3.2 Step 1— Categorize Connections by Groups


4.3.3 Step 2— Select Samples of Connections for Inspection




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4.3.3.1 Method A - Random Selection


4.3.3.2 Method B - Deterministic Selection


4.3.3.3 Method C - Analytical Selection


4.3.4 Step 3— Inspect the Selected Samples of Connections


Commentary: The sample size suggested for inspection in the methods of Section4.3.3 are based on full inspection using both visual (Section 5.3.1) and NDTtechniques (Section 5.3.2) at all connections in the sample. Other methods ofselection and inspection may be used as provided in Section 4.3, with theapproval of the building official. One such approach might be the visual-onlyinspection of the bottom girder flange to column connection, but with theinspection of a large fraction of the total connections in the group, possiblyincluding all of them. If properly performed, such an inspection procedure woulddetect almost all instances of the most severe damage but would not detect welddefects (W1a), or root cracking (W1b), nor lamellar damage in columns (C5).The occurrence of a few of these conditions, randomly scattered through thebuilding would not greatly affect the assessment of the building’s post-earthquakecondition, or the calculation of the damage index. However, if a large number ofsuch defects were present in the building, this would be significant to the overallassessment. Therefore, such an inspection approach should probably includeconfirming NDT investigations of at least a representative sample of the totalconnections investigated. If within that sample, significant incidence of visuallyhidden damage is found, then full NDT investigations should be performed, assuggested by these Interim Guidelines. Similarly, if visual damage is found at thebottom flange, then complete connection inspection should be performed todetermine if other types of damage are also present.

4.3.4.1 Damage Characterization

Characterize the observed damage at each of the inspected connections by assigning aconnection damage index, dj, obtained either from Table 4-3a or Table 4-3b. Table 4-3a presentsdamage indices for individual classes of damage and a rule for combining indices where aconnection has more than one type of damage. Table 4-3b provides combined indices for themore common combinations of damage.



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Table 4-3a - Connection Damage Indices

Type Location Description1 Index2djG1 Girder Buckled Flange 4G2 Girder Yielded Flange 1G3 Girder Top or Bottom Flange fracture in HAZ 8G4 Girder Top or Bottom Flange fracture outside HAZ 8G5 Girder Top and Bottom Flange fracture 10G6 Girder Yielding or Buckling of Web 4G7 Girder Fracture of Web 10G8 Girder Lateral-torsional Buckling 8C1 Column Incipient flange crack (detectable by UT) 4C2 Column Flange tear-out or divot 8C3 Column Full or partial flange crack outside HAZ 8C4 Column Full or partial flange crack in HAZ 8C5 Column Lamellar flange tearing 6C6 Column Buckled Flange 8C7 Column Fractured column splice 8W1a CJP weld Minor root indication - thickness <3/16” or tf/4; width < bf/4 01W1b CJP weld Root indication - thickness > 3/16” or tf/4 or width > bf/4 04W2 CJP weld Crack through weld metal thickness 8W3 CJP weld Fracture at girder interface 8W4 CJP weld Fracture at column interface 8W5 CJP weld Root indication— non-rejectable 0S1a Shear tab Partial crack at weld to column (beam flanges sound) 4S1b Shear tab Partial crack at weld to column (beam flange cracked) 8S2a Shear tab Crack in Supplemental Weld (beam flanges sound) 1S2b Shear tab Crack in Supplemental Weld (beam flange cracked) 8S3 Shear tab Fracture through tab at bolt holes 10S4 Shear tab Yielding or buckling of tab 6S5 Shear tab Damaged, or missing bolts4 6S6 Shear tab Full length fracture of weld to column 10P1 Panel Zone Fracture, buckle, or yield of continuity plate3 4P2 Panel Zone Fracture of continuity plate welds3 4P3 Panel Zone Yielding or ductile deformation of web3 1P4 Panel Zone Fracture of doubler plate welds3 4P5 Panel Zone Partial depth fracture in doubler plate3 4P6 Panel Zone Partial depth fracture in web3 8P7 Panel Zone Full (or near full) depth fracture in web or doubler plate3 8P8 Panel Zone Web buckling3 6P9 Panel Zone Fully severed column 10Notes To Table 4-3a:

1. See Figures 3-2 through 3-6 for illustrations of these types of damage.2. Where multiple damage types have occurred in a single connection, then:

a. Sum the damage indices for all types of damage with d=1 and treat as one type. If multiple types stillexist; then:



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b. For two types of damage refer to Table 4-3b. If the combination is not present in Table 4-3b and thedamage indices for both types are greater than or equal to 4, use 10 as the damage index for theconnection. If one is less than 4, use the greater value as the damage index for the connection.

c. If three or more types of damage apply and at least one is greater than 4, use an index value of 10,otherwise use the greatest of the applicable individual indices.

3. Panel zone damage should be reflected in the damage index for all moment connections attached to thedamaged panel zone within the assembly.

4. Missing or loose bolts may be a result of construction error rather than damage. The condition of the metalaround the bolt holes, and the presence of fireproofing or other material in the holes can provide clues to this.Where it is determined that construction error is the cause, the condition should be corrected and a damageindex of “0” assigned.

Table 4-3b - Connection Damage Indices for Common Damage Combinations1


Shear TabDamage

DamageIndex


Shear TabDamage

DamageIndex

G3 or G4 S1a 8 C5 S1a 6S1b 10 S1b 10S2a 8 S2a 6S2b 10 S2b 10S3 10 S3 10S4 10 S4 10S5 10 S5 10S6 10 S6 10

C2 S1a 8 W2, W3, or W4 S1a 8S1b 10 S1b 10S2a 8 S2a 8S2b 10 S2b 10S3 10 S3 10S4 10 S4 10S5 10 S5 10S6 10 S6 10

C3 or C4 S1a 8S1b 10S2a 8S2b 10S3 10S4 10S5 10S6 10

1. See Table 4-3a, footnote 2 for combinations other than those contained in this table.

More complete descriptions (including sketches) of the various types of damage are providedin Section 3.1. When the engineer can show by rational analysis that other values for the relativeseverities of damage are appropriate, these may be substituted for the damage indices provided in



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the tables. A full reporting of the basis for these different values should be provided to thebuilding official, upon request.

Commentary: The connection damage indices provided in Table 4-3 (rangingfrom 0 to 10) represent judgmental estimates of the relative severities of thisdamage. An index of 0 indicates no damage and an index of 10 indicates verysevere damage.

When initially developed, these connection damage indices wereconceptualized as estimates of the connection’s lost capacity to reliablyparticipate in the building’s lateral-force-resisting system in future earthquakes(with 0 indicating no loss of capacity and 10 indicating complete loss ofcapacity). However, due to the limited data available, no direct correlationbetween these damage indices and the actual residual strength and stiffness of adamaged connection was ever made. They do provide a convenient measure,however, of the extent of damage that various connections in a building haveexperienced.

When FEMA-267 was first published, weld root discontinuities, Type W1a anddefects, type W1b, were classified as damage in Table 4-3a with damage indicesof 1 and 4, respectively assigned. Recent evidence and investigations, however,suggest strongly that these W1 conditions are not likely to be damage, and alsoare difficult to reliably detect. As a result, with the publication of InterimGuidelines Advisory No. 2, the damage indices for these conditions has beenreduced to a null value, consistent with classifying them as pre-existingconditions, rather than damage.

It should be noted that the reduced damage index associated with theseconditions is not intended to indicate that these are not a concern with regard tofuture performance of the building. In particular, type W1b conditions can serveas ready initiators for the types of brittle fractures associated with the otherdamage types and connections having such conditions are more susceptible tofuture earthquake-induced damage than connections that do not have theseconditions. Correction of these conditions should generally be considered anupgrade or modification, rather than a damage repair.

4.3.5 Step 4— Inspect Connections Adjacent to Damaged Connections


4.3.6 Step 5— Determine Average Damage Index for Each Group




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4.3.7 Step 6— Determine the Probability that the Connections in a Group at a Floor Level Sustained Excessive Damage


4.3.7.1 Some Connections in Group Not Inspected


4.3.7.2 All Connections in Group Inspected


4.3.8 Step 7— Determine Recommended Recovery Strategies for the Building


4.3.9 Step 8 - Evaluation Report


4.4 Alternative Group Selection for Torsional Response


4.5 Qualified Independent Engineering Review


4.5.1 Timing of Independent Review


4.5.2 Qualifications and Terms of Employment


4.5.3 Scope of Review


4.5.4 Reports




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4.5.5 Responses and Corrective Actions


4.5.6 Distribution of Reports


4.5.7 Engineer of Record


4.5.8 Resolution of Differences



Post-Earthquake Inspection

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5. POST-EARTHQUAKE INSPECTION

When required by the building official, or recommended by the Interim Guidelines in Chapter4, post-earthquake inspections of buildings may be conducted in accordance with the InterimGuidelines of this Chapter. In order to determine, with certainty, the actual post-earthquakecondition of a building, it is necessary to inspect all elements and their connections. However, itis permissible to select An an appropriate sample (or samples) of WSMF connections should beselected for inspection in accordance with the Chapter 4 Guidelines. These connections, andothers deemed appropriate by the engineer, should be subjected to visual inspection (VI) andsupplemented by non-destructive testing (NDT) as required by this Chapter.

Commentary: The only way to be certain that all damage sustained by a buildingis detected is to perform complete inspections of every structural element andconnection. In most cases, such exhaustive post-earthquake inspections would beboth economically impractical and also unnecessary. As recommended by theseguidelines, the purpose of post-earthquake inspections is not to detect all damagethat has been sustained by a building, but rather, to detect with reasonablecertainty, that damage likely to result in a significant degradation in thebuilding’s ability to resist future loading. The connection sampling process,suggested by Chapter 4 of these Interim Guidelines was developed to provide alow probability that damage in buildings that had sustained a substantialreduction in load carrying capacity would be overlooked while avoiding theperformance of exhaustive investigations of buildings that have sustainedrelatively insignificant damage.

Where greater certainty in the detection of damage is desired for a building, amore extensive program of inspection can be conducted. For those cases inwhich it is desired to perform an analytical determination of the residual loadcarrying capacity of the structure, complete inspections of elements andconnections should be performed so that an analytical model of the building canbe developed that reasonably represents its post-earthquake condition.

5.1 Connection Types Requiring Inspection

5.1.1 Welded Steel Moment Frame (WSMF) Connections

The inspection of a WSMF connection should start with visual inspection of the weldedbottom beam flange to column flange joint and the base materials immediately adjacent to thisjoint. If damage to this joint is apparent, or suspected, then inspections of that connection shouldbe extended to include the complete joint penetration (CJP) groove welds connecting both topand bottom beam flanges to the column flange, including the backing bar and the weld accessholes in the beam web; the shear tab connection, including the bolts, supplemental welds and



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beam web; the column's web panel zone, including doubler plates; and the continuity plates andcontinuity plate welds (See Figure 3-1). In addition, where visual inspection indicates potentialconcealed damage, visual inspection should be supplemented with other methods ofnondestructive testing.

Commentary: The largest concentration of reported damage following theNorthridge Earthquake occurred at the welded joint between the bottom girderflange and column, or in the immediate vicinity of this joint. To a much lesserextent, damage was also observed in some buildings at the joint between the topgirder flange and column. If damage at either of these locations is substantial (dj

per Chapter 4 greater than 5), then damage is also commonly found in the panelzone or shear tab areas.

When originally published,These these Interim Guidelines recommendedcomplete inspection, by visual and NDT assisted means, of all of these potentialdamage areas for a small representative sample of connections. This practice iswas consistent with that followed by most engineers in the Los Angeles area,following the Northridge Earthquake. It requires removal of fireproofing from arelatively large surface of the steel framing, which at most connections will beundamaged.

In the time since the Interim Guidelines were first published, extensiveinvestigations have been conducted of the statistical distribution of damagesustained by buildings in the Northridge earthquake, the nature of this damageand the effect of this damage on the future load-carrying capacity of thebuildings. These investigations strongly suggest that the W1a and W1bconditions at the weld root are unlikely to be earthquake damage, but rather,conditions of discontinuity and defects from the original construction. Further,studies have shown that NDT methods are generally unreliable in the detection ofthese conditions. As a result, the current recommendation is not to conductexhaustive NDT investigations of connections in order to discover hiddendamage, as was originally recommended.

In a series of analytical investigations of the effect of moment-resistingconnection damage on building behavior, it was determined that even if a largenumber of connections experience fracture at one beam flange to column joint,there is relatively little increase in the probability of global collapse in a futureearthquake. Similarly, these investigations indicate that if both the top andbottom beam flange to column joints fracture in a large a number of connections,a very significant increase in the probability of global building collapse occurs. Therefore, to reduce the costs associated with post-earthquake inspections, withthe publication of Interim Guidelines Advisory No.2 it is recommended that post-earthquake inspections initially be limited to visual inspection of the beam bottom



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flange to column joint region. If there is evidence of potential damage in thisregion that is not directly observable by visual means, for example, a gap betweenthe weld backing and column flange, then supplemental investigations of this jointshould be conducted using NDT. Similarly, if it is determined that fractures haveoccurred at the beam bottom flange joint, then inspections of that connectionshould be extended to encompass the entire connection including the top beamflange joint, the shear tab and column panel zone. This approach was permittedas an alternate, in the original publication of the Interim Guidelines.

Some engineers have suggested an alternative approach consisting of visual -only inspections, limited to the girder bottom flange to column joint, but for avery large percentage of the total connections in the building. These bottomflange joint connections can be visually inspected with much less fireproofingremoved from the framing surfaces. When significant damage is found at theexposed bottom connection, then additional fireproofing is removed to allow fullexposure of the connection and inspection of the remaining surfaces. Theseengineers feel that by inspecting more connections, albeit to a lesser scope thanrecommended in these Interim Guidelines, their ability to locate the most severeoccurrences of damage in a building is enhanced. These engineers use NDTassisted inspection on a very small sample of the total connections exposed toobtain an indication of the likelihood of hidden problems including damage types.

If properly executed, such an approach can provide sufficient information toevaluate the post-earthquake condition of a building and to make appropriateoccupancy, structural repair and/or modification decisions. It is important thatthe visual inspector be highly trained and that visual inspections be carefullyperformed, preferably by a structural engineer. Casual observation may missclues that hidden damage exists. If, as a result of the partial visual inspection,there is any reason to believe that damage exists at a connection (such as smallgaps between the CJP weld backing and column face), then complete inspectionof the suspected connection, in accordance with the recommendations of theseInterim Guidelines should be performed. If this approach is followed, it isrecommended that a significantly larger sample of connections than otherwiserecommended by these Interim Guidelines, perhaps nearly all of the connections,be inspected.

5.1.2 Gravity Connections


5.1.3 Other Connection Types




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5.2 Preparation

5.2.1 Preliminary Document Review and Evaluation

5.2.1.1 Document Collection and Review


5.2.1.2 Preliminary Building Walk-Through.


5.2.1.3 Structural Analysis


5.2.1.4 Vertical Plumbness Check


5.2.2 Connection Exposure

Pre-inspection activities to expose and prepare a connection for inspection should include thelocal removal of suspended ceiling panels or (as applicable) local demolition of permanent ceilingfinish to access the connection; and cleaning of sufficient fireproofing from the beam and columnsurfaces to allow visual observation of the area to be inspected. If initial inspections are to belimited to the beam bottom flange to column joint and the surrounding material, fireproofingshould be removed from the connection as indicated in Figure 5-1a. Removal of fireproofing needonly be sufficient to permit observation of the surfaces of base and weld metals. Wire brushingand cleaning to remove all particles of fireproofing material is not necessary unless ultrasonictesting of the joint area is to be conducted. In the event that damage is found at the bottom beamflange to column joint, then additional fireproofing should be removed, as indicated in Figure 5-1b, to expose the column panel zone, the column flange, continuity plates, beam web and flanges. The extent of the removal of fireproofing should be sufficient to allow adequate inspection of thesurfaces to be inspected. Figure 5-1b suggests a pattern that will allow both visual and NDTinspection of the top and bottom beam flange to column joints, the beam web and shearconnection, column panel zone and continuity plates, and column flanges in the areas of highestexpected demands. The maximum extent of the removal of fireproofing need not be greater thana distance equal to the beam depth "d" into the beam span to expose evidence of any yielding.



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6”

6”6”

Fireproofing

Exposed surfaces

Figure 5-1a Recommended Zone for Fireproofing Removal for Initial Inspections

6”

6”12”

Fireproofing

Figure 5-1b Recommended Zone for Removal of Fireproofing for Complete Inspections

Commentary: If inspection is to be limited to visual observation of the surfaces ofthe base metal and welds, cleaning of fireproofing need only be sufficient toexpose these surfaces. However, if ultrasonic testing is to be performed, thesurface over which the scanning will be performed must be free Cleaning of weldareas and removal of mill scale and weld spatter. Such cleaning should be donewith care, preferably using a power wire brush, to ensure a clean surface thatdoes not affect the accuracy of ultrasonic testing. The resulting surface finishshould be clean, free of mill scale, rust and foreign matter. The use of a chiselshould be avoided to preclude scratching the steel surfaces which could bemistaken for yield lines. Sprayed-on fireproofing on WSMFs erected prior toabout 19801970 is likely to contain asbestos and should be handled according to



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applicable standards for the removal of hazardous materials. Health hazardsassociated with asbestos were recognized by industry in the late 1960s and by1969, most commercial production of asbestos containing materials had ceased. In April, 1973, the federal government formally prohibited the production ofasbestos containing materials with the adoption of the National EmissionStandards for Hazardous Air Pollutants. Allowing for shelf life of materialsproduced prior to that date, it should be considered possible that buildingsconstructed prior to 1975 contain some asbestos hazards. To preclude physicalexposure to hazardous materials and working conditions in such buildings, thestructural engineer should require by contractual agreement with the buildingowner, prior to the start of the inspection program, that the building ownerdeliver to the structural engineer for his/her review and files a laboratorycertificate that confirms the absence of asbestos in structural steel fireproofing,local pipe insulation, ceiling tiles, and drywall joint compound.

The pattern of fireproofing removal indicated in Figure 5-1 is adequate toallow visual and UT inspection of the top and bottom girder flange to columnjoints, the beam web and shear connection and the column panel zone. Asdiscussed in the commentary to Section 5.1.1, some engineers prefer to initiallyinspect only the bottom beam flange to column joint. In such cases, the initialremoval of fireproofing can be more limited than indicated in the figure. If afterinitial inspection, damage at a connection is suspected, then full removal, asindicated in the figure, should be performed to allow inspection of all areas of theconnection.

5.3 Inspection Program

5.3.1 Visual Inspection (VI)


5.3.1.1 Top Flange


5.3.1.2 Bottom Flange


5.3.1.3 Column and Continuity Plates




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5.3.1.4 Beam Web Shear Connection


5.3.2 Nondestructive Testing (NDT)

NDT should may be used to supplement the visual inspection of connections selected inaccordance with the Interim Guidelines of Chapter 4. The testing agency and NDT personnelperforming this work should conform to the qualifications indicated in Chapter 11 of these InterimGuidelines. The following NDT techniques should may be used at the top and bottom of eachconnection, where accessible, to supplement visual inspection: These techniques should be usedwhenever visual inspection indicates the potential for damage that is not directly observable.

a) Magnetic particle testing (MT) of the beam flange to column flange weld surfaces may beused to confirm the presence of suspected surface cracks based on visual evidence. Wherefractures are evident from visual inspection, MT should be used to confirm the lateralextent of the fracture.All surfaces which were visually inspected should be tested using themagnetic particle technique.

Commentary: The color of powder should be selected to achieve maximumcontrast to the base and weld metal under examination. The test may be furtherenhanced by applying a white coating made specifically for MT or by applyingpenetrant developer prior to the MT examination. This background coatingshould be allowed to thoroughly dry before performing the MT.

b) Ultrasonic testing (UT) may be used to detect the presence of hidden fractures, wherevisual inspection reveals the potential for such fractures. of all faces at the beam flangewelds and adjacent column flanges (extending at least 3 inches above and below thelocation of the CJP weld, along the face of the column, but not less than 1-1/2 times thecolumn flange thickness).

Commentary: The purpose of UT is to 1) locate and describe the extent ofinternal defects not visible on the surface and 2) to determine the extent of cracksobserved visually and by MT. These guidelines recommend the use of visualinspection as the primary tool for detecting earthquake damage (See commentaryto Sec. 5..1.1). UT can be a useful technique for confirmation of the presence ofsuspected fractures at the beam flange to column flange joints. Visual evidencethat may suggest the need for such testing could include apparent separation ofthe base of the weld backing from the face of the column.

Requirements and acceptance criteria for NDT should be as given in AWS D1.1-98 Sections 6and 8. Acceptance or rejection of planar weld discontinuity (cracks, slag inclusion, or lack offusion), including root indications, should, as a minimum, be consistent with AWS DiscontinuitiesSeverity Class designations of cracks and defects per Table 8.26.2 of AWS D1.1-98 for Static



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Structures. Beam flange welds should be tested as "tension welds" per AWS D1.1 Table 8.15.3,Note 3. Backing bars need not be removed prior to performing UT.

Commentary: The value of UT for locating small discontinuities at the root ofbeam flange to column flange welds when the backing is left in place is notuniversally accepted. The reliability of this technique is particularly questionableat the center of the joint, where the beam web obscures the signal. There havebeen a number of reported instances of UT detected indications which were notfound upon removal of the backing, and similarly, there have been reportedinstances of defects which were missed by UT examination but were evident uponremoval of the backing. The smaller the defect, the less likely it is that UT alonewill reliably detect its presence.

Despite the potential inaccuracies of this technique, it is the only methodcurrently available, short of removal of the backing, to find subsurface damage inthe welds. It is also the most reliable method for finding lamellar problems in thecolumn flange (type C4 and C5 damage) opposite the girder flange. Removal ofweld backing at these connections results in a significant cost increase that isprobably not warranted unless UT indicates widespread, significant defectsand/or damage in the building.

The proper scanning techniques, beam angle(s) and transducer sizes should be used asspecified in the written UT procedure contained in the Written Practice, prepared in accordancewith Section 5.3.3 of these Interim Guidelines. The acceptance standard should be that specifiedin the original contract documents, but in no case should it be less than the acceptance criteria ofAWS D1.1, Chapter 8, for Statically Loaded Structures.

The base metal should be scanned with UT for cracks. Cracks which have propagated to thesurface of the weld or beam and column base metal will probably have been detected by visualinspection and magnetic particle tests performed earlier. The purpose of ultrasonic testing of thebase metal is to:

1. Locate and describe the extent of internal indications not apparent on the surface and,

2. Determine the extent of cracks found visually and by magnetic particle test.

Commentary: Liquid dye penetrant testing (PT) may be used where MT isprecluded due to geometrical conditions or restricted access. Note that morestringent requirements for surface preparation are required for PT than MT, perAWS D1.1.

If practical, NDT should be performed across the full width of the bottombeam flange joint. However, if there are no discontinuity signals from UT of



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accessible faces on one side of the bottom flange weld, obstructions on the otherside of the connection need not be removed for testing of the bottom flange weld.

Slabs, flooring and roofing need not be removed to permit NDT of the topflange joint unless there is significant visible damage at the bottom beam flange,adjacent column flange, column web, or shear connection. Unless such damageis present, NDT of the top flange should be performed as permitted, without localremoval of the diaphragms or perimeter wall obstructions.

It should be noted that UT is not 100% effective in locating discontinuitiesand defects in CJP beam flange to column flange welds. The ability of UT toreliably detect such defects is very dependent on the skill of the operator and thecare taken in the inspection. Even under perfect conditions, it is difficult toobtain reliable readings of conditions at the center of the beam flange to columnflange connection as return signals are obscured by the presence of the beamweb. If backing is left in place on the welds, UT becomes even less reliable. There have been a number of reported instances in which UT indicated apparentdefects, that were found not to exist upon removal of the backing. Similarly, UThas failed in some cases to locate defects that were later discovered upon removalof the backing. Additional information on UT may be found in AWS B1.10.


5.3.4 Post-Earthquake Field Inspection Report


5.3.5 Written Report




5-10



Post-earthquake Repair and Modification

6-1

6. POST-EARTHQUAKE REPAIR AND MODIFICATION

6.1 Scope


6.2 Shoring


6.3 Repair Details


6.4 Preparation


6.5 Execution


6.6 STRUCTURAL MODIFICATION

6.6.1 Definition of Modification


6.6.2 Damaged vs. Undamaged Connections


6.6.3 Criteria

Connection modification intended to permit inelastic frame behavior should be proportionedso that the required plastic deformation of the frame may be accommodated through thedevelopment of plastic hinges at pre-determined locations within the girder spans, as indicated inFigure 6-12 Figure 6.6.3-1. Beam-column connections should be designed with sufficientstrength (through the use of cover plates, haunches, side plates, etc.) to force development of theplastic hinge away from the column face. This condition may also be attained through localweakening of the beam section, at the desired location for plastic hinge formation. All elementsof the connection should have adequate strength to develop the forces resulting from the



6-2

formation of the plastic hinge at the predetermined location, together with forces resulting fromgravity loads.

Plastic Hinges


L’

L

h

drift angle - θ

Figure 6-12 Figure 6.6.3-1 - Desired Plastic Frame Behavior

Commentary: Nonlinear deformation of frame structures is typicallyaccommodated through the development of inelastic flexural or shear strainswithin discrete regions of the structure. At large inelastic strains these regionscan develop into plastic hinges, which can accommodate significant concentratedrotations at constant (or nearly constant) load through yielding at tensile fibersand buckling at compressive fibers. If a sufficient number of plastic hingesdevelop in a frame, a mechanism is formed and the frame can deform laterally ina plastic manner. This behavior is accompanied by significant energydissipation, particularly if a number of members are involved in the plasticbehavior, as well as substantial local damage to the highly strained elements.The formation of hinges in columns, as opposed to beams, is undesirable, as thisresults in the formation of weak story mechanisms with relatively few elementsparticipating, and consequently little energy dissipation occurring. In addition,such mechanisms also result in local damage to critical gravity load bearingelements.

The prescriptive connection contained in the UBC and NEHRP RecommendedProvisions prior to the Northridge Earthquake was based on the assumeddevelopment of plastic hinge zones within the beams at adjacent to the face of thecolumn, or within the column panel zone itself. If the plastic hinge develops inthe column panel zone, the resulting column deformation results in very largesecondary stresses on the beam flange to column flange joint, a condition whichcan contribute to brittle failure. If the plastic hinge forms in the beam, at the faceof the column, this can result in very large through-thickness strain demands on



6-3

the column flange material and large inelastic strain demands on the weld metaland surrounding heat affected zones stress and strain demands on the weldedbeam flange to column flange joint. These conditions can also lead to brittlejoint failure. Although ongoing research may reveal conditions of materialproperties, design and detailing configurations that permit connections withyielding occurring at the column face to perform reliably, for the present, it isrecommended In order to achieve more reliable performance, it is recommendedthat the connection of the beam to the column be modified to be sufficientlystrong to force the inelastic action (plastic hinge) away from the column face.Plastic hinges in steel beams have finite length, typically on the order of half thebeam depth. Therefore, the location for the plastic hinge should be shifted atleast that distance away from the face of the column. When this is done, theflexural demands on the columns are increased. Care must be taken to assurethat weak column conditions are not inadvertently created by local strengtheningof the connections.

It should be noted that connection modifications of the type described above,while believed to be effective in preventing brittle connection fractures, will notprevent structural damage from occurring. Brittle connection fractures areundesirable because they result in a substantial reduction in the lateral-force-resisting strength of the structure which, in extreme cases, can result in instabilityand collapse. Connections modified as described in these Interim Guidelinesshould experience many fewer such brittle fractures than unmodified connections.However, the formation of a plastic hinge within the span of a beam is not acompletely benign event. Beams which have formed such hinges may exhibitlarge buckling and yielding deformation, damage which typically must berepaired. The cost of such repairs could be comparable to the costs incurred inrepairing fracture damage experienced in the Northridge Earthquake. Theprimary difference is that life safety protection will be significantly enhanced andmost structures that have experienced such plastic deformation damage shouldcontinue to be safe for occupancy while repairs are made.

If the types of damage described above are unacceptable for a given building,then alternative methods of structural modification should be considered that willreduce the plastic deformation demands on the structure during a strongearthquake. Appropriate methods of achieving such goals include the installationof supplemental braced frames, energy dissipation systems, and similarsystematic modifications of the building’s basic lateral force resisting system.

It is important to recognize that in frames with relatively short bays, theflexural hinging indicated in Figure 6.6.3-1 may not be able to form. If theeffective flexural length (L’ in the figure) of beams in a frame becomes too short,then the beams or girders will yield in shear before zones of flexural plasticity



6-4

can form, resulting in an inelastic behavior that is more like that of aneccentrically braced frame than that of a moment frame. This behavior mayinadvertently occur in frames in which relatively large strengthened connections,such as haunches, cover plates or side plates have been used on beams withrelatively short spans. This behavior is illustrated in Figure 6.6.3-2.

The guidelines contained in this section are intended to address the design offlexurally dominated moment resisting frames. When utilizing these guidelines, itis important to confirm that the configuration of the structure is such that thepresumed flexural hinging can actually occur. It is possible that shear yielding offrame beams, such as that schematically illustrated in Figure 6.6.3-2 may be adesirable behavior mode. However, to date, there has not been enough researchconducted into the behavior of such frames to develop recommended designguidelines. If modifications to an existing frame result in such a configurationdesigners should consider referring to the code requirements for eccentricallybraced frames. Particular care should be taken to brace the shear link of suchbeams against lateral-torsional buckling and also to adequately stiffen the websto avoid local buckling following shear plastification.

Shear Link

Shear Link

Figure 6.6.3-2 Shear Yielding Dominated Behavior of Short Bay Frames

6.6.4 Strength and Stiffness

6.6.4.1 Strength

When these Interim Guidelines require determination of the strength of a framing element orcomponent, this shall be calculated in accordance with the criteria contained in UBC-94, Section



6-5

2211.4.2 {NEHRP-91 Section 10.2, except that the factor φ should be taken as 1.0}, restated asfollows:


Flexure Ms = Z Fy


Alternatively, the criteria contained in the 1997 edition of the AISC Seismic Provisions forStructural Steel Buildings (AISC, 1997) may be used.

Commentary: At the time the Interim Guidelines were first published, they werebased on the 1994 edition of the Uniform Building Code and the 1994 edition ofthe NEHRP Provisions. In the time since that initial publication, more recenteditions of both documents have been published, and codes based on thesedocuments have been adopted by some jurisdictions. In addition, the AmericanInstitute of Steel Construction has adopted a major revision to its SeismicProvisions for Structural Steel Buildings (AISC Seismic Provisions), largelyincorporating, with some modification, the recommendations contained in theInterim Guidelines. This updated edition of the AISC Seismic Provisions hasbeen incorporated by reference into the 1997 edition of the NEHRP Provisionsand has also been adopted by some jurisdictions as an amendment to the modelbuilding codes. Structural upgrades designed to comply with the requirements ofthe 1997 AISC Seismic Provisions may be deemed to comply with the intent ofthese Interim Guidelines. Where reference is made herein to the requirements ofthe 1994 Uniform Building Code or 1994 NERHP Provisions, the parallelprovisions of the 1997 editions may be used instead, and should be used in thosejurisdictions that have adopted codes based on these updated standards.

Partial penetration welds are not recommended for tension applications incritical connections resisting seismic induced stresses. The geometry of partialpenetration welds creates a notch-like condition that can initiate brittle fractureunder conditions of high tensile strain.

Many WSMF structures are constructed with concrete floor slabs that areprovided with positive shear attachment between the slab and the top flanges ofthe girders of the moment-resisting frames. Although not generally accounted forin the design of the frames, the resulting composite action can increase the



6-6

effective strength of the girder significantly, particularly at sections wherecurvature of the girder places the top flange into compression. Although thiseffect is directly accounted for in the design of composite systems, it is typicallyneglected in the design of systems classified as moment resisting steel frames.The increased girder flexural strength caused by this composite action can resultin a number of effects including the unintentional creation of weak column -strong beam and weak panel zone conditions. In addition, this composite effecthas the potential to reduce the effectiveness of reduced section or “dog-bone”type connection assemblies. Unfortunately, very little laboratory testing of largescale connection assemblies with slabs in place has been performed to date andas a result, these effects are not well quantified. In keeping with typicalcontemporary design practice, the design formulae provided in these Guidelinesneglect the strengthening effects of composite action. Designers should, however,be alert to the fact that these composite effects do exist. Similar, and perhapsmore severe, effects may also exist where steel beams support masonry orconcrete walls.

6.6.4.2 Stiffness

Calculation of frame stiffness for the purpose of determining interstory drift under theinfluence of the design lateral forces should be based on the properties of the bare steel frame,neglecting the effects of composite action with floor slabs. The stiffening effects of connectionreinforcements (e.g.: haunches, side plates, etc.) may be considered in the calculation of overallframe stiffness and drift demands. When reduced beam section connections are utilized, thereduction in overall frame stiffness, due to local reductions in girder cross section, should beconsidered.

Commentary: For design purposes, frame stiffness is typically calculatedconsidering only the behavior of the bare frame, neglecting the stiffening effectsof slabs, gravity framing, and architectural elements such as walls. The resultingcalculation of building stiffness and period typically underestimates the actualproperties, substantially. Although this approach can result in unconservativeestimates of design force levels, it typically produces conservative estimates ofinterstory drift demands. Since the design of most moment-resisting frames arecontrolled by considerations of drift, this approach is considered preferable tomethods that would have the potential to over-estimate building stiffness. Also,many of the elements that provide additional stiffness may be subject to rapiddegradation under severe cyclic lateral deformation, so that the bare framestiffness may provide a reasonable estimate of the effective stiffness under longduration ground shaking response.

Notwithstanding the above, designers should be alert to the fact thatunintentional stiffness introduced by walls and other non-structural elements can



6-7

significantly alter the behavior of the structure in response to ground shaking. Ofparticular concern, if these elements are not uniformly distributed throughout thestructure, or isolated from its response, they can cause soft stories and torsionalirregularities, conditions known to result in poor behavior.


The plastic rotation capacity of modified connections should reflect realistic estimates of therequired level of plastic rotation demand. In the absence of detailed calculations of rotationdemand, connections should be shown to be capable of developing a minimum plastic rotationcapacity on the order of 0.025 to 0.030 radian. The demand may be lower when braced frames,supplemental damping, base isolation, or other elements are introduced into the moment framesystem, to control its lateral deformation; when the design ground motion is relatively low in therange of predominant periods for the structure; and when the frame is sufficiently strong and stiff.

As used in these Guidelines, plastic rotation is defined as the plastic chord rotation angle. Theplastic chord rotation angle is calculated using the rotated coordinate system shown in Fig. 6.6.5-1 as the plastic deflection of the beam or girder, at the point of inflection (usually at the center ofits span,) ∆CI, divided by the distance between the center of the beam span and the centerline ofthe panel zone of the beam column connection, LCL. This convention is illustrated in Figure 6.6.5-1.

It is important to note that this definition of plastic rotation is somewhat different than theplastic rotation that would actually occur within a discrete plastic hinge in a frame model similarto that shown in Figure 6.6.3-1. These two quantities are related to each other, however, and ifone of them is known, the other may be calculated from Eq. 6.6.5-1.

cL

cL

Beam span center line

∆CL

LCL

θ pCL

CLL= ∆

Plastic hinge

lh

Figure 6.6.5-1 Calculation of Plastic Rotation Angle



6-8

( )θ θp ph

CL h

CL

L lL

= − (6.6.5-1)

where: θp is the plastic chord angle rotation, as used in these Guidelines θph is the plastic rotation, at the location of a discrete hinge LCL is the distance from the center of the beam span to the center of the beam-column assembly panel zone lh is the assumed location of the discrete plastic hinge relative to the center of the beam-column assembly panel zone

If calculations are performed to determine the required connection plastic rotation capacity,the capacity should be taken somewhat greater than the calculated deformation demand, due tothe high variability and uncertainty inherent in predictions of inelastic seismic response. Untilbetter guidelines become available, a required plastic rotation capacity on the order of 0.005radians greater than the demand calculated for the design basis earthquake (or if greaterconservatism is desired - the maximum capable considered earthquake) is recommended.Rotation demand calculations should consider the effect of plastic hinge location within the beamspan, as indicated in Figure 6-12 Figure 6.6.3-1, on plastic rotation demand. Calculations shouldbe performed to the same level of detail specified for nonlinear dynamic analysis for base isolatedstructures in UBC-94 Section 1655 {NEHRP-94 Section 2.6.4.4}. Ground motion time historiesutilized for these nonlinear analyses should satisfy the scaling requirements of UBC-94 Section1655.4.2 {NEHRP-94 Section 2.6.4.4} except that instead of the base isolated period, TI, thestructure period, T, calculated in accordance with UBC-94 Section 1628 {NEHRP-94 Section2.3.3.1} should be used.

Commentary. When the Interim Guidelines were first published, the plasticrotation was defined as that rotation that would occur at a discrete plastic hinge,similar to the definition of θph. in Eq. 6.6.5-1, above. In subsequent testing ofprototype connection assemblies, it was found that it is often very difficult todetermine the value of this rotation parameter from test data, since actual plastichinges do not occur at discrete points in the assembly and because some amountof plasticity also occurs in the panel zone of many assemblies. The plastic chordangle rotation, introduced in Interim Guidelines Advisory No. 1, may morereadily be obtained from test data and also more closely relates to the driftexperienced by a frame during earthquake response.

Traditionally, structural engineers have calculated demand in momentframes by sizing the members for strength and drift using code forces (eitherequivalent static or reduced dynamic forces) and then "developing the strength ofthe members." Since 1988, "developing the strength" has been accomplished byprescriptive means. It was assumed that the prescribed connections would bestrong enough so that the girder would yield (in bending), or the panel zone



6-9

would yield (in shear) in a nearly perfectly plastic manner producing the plasticrotations necessary to dissipate the energy of the earthquake. It is now knownthat the prescriptive connection is often incapable of behaving in this manner.

In the 1994 Northridge earthquake, many moment-frame connectionsfractured with little evidence of plastic hinging of the girders or yielding of thecolumn panel zones. Testing of moment frame connections both prior to andsubsequent to the earthquake suggests that the standard welded flange-bolted webconnection is unable to reliably provide plastic rotations beyond about 0.005radian for all ranges of girder depths and often fails below that level. Thus, forframes designed for code forces and for the code drift limits, new connectionconfigurations must be developed to reliably accommodate such rotation withoutbrittle fracture.

In order to develop reasonable estimates of the plastic rotation demands on aframe’s connections, it is necessary to perform inelastic time history analyses.For regular structures, approximations of the plastic rotation demands can beobtained from linear elastic analyses. Analytical research (Newmark and Hall -1982) suggests that for structures having the dynamic characteristics of mostWSMF buildings, and for the ground motions typical of western US earthquakes,the total frame deflections obtained from an unreduced (no R or Rw factor)dynamic analysis provide an approximate estimate of those which would beexperienced by the inelastic structure. For the typical spectra contained in thebuilding code, this would indicate expected drift ratios on the order of 1%. Thedrift demands in a real structure, responding inelastically, tend to concentrate ina few stories, rather than being uniformly distributed throughout the structure’sheight. Therefore, it is reasonable to expect typical drift demands in individualstories on the order of 1.5% to 2% of the story height. As a roughapproximation, the drift demand may be equated to the joint rotation demand,yielding expected rotation demands on the order of perhaps 2%. Since there isconsiderable variation in ground motion intensity and spectra, as well as theinelastic response of buildings to these ground motions, conservatism in selectionof an appropriate connection rotation demand is warranted.

In recent testing of large scale subassemblies incorporating modifiedconnection details, conducted by SAC and others, when the connection design wasable to achieve a plastic rotation demand of 0.025 radians or more for severalcycles, the ultimate failure of the subassembly generally did not occur in theconnection, but rather in the members themselves. Therefore, the statedconnection capacity criteria would appear to result in connections capable ofproviding reliable performance.

It should be noted that the connection assembly capacity criteria for themodification of existing buildings, recommended by these Interim Guidelines, is



6-10

somewhat reduced compared to that recommended for new buildings (Chapter 7).This is typical of approaches normally taken for existing structures. For newbuildings, these Interim Guidelines discourage building-specific calculation ofrequired plastic rotation capacity for connections and instead, encourage thedevelopment of highly ductile connection designs. For existing buildings, such anapproach may lead to modification designs that are excessively costly, as well asthe modification of structures which do not require such modification.Consequently, an approach which permits the development of semi-ductileconnection designs, with sufficient plastic rotation capacity to withstand theexpected demands from a design earthquake is adopted. It should be understoodthat buildings modified to this reduced criteria will not have the same reliabilityas new buildings, designed in accordance with the recommendations of Chapter7. The criteria of Chapter 7 could be applied to existing buildings, if superiorreliability is desired.

When performing inelastic frame analysis, in order to determine the requiredconnection plastic rotation capacity, it is important to accurately account for thelocations at which the plastic hinges will occur. Simplified models, whichrepresent the hinge as occurring at the face of the column, maywill underestimatethe plastic rotation demand. This problem becomes more severe as the columnspacing, L, becomes shorter and the distance between plastic hinges, L’, agreater portion of the total beam span. Eq. 6.6.5-1 may be used to convertcalculated values of plastic rotation at a hinge remotely located from the column,to the chord angle rotation, used for the definition of acceptance criteriacontained in these Guidelines. In extreme cases, the girder will not form plastichinges at all, but instead, will develop a shear yield, similar to an eccentricbraced frame.

6.6.6 Connection Qualification and Design

Modified girder-column connections may be qualified by testing or designed usingcalculations. Qualification by testing is the preferred approach. Preliminary designs ofconnections to be qualified by test may be obtained using the calculation procedures of Section6.6.6.3. The procedures of that section may also be used to calibrate previous tests of similarconnection configurations to slightly different applications, by extrapolation. Extrapolation of testresults should be limited to connections of elements having similar geometries and materialspecifications as the tested connections. Designs based on calculation alone should be subject toqualified independent third party review.

Commentary: Because of the cost of testing, use of calculations for interpolationor extrapolation of test results is desirable. How much extrapolation should beaccepted is a difficult decision. As additional testing is done, more informationmay be available on what constitutes "conservative" testing conditions, thereby



6-11

allowing easier decisions relative to extrapolating tests to actual conditions whichare likely to be less demanding than the tests. For example, it is hypothesizedthat connections of shallower, thinner flanged members are likely to be morereliable than similar connections consisting of deeper, thicker flanged members.Thus, it may be possible to test the largest assemblages of similar details andextrapolate to the smaller member sizes? - at least within comparable membergroup families. However, there is evidence to suggest that extrapolation of testresults to assemblies using members of reduced size is not always conservative.In a recent series of tests of cover plated connections, conducted at the Universityof California at San Diego, a connection assembly that produced acceptableresults for one family of beam sizes, W24, did not behave acceptably when thebeam depth was reduced significantly to W18. In that project, the change inrelative flexibilities of the members and connection elements resulted in a shift inthe basic behavior of the assembly and initiation of a failure mode that was notobserved in the specimens with larger member sizes. In order to minimize thepossibility of such occurrences, when extrapolation of test results is performed, itshould be done with a basic understanding of the behavior of the assembly, andthe likely effects of changes to the assembly configuration on this behavior. Testresults should not be extrapolated to assembly configurations that are expected tobehave differently than the tested configuration. Extrapolation or interpolationof results with differences in welding procedures, details or material properties iseven more difficult.

6.6.6.1 Qualification Test Protocol


6.6.6.2 Acceptance Criteria


a) The connection should develop beam plastic rotations as indicated in Section 6.6.5, forat least one complete cycle.

b) The connection should develop a minimum strength equal to 80% of the plasticstrength of the girder, calculated using minimum specified yield strength Fy,throughout the loading history required to achieve the required plastic rotationcapacity, as indicated in a), above.

c) The connection should exhibit ductile behavior throughout the loading history. Aspecimen that exhibits a brittle limit state (e.g. complete flange fracture, columncracking, through-thickness failures of the column flange, fractures in welds subject to



6-12

tension, shear tab cracking, etc. ) prior to reaching the required plastic rotation shall beconsidered unsuccessful.

d) Throughout the loading history, until the required plastic rotation is achieved, theconnection should be judged capable of supporting dead and live loads required by thebuilding code. In those specimens where axial load is applied during the testing, thespecimen should be capable of supporting the applied load throughout the loadinghistory.

The evaluation of the test specimen’s performance should consistently reflect the relevant limitstates. For example, the maximum reported moment and the moment at the maximum plasticrotation are unlikely to be the same. It would be inappropriate to evaluate the connection usingthe maximum moment and the maximum plastic rotation in a way that implies that they occurredsimultaneously. In a similar fashion, the maximum demand on the connection should beevaluated using the maximum moment, not the moment at the maximum plastic rotation unless thebehavior of the connection indicated that this limit state produced a more critical condition in theconnection.

Commentary: Many connection configurations will be able to withstandplastic rotations on the order of 0.025 radians or more, but will have sustainedsignificant damage and degradation of stiffness and strength in achieving thisdeformation. The intent of the acceptance criteria presented in this Section is toassure that when connections experience the required plastic rotation demand,they will still have significant remaining ability to participate in the structure’slateral load resisting system.

In evaluating the performance of specimens during testing, it is important todistinguish between brittle behavior and ductile behavior. It is not uncommon forsmall cracks to develop in specimens after relatively few cycles of inelasticdeformation. In some cases these initial cracks will rapidly lead to ultimatefailure of the specimen and in other cases they will remain stable, perhapsgrowing slowly with repeated cycles, and may or may not participate in theultimate failure mode. The development of minor cracks in a specimen, prior toachievement of the target plastic rotation demand should not be cause forrejection of the design if the cracks remain stable during repeated cycling.Similarly, the occurrence of brittle fracture at inelastic rotations significantly inexcess of the target plastic rotation should not be cause for rejection of thedesign.

6.6.6.3 Calculations




6-13

6.6.6.3.1 Material Strength Properties

In the absence of project specific material property information (for example, mill testreports), the values listed in Table 6-3 Table 6.6.6.3.1-1 should be used to determine the strengthof steel shape and plate for purposes of calculation. The permissible strength for weld metalshould be taken in accordance with the building code.

Table 6-3Table 6.6.6.3-1 - Properties for Use in Connection Modification Design

Material Fy (ksi) Fy m (ksi) Fu (ksi)A36 Beam 36 1 1

Dual Certified Beam Axial, Flexural Shape Group 1 Shape Group 2 Shape Group 3 Shape Group 4 Through-Thickness

50

-

552

582

572

542

-

65 min.

Note 3A572 Column/Beam Axial, Flexural Shape Group 1 Shape Group 2 Shape Group 3 Shape Group 4 Shape Group 5 Through-Thickness

50

-

582

582

572

572

552

-

65 min.

Note 3A992 Structural Shape1 Use same values as for A572, Gr. 50Notes:1. See Commentary2. Based on coupons from web. For thick flanges,

the Fy flange is approximately 0.95 Fy web.3. See Commentary

Commentary: Table 6-3, Note 1 - The material properties for steel nominallydesignated on the construction documents as ASTM A36 can be highly variableand in recent years, steel meeting the specified requirements for both ASTM A36and A572 has routinely been incorporated in projects calling for A36 steel.Consequently, unless project specific data is available to indicate the actualstrength of material incorporated into the project, the properties for ASTM A572steel should be assumed when ASTM A36 is indicated on the drawings, and theassumption of a higher yield stress results in a more severe design condition.

The ASTM A992 specification was specifically developed by the steel industryin response to expressed concerns of the design community with regard to thepermissible variation in chemistry and mechanical properties of structural steelunder the A36 and A572 specifications. This new specification, which wasadopted in late 1998, is very similar to ASTM A572, except that it includessomewhat more restrictive limits on chemistry and on the permissible variation in



6-14

yield and ultimate tensile stress, as well as the ratio of yield to tensile strength.At this time, no statistical data base is available to estimate the actualdistribution of properties of material produced to this specification. However, theproperties are likely to be very similar, albeit with less statistical scatter, to thoseof material recently produced under ASTM A572, Grade 50.

Table 6-3Table 6.6.6.3-1, Note 3 - In the period immediately following theNorthridge earthquake, the Seismology Committee of the Structural EngineersAssociation of California and the International Conference of Building Officialsissued Interim Recommendation No. 2 (SEAOC-1995) to provide guidance on thedesign of moment resisting steel frame connections. Interim RecommendationNo. 2 included a recommendation that the through-thickness stress demand oncolumn flanges be limited to a value of 40 ksi, applied to the projected area ofbeam flange attachment. This value was selected somewhat arbitrarily, to ensurethat through-thickness yielding did not initiate in the column flanges of moment-resisting connections and because it was consistent with the successful tests ofassemblies with cover plates conducted at the University of Texas at Austin(Engelhardt and Sabol - 1994), rather than being the result of a demonstratedthrough-thickness capacity of typical column flange material. Despite thesomewhat arbitrary nature of the selection of this value, its use often controls theoverall design of a connection assembly including the selection of cover platethickness, haunch depth, and similar parameters.

It would seem to be important to prevent the inelastic behavior of connectionsfrom being controlled by through-thickness yielding of the column flanges. Thisis because it would be necessary to develop very large local ductilities in thecolumn flange material in order to accommodate even modest plastic rotationdemands on the assembly. However, extensive investigation of the through-thickness behavior of column flanges in a “T” joint configuration reveals thatneither yielding, nor through-thickness failure are likely to occur in theseconnections. Barsom and Korvink (1997) conducted a statistical survey ofavailable data on the tensile strength of rolled shape material in the through-thickness direction. These tests were generally conducted on small diametercoupons, extracted from flange material of heavy shapes. The data indicates thatboth the yield stress and ultimate tensile strength of this material in the through-thickness direction is comparable to that of the material in the direction parallelto rolling. However, it does indicate somewhat greater scatter, with a number ofreported values where the through-thickness strength was higher, as well as lowerthan that in the longitudinal direction. Review of this data indicates with highconfidence that for small diameter coupons, the yield and ultimate tensile valuesof the material in a through-thickness direction will exceed 90% and 80%respectively of the comparable values in the longitudinal direction. theThe causesfor through-thickness failures of column flanges (types C2, C4, and C5), observed



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both in buildings damaged by the Northridge Earthquake and in some testspecimens, are not well understood. They are thought to be a function of themetallurgy and “purity” of the steel; conditions of loading including the presenceof axial load and rate of loading application; conditions of tri-axial restraint;conditions of local hardening and embrittlement within the weld’s heat affectedzone; stress concentrations induced by the presence of backing bars and defectsat the root of beam flange to column flange welds; and by the relationship of theconnection components as they may affect flange bending stresses and flangecurvature induced by panel zone yielding. Given the many complex factors whichcan affect the through-thickness strength of the column flange, determination of areliable basis upon which to set permissible design stresses will requiresignificant research. Such research is currently being conducted under the SACphase II program.

While this statistical distribution suggests the likelihood that the through-thickness strength of column flanges could be less than the flexural strength ofattached beam elements, testing of more than 40 specimens at Lehigh Universityindicates that this is not the case. In these tests, high strength plates,representing beam flanges and having a yield strength of 100 ksi were welded tothe face of A572, Grade 50 and A913, Grade 50 column shapes, to simulate theportion of a beam-column assembly at the beam flange. These specimens wereplaced in a universal testing machine and loaded to produce high through-thickness tensile stresses in the column flange material. The tests simulated awide range of conditions, representing different weld metals as well and also toinclude eccentrically applied loading. In 40 of 41 specimens tested, the assemblystrength was limited by tensile failure of the high strength beam flange plate asopposed to the column flange material. In the one failure that occurred withinthe column flange material, fracture initiated at the root of a low-toughness weld,at root defects that were intentionally introduced to initiate such a fracture.

The behavior illustrated by this test series is consistent with mechanics ofmaterials theory. At the joint of the beam flange to column flange, the material isvery highly restrained. As a result of this, both the yield strength and ultimatetensile strength of the material in this region is significantly elevated. Underthese conditions, failure is unlikely to occur unless a large flaw is present that canlead to unstable crack propagation and brittle fracture. In light of this evidence,Interim Guidelines Advisory No. 2 deletes any requirement for evaluation ofthrough-thickness flange stress in columns.

Interim Recommendation No. 2 (SEAOC-1995) included a value of 40 ksi,applied to the projected area of beam flange attachment, for the through-thickness strength to be used in calculations. This value was selected because itwas consistent with the successful tests of cover plated assemblies conducted at



6-16

the University of Texas at Austin (Engelhardt and Sabol - 1994). However,because of the probable influence of all the factors noted above, this value canonly be considered to reflect the specific conditions of those tests and specimens.

Although reduced stresses at the column face produced acceptable results inthe University of Texas tests, the key to that success was more likely the result offorcing the plastic hinge away from the column than reduction of the through-thickness stress by the cover plates. Reduction of through-thickness columnflange stress to ever lower levels by the use of thicker cover plates is notrecommended, since such cover plates will result in ever higher forces on the faceof the column flange as well as larger weldments with potential for enlarged heataffected zones, higher residual stresses and conditions of restraint.

Since the initial publication of the Interim Guidelines, a significant number oftests have been performed on reduced beam section connections (See section7.5.3), most of which employed beam flanges which were welded directly to thecolumn flanges using improved welding techniques, but without reinforcementplates. No through-thickness failures occurred in these tests despite the fact thatcalculated through-thickness stresses at the root of the beam flange to columnflange joint ranged as high as 58 ksi. The successful performance of these weldedjoints is most probably due to the shifting of the yield area of the assembly awayfrom the column flange and into the beam span. Based on the indications of theabove described tests, and noting the undesirability of over reinforcingconnections, it is now suggested that a maximum through-thickness stress of0.9Fyc may be appropriate for use with connections that shift the hinging awayfrom the column face. Notwithstanding this recommendation, engineers are stillcautioned to carefully consider the through-thickness issue when these otherpreviously listed conditions which are thought to be involved in this type offailure are prevalent.

Notwithstanding all of the above, successful tests using cover plates and othermeasures of moving hinges (and coincidentally reducing through-thickness stress)continue to be performed. In the interim, structural engineers choosing to utilizeconnections relying on through-thickness strength should recognize that despitethe successful testing, connections relying on through-thickness strength can notbe considered to be fully reliable until the influence of the other parametersdiscussed above can be fully understood. A high amount of structuralredundancy is recommended for frames employing connections which rely onthrough-thickness strength of the column flange.

6.6.6.3.2 Determine Plastic Hinge Location

The desired location for the formation of plastic hinges should be determined as a basicparameter for the calculations. For beams with gravity loads representing a small portion of the



6-17

total flexural demand, the location of the plastic hinge may be assumed to occur as indicated inTable 6.6.6.3.2-1 and illustrated in Figure 6.6.6.3.2-1, at a distance equal to 1/3 of the beam depthfrom the edge of the reinforced connection (or start of the weakened beam section), unlessspecific test data for the connection indicates that a different value is appropriate. Refer to Figure6-13.

Table 6.6.6.3.2-1 Plastic Hinge Location - Strengthened Connections

Connection Type Reference Section Hinge Location “sh”

Cover plates Sect. 7.9.1 d/4 beyond end of cover plates

Haunches Sect. 7.9.3, 7.9.4 d/3 beyond toe of haunch

Vertical Ribs Sect. 7.9.2 d/3 beyond toe of ribs

L

Bea

m d

epth

- d

Edge

of r

einf

orce

dco

nnec

tion

Edge

of r

einf

orce

dco

nnec

tion

sh=d/3

L’

Plastichinge

Connectionreinforcementsh=

d/4

Figure 6-13 Figure 6.6.6.3.2-1 - Location of Plastic Hinge

Commentary: The suggested locations for the plastic hinge, at a distance d/3away from the end of the reinforced section indicated in Table 6.6.6.3.2-1 andFigure 6.6.6.3.2-1 are is based on the observed behavior of test specimens, withno significant gravity load present. If significant gravity load is present, this canshift the locations of the plastic hinges, and in the extreme case, even change theform of the collapse mechanism. If flexural demand on the girder due to gravityload is less than about 30% of the girder plastic capacity, this effect can safely beneglected, and the plastic hinge locations taken as indicated. If gravity demandssignificantly exceed this level then plastic analysis of the girder should be



6-18

performed to determine the appropriate hinge locations. Note that in zones ofhigh seismicity (UBC Zones 3 and 4, and NEHRP Map Areas 6 and 7) gravityloading on the girders of earthquake resisting frames typically has a very smalleffect.

6.6.6.3.3 Determine Probable Plastic Moment at Hinges

The probable value of the plastic moment, Mpr, at the location of the plastic hinges should bedetermined from the equation:

M 0.95 Z Fpr b ya= a (6-1)

M 1.1Z Fpr b ya= (6.6.6.3.3-1)

where: α is a coefficient that accounts for the effects of strain hardening and modelinguncertainty, taken as:

1.1 when qualification testing is performed or calculations are correlated with previous qualification testing

1.3 when design is based on calculations, alone.

Fya is the actual yield stress of the material, as identified from mill test reports. Wheremill test data for the project is not traceable to specific framing elements, theaverage of mill test data for the project for the given shape may be used. Whenmill test data for the project is not available, the value of Fym, fromtable 6-3Table 6.6.6.3-1 may be used.


Commentary: The 1.10.95 factor, in equation 6.6.6.3.3-1, is used to adjustaccount for two effects. First, it is intended to account for the typical differencebetween the yield stress in the beam web, where coupons for mill certificationtests are normally extracted, andto the value in the beam flange. Beam flanges,being comprised of thicker material, typically have somewhat lower yieldstrengths than do beam web material. Second, it is intended toThe factor of 1.1recommended to account for strain hardening, or other sources of strength aboveyield, and agrees fairly well with available test results. It should be noted that the1.1 factor could underestimate the over-strength where significant flangebuckling does not act as the gradual limit on the connection. Nevertheless, the1.1 factor seems a reasonable expectation of over-strength considering thecomplexities involved.



6-19

Connection designs that result in excessive strength in the girder connectionrelative to the column or excessive demands on the column panel zone are notexpected to produce superior performance. There is a careful balance that mustbe maintained between developing connections that provide for an appropriateallowance for girder overstrength and those that arbitrarily increase connectiondemand in the quest for a “conservative” connection design. The factorssuggested above were chosen in an attempt to achieve this balance, and arbitraryincreases in these values are not recommended.

When the Interim Guidelines were first published, Eq. 6.6.6.3.3-1 included acoefficient, α, intended to account both for the effects of strain hardening andalso for modeling uncertainty when connection designs were based oncalculations as opposed to a specific program of qualification testing. The intentof this modeling uncertainty factor was twofold: to provide additionalconservatism in the design when specific test data for a representative connectionwas not available, and also as an inducement to encourage projects to undertakeconnection qualification testing programs. After the Interim Guidelines had beenin use for some time, it became apparent that this approach was not an effectiveinducement for projects to perform qualification testing, and also that the use ofan overly large value for the α coefficient often resulted in excessively largeconnection reinforcing elements (cover plates, e.g.) and other design features thatdid not appear conducive to good connection behavior. Consequently, it wasdecided to remove this modeling uncertainty factor from the calculation of theprobable strength of an assembly.

6.6.6.3.4 Determine Beam Shear

The shear in the beam at the location of the plastic hinge should be determined. A free bodydiagram of that portion of the beam located between plastic hinges is a useful tool for obtainingthe shear at each plastic hinge. Figure 6-14Figure 6.6.3.4-1 provides an example of such acalculation.



6-20

L

sh

L’

Plastichinge P

L/2

P

Mpr MprL’Vp


“A”

VAw


Figure 6-14 Figure 6.6.3.4-1 - Sample Calculation of Shear at Plastic Hinge

6.6.6.3.5 Determine Strength Demands on Connection

In order to complete the design of the connection, including sizing the various plates andjoining welds which make up the connection, it is necessary to determine the shear and flexuralstrength demands at each critical section. These demands may be calculated by taking a free bodyof that portion of the connection assembly located between the critical section and the plastichinge. Figure 6-15 Figure 6.6.3.5-1 demonstrates this procedure for two critical sections, for thebeam shown in Figure 6-14Figure 6.6.3.4-1.

Plastichinge

Vp

Mpr

Plastichinge

Vp

Mpr

x

Mf

x+dc/2

dc

Mf=Mpr +Vpx

Mc

Mc=Mpr +Vp(x+dc/2)


Figure 6-15 Figure 6.6.3.5-1 - Calculation of Demands at Critical Sections



6-21

Commentary: Each unique connection configuration may have different criticalsections. The vertical plane that passes through the joint between the beamflanges and column (if such joining occurs) will typically define at least one suchcritical section, used for designing the joint of the beam flanges to the column, aswell as evaluating shear demands on the column panel zone. A second criticalsection occurs at the center line of the column. Moments calculated at this pointare used to check weak beam - strong column conditions. Other critical sectionsshould be selected as appropriate to the connection configuration.

6.6.6.3.6 Check for Strong Column - Weak Beam Condition

Buildings which form sidesway mechanisms through the formation of plastic hinges in thebeams can dissipate more energy than buildings that develop mechanisms consisting primarily ofplastic hinges in the columns. Therefore, if an existing building’s original design was such thathinging would occur in the beams rather than the columns, care should be taken not to alter thisbehavior with the addition of connection reinforcement. To determine if the desired strongcolumn - weak beam condition exists, the connection assembly should be checked to determine ifthe following equation is satisfied:

Z (F f ) M 1.0c yc a c− >∑ ∑ (6.6.6.3.6-12)

where: Zc is the plastic modulus of the column section above and below the connectionFyc is the minimum specified yield stress for the column above and belowfa is the axial load in the column above and belowΣMc is the moment calculated at the center of the column in accordance with

Section 6.6.6.3.5 sum of the column moments at the top and bottom of the panel zone, respectively, resulting from the development of the probable beam plastic moments, Mpr, within each beam in the connection.

Commentary: Equation 6.6.6.3.6-12 is based on the building code provisions forstrong column - weak beam design. The building code provisions for evaluatingstrong column - weak beam conditions presume that the flexural stiffness of thecolumns above and below the beam are approximately equal, that the beams willyield at the face of the column, and that the depth of the columns and beams aresmall relative to their respective span lengths. This permits the code to use arelatively simple equation to evaluate strong column - weak beam conditions inwhich the sum of the flexural capacities of columns at a connection are comparedto the sums of the flexural capacities in the beams. The first publication of theInterim Guidelines took this same approach, except that the definition of ΣMc wasmodified to explicitly recognize that because flexural hinging of the beams wouldoccur at a location removed from the face of the column, the moments deliveredby the beams to the connection would be larger than the plastic moment strengthof the beam. In this equation, ΣMc was taken as the sum of the moments at the



6-22

center of the column, calculated in accordance with the procedures of Sect.6.6.3.5.

This simplified approach is not always appropriate. If non-symmetricalconnection configurations are used, such as a haunch on only the bottom side ofthe beam, this can result in an uneven distribution of stiffness between the twocolumn segments, and premature yielding of the column, either above, or below,the beam-column connection. Also, it was determined that for connectionconfigurations in which the panel zone depth represents a significant fraction ofthe total column height, such as can occur in some haunched and side-platedconnections, the definition of ΣMc contained in the initial printing of theGuidelines could lead to excessive conservatism in determining whether or not astrong column - weak beam condition exists in a structure. Consequently, InterimGuidelines Advisory No. 1 adopted the current definition of ΣMc for use in thisevaluation. This definition requires that the moments in the column, at the topand bottom of the panel zone be determined for the condition when a plastichinge has formed at all beams in the connection. Figure 6.6.6.3.6-1 illustrates amethod for determining this quantity. In such cases, When evaluation indicatesthat a strong column - weak beam condition does not exist, a plastic analysisshould be considered to determine if an undesirable story mechanism is likely toform in the building.

(L-L’)/2

d ph t

h b

Mpr

Vp

Vp

Mpr

Vc

Vc+Vf

Mct

Mcb

assumed point of zero moment

Note:The quantities Mpr, Vp, L, and L’ areas previously identified. Vf is the incremental shear distributedto the column at the floor level.Other quantities are as shown.

Vf

( )[ ] ( )

( )

VM V L L V h d

h d h

M V h

M V V h

M M M

cpr p f b p

b p t

ct c t

cb c f b

c ct cb

=+ − − +

+ +== +

= +

∑

∑

' ) / /2 2

Figure 6.6.6.3.6-1 Calculation of Column Moment for Strong Column Evaluation



6-23

6.6.6.3.7 Check Column Panel Zone

The adequacy of the shear strength of the column panel zone should be checked. For thispurpose, the term 0.8ΣMf should be substituted for the term 0.8ΣMs in UBC-94 Section2211.7.2.1 {0.9ΣφbMp in NEHRP-91 Section 10.10.3.1} repeated below for convenience ofreference. Mf is the calculated moment at the face of the column, when the beam mechanismforms, calculated as indicated in Section 6.6.6.3.5, above. In addition, it is recommended not touse the alternative design criteria indicated in UBC-94 Section 2211.7.2.1 (NEHRP-91 Sect.10.10.3.1), permitting panel zone shear strength to be proportioned for the shear induced bybending moments from gravity loads plus 1.85 times the prescribed seismic forces. Forconvenience of reference, UBC-94 Section 2211.7.2.1 is reproduced below, edited, to indicate therecommended application:

2211.7.2.1 Strength (edited). The panel zone of the joint shall be capable of resisting theshear induced by beam bending moments due to gravity loads plus 1.85 times theprescribed seismic forces, but the shear strength need not exceed that required to develop0.8ΣMs 0.8ΣMf of the girders framing into the column flanges at the joint. The joint panelzone shear strength may be obtained from the following formula:

V 0.55F d t 13b td d ty c

c c f2

b c

= +

(11-1)

where: bc = width of column flangedb = the depth of the beam (including haunches or cover plates)dc = the depth of the columnt = the total thickness of the panel zone including doubler platestcf = the thickness of the column flange

Commentary: The effect of panel zone shear yielding on connection behavior isnot well understood. In the past, panel zone shear yielding has been viewed as abenign mechanism that permits overall frame ductility demands to beaccommodated while minimizing the extent of inelastic behavior required of thebeam and beam flange to column flange joint. The criteria permitting panel zoneshear strength to be proportioned for the shears resulting from moments due togravity loads plus 1.85 times the design seismic forces was adopted by the codespecifically to encourage designs with weak panel zones. However, during recenttesting of large scale connection assemblies with weak panel zones, it has beennoted that in order to accommodate the large shear deformations that occur inthe panel zone, extreme “kinking” deformations were induced into the columnflanges at the beam flange to column flange welded joint. While this did not leadto premature joint failure in all cases, it is believed to have contributed to suchpremature failures in at least some of the specimens. The recommendations ofthis section are intended to result in stronger panel zones than previously



6-24

permitted by the code, thereby avoiding potential failures due to this kinkingaction on the column flanges.

6.6.7 Modification Details


6.6.7.1 Haunch at Bottom Flange

Figure 6-166.6.7.1-1 illustrates the basic configuration for a connection modificationconsisting of the addition of a welded haunch at the bottom beam flange. Several tests of such amodification were conducted by Uang under the SAC phase I project (Uang, 1995). Followingthat work, additional research on the feasibility of improving connection performance with weldedhaunches was conducted under a project that was jointly sponsored by NIST and AISC (NIST,1998). As indicated in the report of that work, the haunched modification improves connectionperformance by altering the basic behavior of the connection. In essence, the haunch creates aprop type support, beneath the beam bottom flange. This both reduces the effective flexuralstresses in the beam at the face of the support, and also greatly reduces the shear that must betransmitted to the column through the beam. Laboratory tests indicate this modification can beeffective when the existing low-toughness welds between the beam bottom flange and column areleft in place, however, more reliable performance is obtained when the top welds are modified. Acomplete procedure for the design of this modification may be found in NIST, 1998. twoalternative configurations of this detail that have been tested (Uang - 1995). The basic concept isto reinforce the connection with the provision of a triangular haunch at the bottom flange. Theintended behavior of both configurations is to shift the plastic hinge from the face of the columnand to reduce the demand on the CJP weld by increasing the effective depth of the section. In onetest, shown on the left of Figure 6-16, the joint between the girder bottom flange and column wascut free, to simulate a condition which might occur if the bottom joint had been damaged, but notrepaired. In a second tested configuration, the bottom flange joint was repaired and the top flangewas replaced with a locally thickened plate, similar to the detail shown in Figure 6-9.

Design Issues: This approach developed acceptable levels of plastic rotation. Acceptable levelsof connection strength were also maintained during large inelastic deformations of the plastichinge. This approach does not require that the top flange be modified, or slab disturbed, unlessother conditions require repair of the top flange, as in the detail on the left of Figure 6-16. Thebottom flange is generally far more accessible than the top flange because a slab does not haveto be removed. In addition, the haunch can be installed at perimeter frames without removal ofthe exterior building cladding. There did not appear to be any appreciable degradation inperformance when the bottom beam flange was not re-welded to the face of the column.Eliminating this additional welding should help reduce the cost of the repair.

Performance is dependent on properly executed complete joint penetration welds at the columnface and at the attachment of the haunch to the girder bottom flange. The joint can be subject tothrough-thickness flaws in the column flange; however, this connection may not be as sensitive



6-25

to this potential problem because of the significant increase in the effective depth of the beamsection which can be achieved. Welding of the bottom haunch requires overhead welding. Theskewed groove welds of the haunch flanges to the girder and column flanges may be difficult toexecute.

Experimental Results: This approach developed excellent levels of plastic rotation. In Specimen1, the bottom flange CJP weld was damaged in a prior test but was not repaired: only the bottomhaunch was added. During the test of specimen 1, a slowly growing crack developed at theunderside of the top flange-web intersection, perhaps exacerbated by significant local bucklingof the top flange. Some of the buckling may be attributed to lateral torsional buckling thatoccurred because the bottom flange was not restrained by a CJP weld. A significant portion ofthe flexural strength was lost during the cycles of large plastic rotation. In the second specimen,the bottom girder flange weld was intact during the haunch testing, and its performance wassignificantly improved compared with the first specimen. The test was stopped when significantlocal buckling led to a slowly growing crack at the beam flange and web intersection. At thistime, it appears that repairing damaged bottom flange welds in this configuration can producebetter performance. Acceptable levels of flexural strength were maintained during largeinelastic deformations of the plastic hinge for both specimens. As reported in NIST, 1998, a totalof 9 beam-column connection tests incorporating bottom haunch modifications of pre-Northridge connections have been tested in the laboratory, including two dynamic tests. Most ofthe connection assemblies tested resisted in excess of 0.02 radians of imposed plastic rotation.However, for those specimens in which the existing low-toughness weld was left in place at thebeam top flange, without modification, connection behavior was generally limited by fracturesgenerating at these welds at relatively low plastic rotations. It may be expected that enhancedperformance can be obtained by replacing or reinforcing these welds as part of the modification.

Figure 6-166.6.7.1-1 - Bottom Haunch Connection Modification



6-26


Specimen 1 UCSD-1R: 0.04 radian (w/o bottom flange weld)Specimen 2 UCSD-3R:0.05 radian (with bottom flange weld)

Specimen UCSD-4R: 0.014 radian (dynamic- limited by test setup)Speciemn UCSD-5R: 0.015 radian (dynamic- limited by test setup)

Girder Size: W36x150Column Size: W14x257Plastic Rotation achieved -

Specimen UCB-RN2: 0.014 radian (no modification of top weld)Specimen UTA-1R: 0.019 radian (partial modification of top weld)Specimen UTA-1RB: 0.028 radian (modified top weld)

Girder Size: W36x150Column Size: W14x455Plastic Rotation achieved-

Specment UTA-NSF4: 0.015 radian (no modification of top weld)


Specimen SFCCC-8: 0.035 radian (cover plated top flange)

6.6.7.2 Top and Bottom Haunch


6.6.7.3 Cover Plate Sections

Figure 6.6.7.3-1 Figure 6-18 illustrates the basic configurations of cover plate connections.The assumption behind the cover plate is that it reduces the applied stress demand on the weld atthe column flange and shifts the plastic hinge away from the column face. Only the connectionwith cover plates on the top of the top flange has been tested. There are no quantitative resultsfor cover plates on the bottom side of the top flange, such as might be used in repair. It is likelythat thicker plates would be required where the plates are installed on the underside of the topflange. The implications of this deviation from the tested configuration should be considered.



6-27

Top &Bottom

Near and Far Sides

Top &Bottom

d

d/2, typical

Figure 6-18 Figure 6.6.7.3-1 - Cover Plate Connection Modification

Design Issues: Following the Northridge earthquake, the University of Texas at Austinconducted a program of research, under private funding, to develop a modified connectionconfiguration for a specific project. Following a series of unsuccessful tests on various types ofconnections,Approximately eight connections similar to that shown in Figure 6-18Figure 6.7.3-1have been were tested (Engelhardt & Sabol - 1994), and have demonstrated the ability toachieve acceptable levels of plastic rotation provided that the beam flange to column flangewelding wasis correctly executed and through-thickness problems in the column flange wereareavoided. Due to the significant publicity that followed these successful tests, as well as theeconomy of these connections relative to some other alternatives, cover plated connectionsquickly became the predominant configuration used in the design of new buildings. As a result,a number of qualification tests have now been performed on different variations of cover platedconnections, covering a wide range of member sizes ranging from W16 to W36 beams, as part ofthe design process for individual building projects. The results of these tests have beensomewhat mixed, with a significant number of failures reported. Although this connection typeappears to be significantly more reliable than the typical pre-Northridge connection, it should beexpected that some connections in buildings incorporating this detail may still be subjected toearthquake initiated fracture damage. Designers should consider using alternative connectiontypes, unless highly redundant framing systems are employed.

The option with the top flange cover plate located on top of the flange can be used onperimeter frames where access to the outer side of the beam is restricted by existing buildingcladding. The option with the cover plate for the top flange located beneath the flange can beinstalled without requiring modification of the slab. In the figures shown, the bottom cover plateis rectangular, and sized slightly wider than the beam flange to allow downhand fillet welding ofthe joint between the two plates. Some configurations using triangular plates at the bottomflange, similar to the top flange have also been tested.



6-28

Designers using this detail are cautioned to be mindful of not making cover plates so thickthat excessively large welds of the beam flange combination to column flange result. As thecover plates increase in size, the weld size must also increase. Larger welds invariably result ingreater shrinkage stresses and increased potential for cracking prior to actual loading. Inaddition, larger welds will lead to larger heat affected zones in the column flange, a potentiallybrittle area.

Performance is dependent on properly executed girder flange welds. The joint can be subjectto through-thickness failures in the column flange. Access to the top of the top flange requiresdemolition of the existing slab. Access to the bottom of the top flange requires overhead weldingand may be problematic for perimeter frames. Costs are greater than those associated withapproaches that concentrate modifications on the bottom flange

Experimental Results: Six of eight connections tested by the University of Texas at Austin wereable to achieve plastic rotations of at least 0.025 radians, or better. These tests were performedusing heavy column sections which forced nearly all of the plastic deformation into the beamplastic hinge; very little column panel zone deformation occurred. Strength loss at the extremelevels of plastic rotation did not reduce the flexural capacity to less than the plastic momentcapacity of the section based on minimum specified yield strength. One specimen achievedplastic rotations of 0.015 radians when a brittle fracture of the CJP weld (type W2 failure)occurred. This may partially be the result of a weld that was not executed in conformance withthe specified welding procedure specification. The second unsuccessful test specimen achievedplastic rotations of 0.005 radian when a section of the column flange pulled out (type C2failure). The successful tests were terminated either when twisting of the specimen threatened todamage the test setup or the maximum stroke of the loading ram was achieved. Since thecompletion of that testing, a number of additional tests have been performed. Data for 18 tests,including those performed by Engelhardt and referenced above, are in the public domain. Atleast 12 other tests have been performed on behalf of private parties, however, the data fromthese tests are not available. Some of those tests exhibited premature fractures.

Quantitative Results: No. of specimens tested: 18Girder Size: W21 x 68 to W36 x 150Column Size: W12 x 106 to W14 x 455, and 426Plastic Rotation achieved-

6 13 Specimens : >.025 radian to 0.05 radian13 Specimens: 0.005 < θp < 0.0250.015 radian (W2 failure)12 Specimens: 0.005 radian (C2 failure)

6.6.7.4 Upstanding Ribs




6-29

6.6.7.5 Side-Plate Connections


6.6.7.6 Bolted Brackets

Heavy bolted brackets, incorporating high strength bolts, may be added to existing weldedconnections to provide an alternative load path for transfer of stress between the beams andcolumns. To be compatible with existing welded connections, the brackets must have sufficientstrength and rigidity to transfer beam stresses with negligible deformation. Pre-tensioning of thebolts or threaded rods attaching the brackets to the column flanges and use of welds or slip-critical connections between the brackets and beam flanges can help to minimize deformationunder load. Reinforcement of the column flanges may be required to prevent local yielding andexcessive deformation of these elements. Two alternative configurations, which may be usedeither to repair an existing damaged, welded connection or to reinforce an existing undamagedconnection are illustrated in Figure 6.6.7.6-1. The developer of these connections offers thebrackets in the form of proprietary steel castings. Several tests of these alternative connectionshave been performed on specimens with beams ranging in size from W16 to W36 sections andwith large plastic rotations successfully achieved. Under a project jointly funded by NIST andAISC, the use of a single bracket at the bottom flange of the beam was investigated. It wasdetermined that significant improvement in connection behavior could be obtained by placing abracket at the bottom beam flange and by replacing existing low-toughness welds at the top flangewith tougher material. NIST, 1998 provides a recommended design procedure for suchconnection modifications.

Design Issues: The concept of bolted bracket connections is similar to that of the riveted “windconnections” commonly installed in steel frame buildings in the early twentieth century. Theprimary difference is that the riveted wind connections were typically limited in strength eitherby flexural yielding of outstanding flanges of the brackets, or shear and tension on the rivets,rather than by flexural hinging of the connected framing. Since the old-style wind connectionscould not typically develop the flexural strength of the girders and also could be quite flexible,they would be classified either as partial strength or partially restrained connections. Followingthe Northridge earthquake, the concept of designing such connections with high strength boltsand heavy plates, to behave as fully restrained connections, was developed and tested by aprivate party who has applied for patents on the concept of using steel castings for this purpose.

Bolted bracket connections can be installed in an existing building without field welding. Sincethis reduces the risk of construction-induced fire, brackets may be installed with somewhat lessdemolition of existing architectural features than with welded connections. In addition, thequality assurance issues related to field welding are eliminated. However, the fabrication of thebrackets themselves does require quality assurance. Quality assurance is also required foroperations related to the drilling of bolt holes for installation of bolts, surface preparation offaying surfaces and for installation and tensioning of the bolts themselves.



6-30

PipePlate

Bolts

High tensilethreaded rod

Bolts

Steelcasting

WARNING: The information presented in this figure is PROPRIETARY. US and ForeignPatents have been applied for. Use of this information is strictly prohibited except as authorizedin writing by the developer. Violators shall be prosecuted in accordance with US and ForeignPatent Intellectual Property Laws.

Figure 6.6.7.6-1 Bolted Bracket Modification

Bolted brackets can be used to repair damaged connections. If damage is limited to the beamflange to column flange welds, the damaged welds should be dressed out by grinding. Anyexisting fractures in base metal should be repaired as indicated in Section 6.3, in order torestore the strength of the damaged members and also to prevent growth of the fractures underapplied stress. Since repairs to base metal typically require cutting and welding, this reducessomewhat the advantages cited above, with regard to avoidance of field welding.

Experimental Results: A series of tests on several different configurations of proprietary heavybolted bracket connections have been performed at Lehigh University (Ksai & Bleiman, 1996) toqualify these connections for use in repair and modification applications. To test repairapplications, brackets were placed only on the bottom beam flange to simulate installations on aconnection where the bottom flange weld in the original connection had failed. In thesespecimens, bottom flange welds were not installed, to approximate the condition of a fullyfractured weld. The top flange welds of these specimens were made with electrodes rated fornotch toughness, to preclude premature failure of the specimens at the top flange. Forspecimens in which brackets were placed at both the top and bottom beam flanges, both weldswere omitted. Acceptable plastic rotations were achieved for each of the specimens tested. Notesting has yet been performed to determine the effectiveness of bolted brackets when applied toan existing undamaged connection with full penetration beam flange to column flange welds withlow toughness or significant defects or discontinuities.

Quantitative Results: No. of specimens tested: 8Girder Size: W16x40 and W36x150Column Size: W12x65 and W14x425Plastic Rotation achieved - 0.05 radians - 0.07 radians



6-31



New Construction

7-1

7. NEW CONSTRUCTION

7.1 Scope

This Chapter presents interim design guidelines for new welded steel moment frames(WSMFs) intended to resist seismic demands through inelastic behavior. The criteria apply to allSMRF structures designed for earthquake resistance and those IMRF and OMRF structureslocated in Uniform Building Code (UBC) Seismic Zones 3 and 4 {National Earthquake HazardsReduction Program (NEHRP) Map Areas 6 and 7} or assigned to 1997 NEHRP Seismic DesignCategories D, E, or F. Light, single-story buildings, the design of which is governed by wind,need not consider these Interim Guidelines. Frames with bolted connections, either fullyrestrained (type FR) or partially restrained (type PR), are beyond the scope of this document. However, the acceptance criteria for connections may be applied to type FR bolted connections aswell.

Commentary: Observation of damage experienced by WSMF buildings in theNorthridge Earthquake and subsequent laboratory testing of large scale beam-column assemblies has demonstrated that the standard details for WSMFconnections commonly used in the past are not capable of providing reliableservice in the post-elastic range. Therefore, structures which are expected toexperience significant post-elastic demands from design earthquakes, or forwhich highly reliable seismic performance is desired, should be designed usingthe Interim Guidelines presented herein.

In order to determine if a structure will experience significant inelasticbehavior in a design earthquake, it is necessary to perform strength checks of theframe components for the combination of dead and live loads expected to bepresent, together with the full earthquake load. Except for structures with specialperformance goals, or structures located within the near field (within 10kilometers) of known active earthquake faults, the full earthquake load may betaken as the minimum design earthquake load specified in the building code, butcalculated using a lateral force reduction coefficient (Rw or R) of unity. If allcomponents of the structure and its connections have adequate strength to resistthese loads, or nearly so, then the structure may be considered to be able to resistthe design earthquake, elastically.

Design of frames to remain elastic under unreduced (Rw {R} taken as unity)earthquake forces may not be an overly oppressive requirement, particularly inmore moderate seismic zones. Most frame designs are currently controlled bydrift considerations and have substantially more strength than the minimumspecified for design by the building code. As part of the SAC Phase 1 research, anumber of modern frame buildings designed with large lateral force reductioncoefficients (Rw = 12, {R = 8}) were evaluated for unreduced forces calculated


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using the standard building code spectra. It was determined that despite thenominally large lateral force reduction coefficients used in the original design,the maximum computed demands from the dynamic analyses were only on theorder of 2 to 3 times those which would cause yielding of the real structures(Krawinkler, et. al. - 1995; Uang, et. al. - 1995; Engelhardt, et. al. - 1995, Hart,et. al. - 1995; Kariotis and Eimani - 1995). Therefore, it is not unreasonable toexpect that OMRF structures (nominally designed with a lateral force reductioncoefficient Rw = 6 {R = 4.5}) could resist the design earthquakes with near elasticbehavior. Regardless of these considerations, better seismic performance can beexpected by designing structures with greater ductility rather than less, andengineers are not encouraged to design structures for elastic behavior usingbrittle or unreliable details..

For structures designed to meet special performance goals, and buildingslocated within the near field of major active faults, full earthquake loadscalculated in accordance with the above procedure may not be adequate. Forsuch structures, the full earthquake load should be determined using a sitespecific ground motion characterization and a suitable analysis procedure. Recent research (Heaton, et. al. - 1995) suggests that the elastic responsespectrum technique, typically used for determining seismic forces for structuraldesign, may not provide an adequate indication of the true earthquake demandsproduced by the large impulsive ground motions common in the near field oflarge earthquake events. Further, this research indicates that frame structures,subjected to such impulsive ground motions can experience very large drifts, andpotential collapse. In an attempt to address this, both the 1997 edition of theUniform Building Code and the 1997 edition of the NEHRP Provisions specifydesign ground motions for structures located close to major active faults that aresubstantially more severe than those contained in earlier codes. While the moresevere ground motion criteria contained in these newer provisions are probablyadequate for the design of most structures, analytical studies conducted by SACconfirm that even structures designed to these criteria can experience very largedrift demands, and potentially collapse, if the dynamic characteristics of theimpulsive loading and those of the structure are matched. Direct nonlinear timehistory analysis, using an appropriate ground motion representation would beone method of more accurately determining the demands on structures located inthe near field. Additional research on these effects is required.

As an alternative to use of the criteria contained in these Interim Guidelines,OMRF structures in zones of high seismicity (UBC seismic zones 3 and 4 andNEHRP map areas 6 and 7) and OMRF structures assigned to 1997 NEHRPSeismic Design Categories D, E or F, may be designed for the connections toremain elastic (Rw or R taken as 1.0) while the beams and columns are designedusing the standard lateral force reduction coefficients specified by the building


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code. Although this is an acceptable approach, it may result in much largerconnections than would be obtained by following these Interim Guidelines.

The use of partially restrained connections may be an attractive andeconomical alternative to the design of frames with fully restrained connections. However, the design of frames with partially restrained connections is beyond thescope of this document. The AISC is currently working on development ofpractical design guidelines for frames with partially restrained connections.

7.2 General - Welded Steel Frame Design Criteria

7.2.1 Criteria

Welded Steel Moment Frame (WSMF) systems should, as a minimum, be designed for theprovisions of the prevailing building code and these Interim Guidelines. Special Moment-Resisting Frames (SMRF)s and Ordinary Moment-Resisting Frames (OMRF)s with FRconnections, should additionally be designed in accordance with either the 1997 edition of theAISC Seismic Provisions for Structural Steel Buildings (AISC, 1997) or the emergency codechange to the 1994 UBC {NEHRP-1994}, restated as follows:

2211.7.1.1. Required Strength {NEHRP-1994 Section 5.2, revision to Ref. 8.2c of Ref. 5.3}

The girder-to-column connections shall be adequate to develop the lesser of the following:


2. The moment corresponding to development of the panel zone shear strength as determined by Formula (11-1).

2211.7.1.3-2 Connection Strength

Connection configurations utilizing welds and high strength bolts shall demonstrate, by approved cyclic test results orcalculation, the ability to sustain inelastic rotations and to develop the strength criteria in Section 2211.7.1.1considering the effects of steel overstrength and strain hardening.

Commentary: At the time the Interim Guidelines were first published, they werebased on the 1994 edition of the Uniform Building Code and the 1994 edition ofthe NEHRP Provisions. In the time since that initial publication, more recenteditions of both documents have been published, and codes based on thesedocuments have been adopted by some jurisdictions. In addition, the AmericanInstitute of Steel Construction has adopted a major revision to its SeismicProvisions for Structural Steel Buildings (AISC Seismic Provisions), largelyincorporating, with some modification, the recommendations contained in theInterim Guidelines. This updated edition of the AISC Seismic Provisions hasbeen incorporated by reference into the 1997 edition of the NEHRP Provisionsand has also been adopted by some jurisdictions as an amendment to the modelbuilding codes. SMRF and OMRF systems that are designed to comply with therequirements of the 1997 AISC Seismic Provisions may be deemed to comply withthe intent of these Interim Guidelines. Where reference is made herein to the


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requirements of the 1994 Uniform Building Code or 1994 NERHP Provisions, theparallel provisions of the 1997 editions may be used instead, and should be usedin those jurisdictions that have adopted codes based on these updated standards.

The 1997 edition NEHRP Provisions and AISC Seismic Provisions introducea new structural system termed an Intermediate Moment Resisting Frame (IMRF). Provisions for IMRF structures include somewhat more restrictive detailing anddesign requirements than those for OMRF structures, and less than those forSMRF structures. The intent was to provide a system that would be moreeconomical than SMRF structures yet have better inelastic response capabilitythan OMRF structures. The SAC project is currently conducting research todetermine if the provisions for the new IMRF system are adequate, but has notdeveloped a position on this at this time.

At this time, no recommendations are made to change the minimum lateralforces, drift limitations or strength calculations which determine member sizingand overall performance of moment frame systems, except as recommended inSections 7.2.2, 7.2.3 and 7.2.4. The design of joints and connections is discussedin Section 7.3. The UBC permits OMRF structures with FR connections,designed for 3/8Rw times the earthquake forces otherwise required, to bedesigned without conforming to Section 2211.7.1. However, this is notrecommended.

7.2.2 Strength and Stiffness

7.2.2.1 Strength



Flexure Ms = Z Fy



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Commentary: Partial penetration welds are not recommended for tensionapplications in critical connections resisting seismic-induced stresses. Thegeometry of partial penetration welds creates a notch-like condition that caninitiate brittle fracture under conditions of high tensile strain.

Many WSMF structures are constructed with concrete floor slabs that areprovided with positive shear attachment between the slab and the top flanges ofthe girders of the moment-resisting frames. Although not generally accounted forin the design of the frames, the resulting composite action can increase theeffective strength of the girder significantly, particularly at sections wherecurvature of the girder places the top flange into compression. Although thiseffect is directly accounted for in the design of composite systems, it is typicallyneglected in the design of systems classified as moment resisting steel frames. The increased girder flexural strength caused by this composite action can resultin a number of effects including the unintentional creation of weak column -strong beam and weak panel zone conditions. In addition, this composite effecthas the potential to reduce the effectiveness of reduced section or “dog-bone”type connection assemblies. Unfortunately, very little laboratory testing of largescale connection assemblies with slabs in place has been performed to date andas a result, these effects are not well quantified. In keeping with typicalcontemporary design practice, the design formulae provided in these Guidelinesneglect the strengthening effects of composite action. Designers should, however,be alert to the fact that these composite effects do exist.

7.2.2.2 Stiffness

Calculation of frame stiffness for the purpose of determining interstory drift under theinfluence of the design lateral forces should be based on the properties of the bare steel frame,neglecting the effects of composite action with floor slabs. The stiffening effects of connectionreinforcements (e.g.: haunches, side plates, etc.) may be considered in the calculation of overallframe stiffness and drift demands. When reduced beam section connections are utilized, thereduction in overall frame stiffness, due to local reductions in girder cross section, should beconsidered.

Commentary: For design purposes, frame stiffness is typically calculatedconsidering only the behavior of the bare frame, neglecting the stiffening effectsof slabs, gravity framing, and architectural elements. The resulting calculation ofbuilding stiffness and period typically underestimates the actual properties,substantially. Although this approach can result in unconservative estimates ofdesign force levels, it typically produces conservative estimates of interstory driftdemands. Since the design of most moment-resisting frames are controlled byconsiderations of drift, this approach is considered preferable to methods thatwould have the potential to over-estimate building stiffness. Also, many of the


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elements that provide additional stiffness may be subject to rapid degradationunder severe cyclic lateral deformation, so that the bare frame stiffness mayprovide a reasonable estimate of the effective stiffness under long durationground shaking response.

Notwithstanding the above, designers should be alert to the fact thatunintentional stiffness introduced by walls and other non-structural elements cansignificantly alter the behavior of the structure in response to ground shaking. Ofparticular concern, if these elements are not uniformly distributed throughout thestructure, or isolated from its response, they can cause soft stories and torsionalirregularities, conditions known to result in poor behavior.

7.2.3 Configuration

Frames should be proportioned so that the required plastic deformation of the frame can maybe accommodated through the development of plastic hinges at pre-determined locations withinthe girder spans, as indicated in Figure 7-1Figure 7.2.3-1. Beam-column connections should bedesigned with sufficient strength (through the use of cover plates, haunches, side plates, etc.) toforce development of the plastic hinge away from the column face. This condition may also beattained through local weakening of the beam section at the desired location for plastic hingeformation.

Plastic Hinges


L’

L

h

drift angle - θ

Figure 7-1 Figure 7.2.3-1 - Desired Plastic Frame Behavior

Commentary: Nonlinear deformation of frame structures is typicallyaccommodated through the development of inelastic flexural or shear strainswithin discrete regions of the structure. At large inelastic strains these regionscan develop into plastic hinges, which can accommodate significant concentratedrotations at constant (or nearly constant) load through yielding at tensile andcompressive fibers and by buckling at compressive fibers. If a sufficient numberof plastic hinges develop in a frame, a mechanism is formed and the frame can


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deform laterally in a plastic manner. This behavior is accompanied by significantenergy dissipation, particularly if a number of members are involved in theplastic behavior, as well as substantial local damage to the highly strainedelements. The formation of hinges in columns, as opposed to beams, isundesirable, as this results in the formation of weak story mechanisms withrelatively few elements participating, so called “story mechanisms” andconsequently little energy dissipation occurring. In addition, such mechanismsalso result in local damage to critical gravity load bearing elements.

The prescriptive connection contained in the UBC and NEHRP RecommendedProvisions prior to the Northridge Earthquake was based on the assumeddevelopment of plastic hinge zones within the beams at adjacent to the face of thecolumn, or within the column panel zone itself. If the plastic hinge develops inthe column panel zone, the resulting column deformation results in very largesecondary stresses on the beam flange to column flange joint, a condition whichcan contribute to brittle failure. If the plastic hinge forms in the beam, at the faceof the column, this can result in very large through-thickness strain demands onthe column flange material and large inelastic strain demands on the weld metaland surrounding heat affected zones. These conditions can also lead to brittlejoint failure. Although ongoing research may reveal conditions of materialproperties, design and detailing configurations that permit connections withyielding occurring at the column face to perform reliably, for the present it isrecommended In order to achieve more reliable performance, it is recommendedthat the connection of the beam to the column be configured to force the inelasticaction (plastic hinge) away from the column face. This can be done either bylocal reinforcement of the connection, or locally reducing the cross section of thebeam at a distance away from the connection. Plastic hinges in steel beams havefinite length, typically on the order of half the beam depth. Therefore, thelocation for the plastic hinge should be shifted at least that distance away fromthe face of the column. When this is done through reinforcement of theconnection, the flexural demands on the columns, for a given beam size, areincreased. Care must be taken to assure that weak column conditions are notinadvertently created by local strengthening of the connections.

It should be noted that some professionals and researchers believe thatconfigurations which permit plastic hinging to occur adjacent to the column facemay still provide reliable service under some conditions. These conditions mayinclude limitations on the size of the connected sections, the use of base and weldmetals with adequate notch toughness, joint detailing that minimizes notch effects,and appropriate control of the relative strength of the beam and columnmaterials. Sufficient research has not been performed to date either to confirmthese suggestions or define the conditions in which they are valid. Researchhowever does indicate that reliable performance can be attained if the plastic


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hinge is shifted away from the column face, as suggested above. Consequently,these Interim Guidelines make a general recommendation that this approach betaken. Additional research should be performed to determine the acceptability ofother approaches.

It should also be noted that reinforced connection (or reduced beam section)configurations of the type described above, while believed to be effective inpreventing brittle connection fractures, will not prevent structural damage fromoccurring. Brittle connection fractures are undesirable because they result in asubstantial reduction in the lateral-force-resisting strength of the structure which,in extreme cases, can result in instability and collapse. Connections configuredas described in these Interim Guidelines should experience many fewer suchbrittle fractures than unmodified connections. However, the formation of aplastic hinge within the span of a beam is not a completely benign event. Beamswhich have formed such hinges may, if plastic rotations are large, exhibitsignificantlarge buckling and yielding deformation, damage which typically mustbe repaired. The cost of such repairs could be comparable to the costs incurredin repairing fracture damage experienced in the Northridge Earthquake. Theprimary difference is that life safety protection will be significantly enhanced andmost structures that have experienced such plastic deformation damage shouldcontinue to be safe for occupancy while repairs are made.

If the types of damage described above are unacceptable for a given building,then alternative structural systems should be considered that will reduce theplastic deformation demands on the structure during a strong earthquake. Appropriate methods of achieving such goals include the installation ofsupplemental braced frames, energy dissipation systems, base isolation systemsand similar structural systems. Framing systems incorporating partiallyrestrained connections may also be quite effective in resisting large earthquakeinduced deformation with limited damage.

It is important to recognize that in frames with relatively short bays, theflexural hinging indicated in Figure 7.2.3-1 may not be able to form. If theeffective flexural length (L’ in the figure) of beams in a frame becomes too short,then the beams or girders will yield in shear before zones of flexural plasticitycan form, resulting in an inelastic behavior that is more like that of aneccentrically braced frame than that of a moment frame. This behavior mayinadvertently occur in frames in which relatively large strengthened connections,such as haunches, cover plates or side plates have been used on beams withrelatively short spans. This behavior is illustrated in Figure 7.2.3-2.

The guidelines contained in this section are intended to address the design offlexurally dominated moment resisting frames. When utilizing these guidelines, itis important to confirm that the configuration of the structure is such that the


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presumed flexural hinging can actually occur. It is possible that shear yielding offrame beams, such as that schematically illustrated in Figure 7.2.3-2 may be adesirable behavior mode. However, to date, there has not been enough researchconducted into the behavior of such frames to develop recommended designguidelines. Designers wishing to utilize such configurations should refer to thecode requirements for eccentrically braced frames. Particular care should betaken to brace the shear link of such beams against lateral-torsional buckling andalso to adequately stiffen the webs to avoid local buckling following shearplastification.

Shear Link

Shear Link

Figure 7.2.3-2 Shear Yielding Dominated Behavior of Short Bay Frames


The plastic rotation capacity of tested connection assemblies should reflect realistic estimatesof the total (elastic and plastic) drift likely to be induced in the frame by earthquake groundshaking, and the geometric configuration of the frame. For frames of typical configuration, andfor ground shaking of the levels anticipated by the building code, a minimum plastic rotationcapacity of 0.03 radian is recommended. As used in these Guidelines, plastic rotation is definedas the plastic chord rotation angle. The plastic chord rotation angle is calculated using the rotatedcoordinate system shown in Fig. 7.2.4-1 as the plastic deflection of the beam or girder, at its pointof inflection (usually the mid-span,) ∆CI, divided by the distance between this mid-span point andthe centerline of the panel zone of the beam column connection, LCL. This convention isillustrated in Figure 7.2.4-1.


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cL

cL

Beam span center line

∆CL

LCL

θ pCL

CLL= ∆

Plastic hinge

lh

Figure 7.2.4-1 Calculation of Plastic Rotation Angle

It is important to note that this definition of plastic rotation is somewhat different than theplastic rotation that would actually occur within a discrete plastic hinge in a frame model similarto that shown in Figure 7.2.3-1. These two quantities are related to each other, however, and ifone of them is known, the other may be calculated from Eq. 7.2.4-1.When the configuration of aframe is such that the ratio L/L’ is greater than 1.25, the plastic rotation demand should be takenas follows:

( )θ θp phCL h

CL

L lL

= −(7.2.4-1)

where: θp is the plastic chord angle rotation, as used in these Guidelines θph is the plastic rotation, at the location of a discrete hinge LCL is the distance from the center of the beam-column assembly panel zone to the center of the beam span lh is the location of the discrete plastic hinge relative to the center of the beam-column assembly panel zone

( )( )θ = + −0.025 1 L L' L' (7-1)

where: L is the center to center spacing of columns, and L’ is the center to center spacing of plastic hinges in the bay under consideration

The indicated rotation demands may be reduced when positive means, such as the use of baseisolation or energy dissipation devices, are introduced into the design to control the building’sresponse. When such measures are taken, nonlinear dynamic analyses should be performed andthe connection demands taken as 0.005 radians greater than the plastic rotation demandscalculated in the analyses. The nonlinear analyses should conform to the criteria specified inUBC-94 Section 1655 {NEHRP-94 Section 2.6.4.2} for nonlinear dynamic analysis of base


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isolated structures. Ground motion time histories utilized for these nonlinear analyses shouldsatisfy the scaling requirements of UBC-94 Section 1655.4.2 {NEHRP-94 Section 2.6.4.4},except that if the building is not base isolated, the structure period T, calculated in accordancewith UBC-94 Section 1628 {NEHRP-94 Section 2.3.3.1} should be substituted for TI. Whenusing methods of nonlinear analysis to establish the plastic rotation demands on frameconnections, the analysis results should not be scaled by the factor Rw (R) or RWi (Ri), asotherwise permitted by the building code.

Commentary: When the Interim Guidelines were first published, the plasticrotation was defined as that rotation that would occur at a discrete plastic hinge,similar to the definition of θph. in Eq. 7.2.4-1, above. In subsequent testing ofprototype connection assemblies, it was found that it is often very difficult todetermine the value of this rotation parameter from test data, since actual plastichinges do not occur at discrete points in the assembly and because some amountof plasticity also occurs in the panel zone of many assemblies. The plastic chordangle rotation, introduced in this advisory, may more readily be obtained fromtest data and also more closely relates to the drift experienced by a frame duringearthquake response.

This change in the definition of plastic rotation does not result in anysignificant change in the acceptance criteria for beam-column assemblyqualification testing. When the Interim Guidelines were first published, theyrecommended an acceptance criteria given by Eq. 7.2.4-2, below:

θ pL L

L= + −

0 025 1.'

'(7.2.4-2)

For typical beam-column assemblies in which the plastic hinge forms relativelyclose to the face of the column, perhaps within a length of 1/2 the beam depth,this typically resulted in a plastic rotation demand of 0.03 radians, as currentlymeasured.

Traditionally, engineers have calculated demand in moment frames by sizingthe members for strength and drift using code forces (either equivalent static orreduced dynamic forces) and then "developing the strength of the members." Since 1988, "developing the strength" has been accomplished by prescriptivemeans based on a review of testing of moment frame connections to that date. Itwas assumed that the prescribed connections would be strong enough that thebeam or girder would yield (in bending), or the panel zone would yield (in shear)in a nearly perfectly plastic manner producing the plastic rotations necessary todissipate the energy of the earthquake.


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A realistic estimate of the interstory drift demand for most structures and mostearthquakes is on the order of 0.015 to 0.025 times the story height for WSMFstructures designed to code allowable drift limits. In such frames, a portion ofthe drift will be due to elastic deformations of the frame, while the balance mustbe provided by inelastic rotations of the beam plastic hinges, by yielding of thecolumn panel zone, or by a combination of the two.

In the 1994 Northridge Earthquake, many moment-frame connectionsfractured with little evidence of plastic hinging of the beams or yielding of thecolumn panel zones. Testing of moment frame connections both prior to andsubsequent to the earthquake suggests that the standard, pre-Northridge, weldedflange-bolted web connection is unable to reliably provide plastic rotationsbeyond about 0.005 radian for all ranges of beam depths and often fails belowthat level. Since the elastic contribution to drift may approach 0.01 radian, thenecessary inelastic contributions will exceed the capability of the standardconnection in many cases. For frames designed for code forces and for the codedrift, the necessary plastic rotational demand may be expected to be on the orderof 0.02 radian or more and new connection configurations should be developed toaccommodate such rotation without brittle fracture.

The recommended plastic rotation connection demand of 0.03 radians wasselected both to provide a comfortable margin against the demands actuallyexpected in most cases and because in recent testing of connection assemblies,specimens capable of achieving this demand behaved in a ductile manner throughthe formation of plastic hinges.

For a given building design, and known earthquake hazard, it is possible tomore accurately estimate plastic rotation demands on frame connections. Thisrequires the use of nonlinear analysis techniques. Analysis software capable ofperforming such analyses is becoming more available and many design officeswill have the ability to perform such analyses and develop more accurateestimates of inelastic demands for specific building designs. However, whenperforming such analyses, care should be taken to evaluate building response formultiple earthquake time histories, representative of realistic ground motions forsites having similar geologic characteristics and proximity to faults as the actualbuilding site. Relatively minor differences in the ground motion time history usedas input in such an analysis can significantly alter the results. Since there issignificant uncertainty involved in any ground motion estimate, it isrecommended that analysis not be used to justify the design of structures withnon-ductile connections, unless positive measures such as the use of baseisolation or energy dissipation devices are taken to provide reliable behavior ofthe structure.


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It has been pointed out that it is not only the total plastic rotation demand thatis important to connection and frame performance, but also the connectionmechanism (for example - panel zone yielding, girder flange yielding/buckling,etc.) and hysteretic loading history. These are matters for further study in thecontinuing research on connection and joint performance.

7.2.5 Redundancy

The frame system should be designed and arranged to incorporate as many moment-resistingconnections as is reasonable into the moment frame.

Commentary: Early moment frame designs were highly redundant and nearlyevery column was designed to participate in the lateral-force-resisting system. Inan attempt to produce economical designs, recent practice often yieldedproduceddesigns which utilized only a few large columns and beams in a small proportionof the building’s frames for lateral resistance, with the balance of the buildingcolumns designed not considered or designed to participate in lateral resistance. This practice led to the need for large welds at the connections and to reliance ononly a few connections for the lateral stability of the building. The resultinglarge framing elements and connections are believed to have exacerbated thepoor performance of the pre-Northridge connection. Further, if only a fewframing elements are available to resist lateral demands, then failure of only afew connections has the potential to result in a significant loss of earthquake-resisting strength. Together, these effects are not beneficial to buildingperformance.

The importance of redundancy to building performance can not be over-emphasized. Even connections designed and constructed according to theimproved procedures recommended by these Interim Guidelines will have somepotential, albeit greatly reduced, for brittle failures. As the number of individualbeams and columns incorporated into the lateral-force-resisting system isincreased, the consequences of isolated connection failures are significantlyreduceds. Further, as more framing elements are activated in the building’sresponse to earthquake ground motion, the building develops greater potential forenergy absorption and dissipation, and greater ability to limit controlearthquake-induced deformations to acceptable levels.

Incorporation of more of the building framing into the lateral-force-resistingsystem will lead to smaller members and therefore an anticipated increase in the reliability of individual connections. It will almost certainly lead to improvedoverall system reliability. Further, recent studies conducted by designers indicatethat under some conditions, redundant framing systems can be constructed aseconomically as non-redundant systems. In these studies, the additional costsincurred in making a greater number of field-welded moment-resisting


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connections in the more redundant frame were balanced by a reduced totaltonnage of steel in the lateral-force-resisting systems, and sometimes reducedfoundation costs as well.

In order to codify the need for more redundant structural systems, the 1997Uniform Building Code has specifically adopted a reliability coefficient, ρx, tiedto the redundancy of framing present in the building. This coefficient, with valuesvarying from 1.0 for highly redundant structures to 1.5 for non-redundantstructures, is applied to the design earthquake forces, E, in the load combinationequations, and has the effect of requiring more conservative design force levelsfor structures with nonredundant systems. The Building Seismic Safety Council’sProvisions Update Committee has also approved a proposal to include such acoefficient in the1997 NEHRP Provisions also includes such a coefficient. Theformulation of this coefficient and its application are very similar in both the1997 Uniform Building Code and 1997 NEHRP Provisions.

As proposed contained in the 1997 NEHRP Provisions, the reliabilitycoefficient is given by the equation:

xArmax

202 −=ρ (7.25-1)

where:

r xmax = the ratio of the design story shear resisted by the single element

carrying the most shear force in the story to the total story shear, for agiven direction of loading. For moment frames, r xmax is taken as the

maximum of the sum of the shears in any two adjacent columns in amoment frame divided by the story shear. For columns common to twobays with moment resisting connections on opposite sides at the levelunder consideration, 70% of the shear in that column may be used in thecolumn shear summation.

Ax = the floor area in square feet of the diaphragm level immediately above thestory.

The 1997 UBC and NEHRP Provisions also require that structures utilizingmoment resisting frames as the primary lateral force resisting system beproportioned such that they qualify for a maximum value of ρx of 1.25. Structureslocated within a few kilometers of major active faults must be configured so as toqualify for a maximum value of ρx of 1.1.


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The most redundant moment-resisting frame systems are distributed frames inwhich all beam-column connections are detailed to be moment resisting. In thesetypes of structures, half of the moment-resisting connections will be to the minoraxis of the column which will typically result in weak column/strong beamframing. The 1994 UBC requirements limit the portion of the building designlateral forces that can be resisted by relative number of weak column/strong beamconnections in the moment frame system. This limitation was adopted to avoidthe design of frames likely to develop story mechanisms as opposed to concernabout the adequacy of moment-resisting connections to the minor axis ofcolumns. However, the limited research data available on such connectionssuggests that they do not behave well.

There is a divergence of opinion among structural engineers on thedesirability of frames in which all beam-column connections are made moment-resisting, including those of beams framing to the minor axis of columns. Use ofsuch systems as a means of satisfying the redundancy recommendations of theseInterim Guidelines requires careful consideration by the structural engineer. Limited testing in the past has indicated that moment connections made to theminor axis of wide flange columns are subject to the same types of fracturedamage experienced by major axis connections. As of this time, there has notbeen sufficient research to suggest methods of making reliable connections to thecolumn minor axis.

7.2.6 System Performance


7.2.7 Special Systems


7.3 Connection Design & Qualification Procedures - General

7.3.1 Connection Performance Intent

The intent of connection design should be to force the plastic hinge away from the face of thecolumn to a pre-determined location within the beam span. This may be accomplished by localreinforcement of the connection itself (cover plates, haunches, side plates, etc.) or by localreductions of the beam section (drilled holes, trimmed flanges, etc.). All elements of theconnection should have adequate strength to develop the forces resulting from the formation ofthe plastic hinge at the predetermined location, together with forces resulting from gravity loads. Section 7.5.2 outlines a design procedure for reinforced connection designs. Section 7.5.3provides a design procedure for reduced section connections.


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7.3.2 Qualification by Testing


7.3.3 Design by Calculation


7.4 Guidelines for Connection Qualification by Testing

7.4.1 Testing Protocol


7.4.2 Acceptance Criteria


a) The connection should develop beam plastic rotations as indicated in Section7.2.4, for at least one complete cycle.

b) The connection should develop a minimum strength equal to the plastic strength ofthe girder, calculated using minimum specified yield strength Fy, tThroughout theloading history required to achieve the required plastic rotation capacity, asindicated in a), above, the connection should develop a minimum moment at thecolumn face as follows:

i) For strengthened connections, the minimum moment at the column faceshould be equal to the plastic moment of the girder, calculated using theminimum specified yield strength, Fy of the girder. If the load limitingmechanism in the test is buckling of the girder flanges, the engineer, uponconsideration of the effect of strength degradation on the structure, mayconsider a minimum of 80% of the nominal strength as acceptable.

ii) For reduced section connection designs, the minimum moment at thecolumn face should be equal to the moment corresponding to developmentof the nominal plastic moment of the reduced section at the reducedsection, calculated using the minimum specified yield strength, Fy of thegirder, and the plastic section modulus for the reduced section. Themoment at the column face should not be less than 80% of the nominalplastic moment capacity of the unreduced girder section.


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c) The connection should exhibit ductile behavior throughout the loading history. Aspecimen that exhibits a brittle limit state (e.g. complete flange fracture, columncracking, through-thickness failures of the column flange, fractures in weldssubject to tension, shear tab cracking, etc. ) prior to reaching the required plasticrotation should be considered unsuccessful.

d) Throughout the loading history, until the required plastic rotation is achieved, theconnection should be judged capable of supporting dead and live loads required bythe building code. In those specimens where axial load is applied during thetesting, the specimen should be capable of supporting the applied load throughoutthe loading history.

The evaluation of the test specimen’s performance should consistently reflect the relevant limitstates. For example, the maximum reported moment and the moment at the maximum plasticrotation are unlikely to be the same. It would be inappropriate to evaluate the connection usingthe maximum moment and the maximum plastic rotation in a way that implies that they occurredsimultaneously. In a similar fashion, the maximum demand on the connection should beevaluated using the maximum moment, not the moment at the maximum plastic rotation unless thebehavior of the connection indicated that this limit state produced a more critical condition in theconnection.

Commentary: While the testing of all connection geometries and membercombinations in any given building might be desirable, it would not be verypractical nor necessary. Test specimens should replicate, within the limitationsassociated with test specimen simplification, the fabrication and weldingprocedures, connection geometry and member size, and potential modes offailure. If the testing is done in a manner consistent with other testing programs,reasonable comparisons can be made. On the other hand, testing is expensiveand it is difficult to realistically test the beam-column connection using actualboundary conditions and earthquake loading histories and rates.

It was suggested in Interim Recommendation No. 2 by the SEAOC SeismologyCommittee that three tested specimens be the minimum for qualification of aconnection. Further consideration has led to the recognition that while three testsmay be desirable, the actual testing program selected should consider theconditions of the project. Since the purpose of the testing program is to "qualifythe connection", and since it is not practical for a given project to do enough teststo be statistically meaningful considering random factors such as material,welder skills, and other variables, arguments can be made for fewer tests ofidentical specimens, and concentration on testing specimens which represent therange of different properties which may occur in the project. Once a connectionis qualified, that is, once it has been confirmed that the connection can work,monitoring of actual materials and quality control to assure emulation of thetested design becomes most important.


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Because of the cost of testing, use of calculations for interpolation orextrapolation of test results is desirable. How much extrapolation should beaccepted is a difficult decision. As additional testing is done, more informationmay be available on what constitutes "conservative" testing conditions, therebyallowing easier decisions relative to extrapolating tests to actual conditions whichare likely to be less demanding than the tests. For example, it is hypothesizedthat connections of shallower, thinner flanged members are likely to be morereliable than similar connections consisting of deeper, thicker flanged members. Thus, it may be possible to test the largest assemblages of similar details andextrapolate to the smaller member sizes - at least within comparable membergroup families. However, there is evidence to suggest that extrapolation of testresults to assemblies using members of reduced size is not always conservative. In a recent series of tests of cover plated connections, conducted at the Universityof California at San Diego, a connection assembly that produced acceptableresults for one family of beam sizes, W24, did not behave acceptably when thebeam depth was reduced significantly, to W18. In that project, the change inrelative flexibilities of the members and connection elements resulted in a shift inthe basic behavior of the assembly and initiation of a failure mode that was notobserved in the specimens with larger member sizes. In order to minimize thepossibility of such occurrences, when extrapolation of test results is performed, itshould be done with a basic understanding of the behavior of the assembly, andthe likely effects of changes to the assembly configuration on this behavior. Testresults should not be extrapolated to assembly configurations that are expected tobehave differently than the tested configuration. Extrapolation or interpolationof results with differences in welding procedures, details or material properties iseven more difficult.

7.5 Guidelines for Connection Design by Calculation

In conditions where it has been determined that design of connections by calculation issufficient, or when calculations are used for interpolation or extrapolation, the followingguidelines should be used.

7.5.1 Material Strength Properties

In the absence of project specific material property information, the values listed in Table 7-1Table 7.5.1-1 should be used to determine the strength of steel shape and plate for purposes ofcalculation. The permissible strength for weld metal should be taken in accordance with thebuilding code. Additional information on material properties may be found in the InterimGuidelines of Chapter 8.


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Table 7-1Table 7.5.1-1 - Properties for Use in Connection Design

Material Fy (ksi) Fy m (ksi) Fu (ksi)A36 36 use values for

Dual Certified58

Dual Certified Beam Axial, Flexural3

Shape Group 1 Shape Group 2 Shape Group 3 Shape Group 4 Through-Thickness

50

-

551

581

571

541

-

65 min.

Note 5A572 Column/Beam Axial, Flexural3

Shape Group 1 Shape Group 2 Shape Group 3 Shape Group 4 Shape Group 5 Through-Thickness

50

-

581

581

571

571

551

-

65 min.

, Note 5A9922 Use same values as ASTM A572A913-50 Axial, Flexural Through-thickness

50-

581

-65 min., Note 5

A913--— 65 Axial, Flexural Through-thickness

65 751(4) 80 min.Note 5

Notes:1. Based on coupons from web. For thick flanges,

the Fy flange is approximately 0.95 Fy web.2. See Commentary3. Values based on (SSPC-1994)4. ASTM A913, Grade 65 material is not recommended for use in the beams of moment resisting frames5. See Commentary

Commentary: Table7.5.1-1 Note 2 - The ASTM A992 specification wasspecifically developed by the steel industry in response to expressed concerns ofthe design community with regard to the permissible variation in chemistry andmechanical properties of structural steel under the A36 and A572 specifications. This new specification, which was adopted in late 1998, is very similar to ASTMA572, except that it includes somewhat more restrictive limits on chemistry andon the permissible variation in yield and ultimate tensile stress, as well as theratio of yield to tensile strength. At this time, no statistical data base is availableto estimate the actual distribution of properties of material produced to thisspecification. However, the properties are likely to be very similar, albeit withless statistical scatter, to those of material recently produced under ASTM A572,Grade 50.


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Table 7.5.1-1 Note 5 -In the period immediately following the Northridgeearthquake, the Seismology Committee of the Structural Engineers Association ofCalifornia and the International Conference of Building Officials issued InterimRecommendation No. 2 (SEAOC-1995) to provide guidance on the design ofmoment resisting steel frame connections. Interim Recommendation No. 2included a recommendation that the through-thickness stress demand on columnflanges be limited to a value of 40 ksi, applied to the projected area of beamflange attachment. This value was selected somewhat arbitrarily, to ensure thatthrough-thickness yielding did not initiate in the column flanges of moment-resisting connections and because it was consistent with the successful tests ofassemblies with cover plates conducted at the University of Texas at Austin(Engelhardt and Sabol - 1994), rather than being the result of a demonstratedthrough-thickness capacity of typical column flange material. Despite thesomewhat arbitrary nature of the selection of this value, its use often controls theoverall design of a connection assembly including the selection of cover platethickness, haunch depth, and similar parameters.

It would seem to be important to prevent the inelastic behavior of connectionsfrom being controlled by through-thickness yielding of the column flanges. Thisis because it would be necessary to develop very large local ductilities in thecolumn flange material in order to accommodate even modest plastic rotationdemands on the assembly. However, extensive investigation of the through-thickness behavior of column flanges in a “T” joint configuration reveals thatneither yielding, nor through-thickness failure are likely to occur in theseconnections. Barsom and Korvink (1997) conducted a statistical survey ofavailable data on the tensile strength of rolled shape material in the through-thickness direction. These tests were generally conducted on small diametercoupons, extracted from flange material of heavy shapes. The data indicates thatboth the yield stress and ultimate tensile strength of this material in the through-thickness direction is comparable to that of the material in the direction parallelto rolling. However, it does indicate somewhat greater scatter, with a number ofreported values where the through-thickness strength was higher, as well as lowerthan that in the longitudinal direction. Review of this data indicates with highconfidence that for small diameter coupons, the yield and ultimate tensile valuesof the material in a through-thickness direction will exceed 90% and 80%respectively of the comparable values in the longitudinal direction. the actualThe causes for through-thickness failures of column flanges (types C2, C4, andC5), observed both in buildings damaged by the Northridge Earthquake and insome test specimens, are not well understood. They are thought to be a functionof the metallurgy and “purity” of the steel; conditions of loading including thepresence of axial load and rate of loading application; conditions of tri-axialrestraint; conditions of local hardening and embrittlement within the weld’s heataffected zone; stress concentrations induced by the presence of backing bars and


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defects at the root of beam flange to column flange welds; and by the relationshipof the connection components as they may affect flange bending stresses andflange curvature induced by panel zone yielding. Given the many complex factorswhich can affect the through-thickness strength of the column flange,determination of a reliable basis upon which to set permissible design stresseswill require significant research. Such research is currently being conductedunder the SAC phase II program.

While this statistical distribution suggests the likelihood that the through-thickness strength of column flanges could be less than the flexural strength ofattached beam elements, testing of more than 40 specimens at Lehigh Universityindicates that this is not the case. In these tests, high strength plates,representing beam flanges and having a yield strength of 100 ksi were welded tothe face of A572, Grade 50 and A913, Grade 50 and 65 column shapes, tosimulate the portion of a beam-column assembly at the beam flange. Thesespecimens were placed in a universal testing machine and loaded to produce highthrough-thickness tensile stresses in the column flange material. The testssimulated a wide range of conditions, representing different weld metals as welland also to include eccentrically applied loading. In 40 of 41 specimens tested,the assembly strength was limited by tensile failure of the high strength beamflange plate as opposed to the column flange material. In the one failure thatoccurred within the column flange material, fracture initiated at the root of a low-toughness weld, at root defects that were intentionally introduced to initiate sucha fracture.

The behavior illustrated by this test series is consistent with mechanics ofmaterials theory. At the joint of the beam flange to column flange, the material isvery highly restrained. As a result of this, both the yield strength and ultimatetensile strength of the material in this region is significantly elevated. Underthese conditions, failure is unlikely to occur unless a large flaw is present that canlead to unstable crack propagation and brittle fracture. In light of this evidence,Interim Guidelines Advisory No. 2 deletes any requirement for evaluation ofthrough-thickness flange stress in columns.

Interim Recommendation No. 2 (SEAOC-1995) included a value of 40 ksi,applied to the projected area of beam flange attachment, for the through-thickness strength to be used in calculations. This value was selected because itwas consistent with the successful tests of assemblies with cover plates conductedat the University of Texas at Austin (Engelhardt and Sabol - 1994). However,because of the probable influence of all the factors noted above, this value canonly be considered to reflect the specific conditions of those tests and specimens.

Although reduced stresses at the column face produced acceptable results inthe University of Texas tests, the key to that success was more likely the result of


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forcing the plastic hinge away from the column than reduction of the through-thickness stress by the cover plates. Reduction of through-thickness columnflange stress to ever lower levels by the use of thicker cover plates is notrecommended, since such cover plates will result in ever higher forces on the faceof the column flange as well as larger weldments with potential for enlarged heataffected zones, higher residual stresses and conditions of restraint.

Since the initial publication of the Interim Guidelines, a significant number oftests have been performed on reduced beam section connections (See section7.5.3), most of which employed beam flanges which were welded directly to thecolumn flanges using improved welding techniques, but without reinforcementplates. No through-thickness failures occurred in these tests despite the fact thatcalculated through-thickness stresses at the root of the beam flange to columnflange joint ranged as high as 58 ksi. The successful performance of these weldedjoints is most probably due to the shifting of the yield area of the assembly awayfrom the column flange and into the beam span. Based on the indications of theabove described tests, and noting the undesirability of over reinforcingconnections, it is now suggested that a maximum through-thickness stress of0.9Fyc may be appropriate for use with connections that shift the hinging awayfrom the column face. Notwithstanding this recommendation, engineers are stillcautioned to carefully consider the through-thickness issue when these otherpreviously listed conditions which are thought to be involved in this type offailure are prevalent. Connections relying on through-thickness strength can notbe considered to be fully reliable until the influence of the other parametersdiscussed above can be fully understood. A high amount of structuralredundancy is recommended for frames employing connections which rely onthrough-thickness strength of the column flange.

Notwithstanding all of the above, successful tests using cover plates and othermeasures of moving hinges (and coincidentally reducing through-thickness stress)continue to be performed. In the interim, engineers choosing to utilizeconnections relying on through-thickness strength should recognize that despitethe successful testing, connections relying on through-thickness strength can notbe considered to be fully reliable until the influence of the other parametersdiscussed above can be fully understood. A high amount of structuralredundancy is recommended for frames employing connections which rely onthrough-thickness strength of the column flange.

7.5.2 Design Procedure - Strengthened Connections

The following procedure may be followed to size the various elements of strengthenedconnection assemblies that are intended to promote formation of plastic hinges within the beamspan by providing a reinforced beam section at the face of the column. Section 7.5.3 provides a


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modified procedure recommended for use in the design of connection assemblies using reducedbeam sections to promote similar inelastic behavior. Begin by selecting Select a connectionconfiguration, such as one of those indicated in Sections 7.9.1, 7.9.2, 7.9.3, 7.9.4, or 7.9.5, thatwill permit the formation of a plastic hinge within the beam span, away from the face of thecolumn, when the frame is subjected to gravity and lateral loads. Then proceed as described inthe following sections. The following procedure should be followed to size the various elementsof the connection assembly:

7.5.2.1 Determine Plastic Hinge Locations

For beams with gravity loads representing a small portion of the total flexural demand, thelocation of the plastic hinge may be assumed to occur as indicated in Table 7.5.2.1-1 at a distanceequal to 1/3 of the beam depth from the edge of the reinforced connection (or start of the reducedbeam section), unless specific test data for the connection indicates that a different location valueis more appropriate. Refer to Figure 7-2Figure 7.5.2.1-1.

Table 7.5.2.1-1 Plastic Hinge Location - Strengthened Connections

Connection Type Reference Section Hinge Location “sh”

Cover plates Sect. 7.9.1 d/4 beyond end of cover plates

Haunches Sect. 7.9.3, 7.9.4 d/3 beyond toe of haunch

Vertical Ribs Sect. 7.9.2 d/3 beyond toe of ribs

L

Bea

m d

epth

- d

Edge

of r

einf

orce

dco

nnec

tion

Edge

of r

einf

orce

dco

nnec

tion

sh=d/3

L’

Plastichinge

Connectionreinforcementsh=

d/4

Figure 7-2 Figure 7.5.2.1-1 - Location of Plastic Hinge

Commentary: The suggested locations for the plastic hinge, at a distance d/3away from the end of the reinforced section (or beginning of reduced section)indicated in Table 7.5.2.1-1 and Figure 7.5.2.1-1 are is based on the observedbehavior of test specimens, with no significant gravity load present. If significant


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gravity load is present, this can shift the locations of the plastic hinges, and in theextreme case, even change the form of the collapse mechanism. If flexuraldemand on the girder due to gravity load is less than about 30% of the girderplastic capacity, this effect can safely be neglected, and the plastic hingelocations taken as indicated. If gravity demands significantly exceed this level,then plastic analysis of the girder should be performed to determine theappropriate hinge locations. In zones of high seismicity (UBC Zones 3 and 4,and NEHRP Map Areas 6 and 7), gravity loading on the girders of earthquakeresisting frames typically has a very small effect, unless tributary areas forgravity loads are large.

7.5.2.2 Determine Probable Plastic Moment at Hinges

Determine the probable value of the plastic moment, Mpr, at the location of the plastic hingesas:

M M Z Fpr p b y= =β β (7.5.2.2-12)

where: ß is a coefficient that adjusts the nominal plastic moment to the estimated hingemoment based on the mean yield stress of the beam material and the estimatedstrain hardening. A value of 1.2 should be taken for β for ASTM A572, A992 andA913 steels. When designs are based upon calculations alone, an additional factoris recommended to account for uncertainty. In the absence of adequate testing ofthe type described above, ß should be taken as 1.4 for ASTM A572 and for A913,Grades 50 and 65 steels. Where adequate testing has been performed ß should bepermitted to be taken as 1.2 for these materials.


Commentary: In order to compute β, the expected yield strength, strainhardening and an appropriate uncertainty factor need to be determined. Thefollowing assumed strengths are recommended:

Expected Yield: The expected yield strength, for purposes of computing (Mpr) may be taken as:

Fye = 0.95 Fym (7.5.2.2-2-3)

The 0.95 factor is used to adjust the yield stress in the beam web, wherecoupons for mill certification tests are normally extracted, to the value in thebeam flange. Beam flanges, being comprised of thicker material, typically havesomewhat lower yield strengths than do beam web material.


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Fy m for various steels are as shown in Table 7-1 Table 7.5.1-1, based on asurvey of web coupon tensile tests (Steel Shape Producers Council - 1994). Theengineer is cautioned that there is no upper limit on the yield point for ASTM A36steel and consequently, dual-certification steel having properties consistent withASTM A572, Grade 50 is routinely supplied when ASTM A36 is specified. Consequently, it is the recommendation here that the design of connections bebased on an assumption of Grade 50 properties, even when A36 steel is specifiedfor beams. It should be noted that at least one producer offers A36 steel with amaximum yield point of 50 ksi in shape sizes ranging up to W 24x62. Refer to thecommentary to Section 8.1.3 for further discussion of steel strength issues.

Strain Hardening: A factor of 1.1 is recommended for use with the mean yieldstress in the foregoing table when calculating the probable plastic momentcapacity Mpr.. The 1.1 factor for strain hardening, or other sources of strengthabove yield, agrees fairly well with available test results. The 1.1 factor couldunderestimate the over-strength where significant flange buckling does not act asa gradual limit on the beam strength. Nevertheless, the 1.1 factor seems areasonable expectation of over-strength considering the complexities involved.

Modeling Uncertainty: Where a design is based on approved cyclic testing, themodeling uncertainty may be taken as 1.0, otherwise the recommended value is1.2. When the Interim Guidelines were first published, the β coefficient includeda 1.2 factor to account for modeling uncertainty when connection designs werebased on calculations as opposed to a specific program of qualification testing. The intent of this factor was twofold: to provide additional conservatism in thedesign when specific test data for a representative connection was not availableand also as an inducement to encourage projects to undertake connectionqualification testing programs. After the Interim Guidelines had been in use forsome time, it became apparent that this approach was not an effective inducementfor projects to perform qualification testing, and also that the use of an overlylarge value for the β coefficient often resulted in excessively large connectionreinforcing elements (cover plates, e.g.) and other design features that did notappear conducive to good connection behavior. Consequently, it was decided toremove this modeling uncertainty factor from the calculation of β.

In summary, for Grade 50 steel, we have:

β = [0.95 (54 ksi to 58 ksi)/50 ksi] (1.1) 1.2) = 1.35 t0 1.45, say 1.4

β = [0.95 (54 ksi to 59 ksi)/50 ksi] (1.1) = 1.13 to 1.21, say 1.2


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7.5.2.3 Determine Shear at the Plastic Hinge

The shear at the plastic hinge should be determined by statics, considering gravity loads actingon the beam. A free body diagram of that portion of the beam between plastic hinges, is a usefultool for obtaining the shear at each plastic hinge. Figure 7-3 Figure 7.5.2.3-1 provides anexample of such a calculation. For the purposes of such calculations, gravity load should be basedon the load combinations required by the building code in use.

L

sh

L’

Plastichinge P

L/2

P

Mpr MprL’Vp


“A”

VAw

Note: Gravity loads can effect the location of the plastic hinges. If 2Mpr /L’is less then the gravity shear in the free body (in this case P/2 + wL’/2), then the plastichinge location will shiftsignificantly and L’ must beadjusted, accordingly

w

Figure 7-3 Figure 7.5.2.3-1- Sample Calculation of Shear at Plastic Hinge

Commentary: The UBC gives no specific guidance on the load combinations touse with strength level calculations while the NEHRP Recommended Provisionsdo specify load factors for the various dead, live and earthquake components ofload. For designs performed in accordance with the UBC, it is customary to useunfactored gravity loads when checking the strength of elements.

7.5.2.4 Determine Strength Demands at Each Critical Section

In order to complete the design of the connection, including sizing the various plates andjoining welds which make up the connection, it is necessary to determine the shear and flexuralstrength demands at each critical section. These demands may be calculated by taking a free bodyof that portion of the connection assembly located between the critical section and the plastichinge. Figure 7-4 Figure 7.5.2.4-1 demonstrates this procedure for two critical sections, for thebeam shown in Figure 7-3 Figure 7.5.2.3-1.


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Plastichinge

Vp

Mpr

Plastichinge

Vp

Mpr

x

Mf

x+dc/2

dc

Mf=Mpr +Vpx

Mc

Mc=Mpr +Vp(x+dc/2)


Figure 7-4 Figure 7.5.2.4-1 - Calculation of Demands at Critical Sections

Commentary: Each unique connection configuration may have different criticalsections. The vertical plane that passes through the joint between the beamflanges and column (if such joining occurs) will typically define at least one suchcritical section, used for designing the joint of the beam flanges to the column, aswell as evaluating shear demands on the column panel zone. A second criticalsection occurs at the center line of the column. Moments calculated at this pointare used to check strong column - weak beam conditions. Other critical sectionsshould be selected as appropriate.

7.5.2.5 Check for Strong Column - Weak Beam Condition

When required by the building code, the connection assembly should be checked to determineif strong column - weak beam conditions are satisfied. In lieu of UBC-94 equation 11-3.1{NEHRP-91 equation 10-3}, the following equation should be used:

Z (F f ) M 1.0c yc a c− >∑ ∑ (7.5.2.5-1-4)

where: Zc is the plastic modulus of the column section above and below the connectionFyc is the minimum specified yield stress for the column above and belowfa is the axial load in the column above and belowΣMc is the moment calculated at the center of the column in accordance with

Section 7.5.2.4 sum of the column moments at the top and bottom of the panel zone, respectively, resulting from the development of the probable beam plastic moments, Mpr, within each beam in the connection.


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Commentary: The building code provisions for evaluating strong column - weakbeam conditions presume that the flexural stiffness of the columns above andbelow the beam are approximately equal, that the beams will yield at the face ofthe column, and that the depth of the columns and beams are small relative totheir respective span lengths. This permits the code to use a relatively simpleequation to evaluate strong column - weak beam conditions in which the sum ofthe flexural capacities of columns at a connection are compared against the sumsof the flexural capacities in the beams. The first publication of the InterimGuidelines took this same approach, except that the definition of ΣMc wasmodified to explicitly recognize that because flexural hinging of the beams wouldoccur at a location removed from the face of the column, the moments deliveredby the beams to the connection would be larger than the plastic moment strengthof the beam. In this equation, ΣMc was taken as the sum of the moments at thecenter of the column, calculated in accordance with the procedures of Sect.7.5.2.4.

(L-L’)/2

d ph t

h b

Mpr

Vp

Vp

Mpr

Vc

Vc+Vf

Mct

Mcb

assumed point of zero moment

Note:The quantities Mpr, Vp, L, and L’ areas previously identified. Vf is the incremental shear distributedto the column at the floor level.Other quantities are as shown.

Vf

( )[ ] ( )

( )

VM V L L V h d

h d h

M V h

M V V h

M M M

cpr p f b p

b p t

ct c t

cb c f b

c ct cb

=+ − − +

+ +== +

= +

∑

∑

' ) / /2 2

Figure 7.5.2.5-1 Calculation of Column Moment for Strong ColumnEvaluation

This simplified approach is not always appropriate. If non-symmetricalconnection configurations are used, such as a haunch on only the bottom side ofthe beam, this can result in an uneven distribution of stiffness between the twocolumn segments, and premature yielding of the column, either above, or below,the beam-column connection. Also, it was determined that for connectionconfigurations in which the panel zone depth represents a significant fraction of


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the total column height, such as can occur in some haunched and side-platedconnections, the definition of ΣMc contained in the initial printing of theGuidelines could lead to excessive conservatism in determining whether or not astrong column - weak beam condition exists in a structure. Consequently, InterimGuidelines Advisory No. 1 adopted the current definition of ΣMc for use in thisevaluation. This definition requires that the moments in the column, at the topand bottom of the panel zone be determined for the condition when a plastichinge has formed at all beams in the connection. Figure 7.5.2.5-1 illustrates amethod for estimating this quantity.

7.5.2.6 Check Column Panel Zone

The adequacy of the shear strength of the column panel zone should be checked. For thispurpose, the term 0.8ΣMf should be substituted for the term 0.8ΣMs in UBC-94 Section2211.7.2.1 {0.9ΣφbMp in NEHRP-91 Section 10.10.3.1}, repeated below for convenience ofreference. Mf is the calculated moment at the face of the column, when the beam mechanismforms, calculated as indicated in Section 7.5.2.4 above. In addition, it is recommended that thealternative design criteria indicated in UBC-94 Section 2211.7.2.1 (NEHRP-91 Sect. 10.10.3.1),permitting panel zone shear strength to be proportioned for the shear induced by bendingmoments from gravity loads plus 1.85 times the prescribed seismic forces, not be used. Forconvenience of reference, UBC-94 Section 2211.7.2.1 is reproduced below, edited, to indicate therecommended application.

2211.7.2.1 Strength (edited). The panel zone of the joint shall be capable of resisting theshear induced by beam bending moments due to gravity loads plus 1.85 times theprescribed seismic forces, but the shear strength need not exceed that required to develop0.8ΣMs 0.8ΣMf of the girders framing into the column flanges at the joint. The joint panelzone shear strength may be obtained from the following formula:

V 0.55F d t3b td d ty c

c c f2

b c

= +

1 (11-1)

where: bc = width of column flangedb = the depth of the beam (including any haunches or cover plates)dc = the depth of the columnt = the total thickness of the panel zone including doubler platestcf = the thickness of the column flange

Commentary: The effect of panel zone shear yielding on connection behavior isnot well understood. In the past, panel zone shear yielding has been viewed as abenign, or even beneficial mechanism that permits overall frame ductilitydemands to be accommodated while minimizing the extent of inelastic behaviorrequired of the beam and beam flange to column flange joint. The criteria


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permitting panel zone shear strength to be proportioned for the shears resultingfrom moments due to gravity loads plus 1.85 times the design seismic forces wasadopted by the code specifically to permit designs with somewhat weak panelzones. However, during recent testing of large scale connection assemblies withweak panel zones, it has been noted that in order to accommodate the large sheardeformations that occur in the panel zone, extreme “kinking” deformations wereinduced into the column flanges at the beam flange to column flange welded joint. While this did not lead to premature joint failure in all cases, it is believed tohave contributed to such premature failures in at least some of the specimens. The recommendations of this section are intended to result in stronger panelzones than previously permitted by the code, thereby avoiding potential failuresdue to this kinking action on the column flanges.

7.5.3 Design Procedure - Reduced Beam Section Connections

The following procedure may be followed to size the various elements of reduced beamsection (RBS) assemblies with circular curved reductions in beam flanges, such as shown inFigure 7.5.3-1., such as those indicated in Section 7.9.6 indicates other configurations for suchconnections, however, the circular curved configuration shown in Figure 7.5.3-1 is currentlypreferred. RBS assemblies are intended to promote the formation of plastic hinges within thebeam span by developing a segment of the beam with locally reduced section properties andstrength. Begin by selecting an RBS configuration, such as one of those indicated in Figure 7.5.3-1, that will permit the formation of a plastic hinge within the reduced section of the beam. Of theconfigurations shown in the figure, the circular curved configuration is preferred.

c l

a

a

R = radius of cut =4a + l8a

2 2

bf

Figure 7.5.3-1 Geometry of Reduced Beam Section

Commentary: Connection assemblies in which inelastic behavior is shifted awayfrom the column face through development of a segment of the beam withintentionally reduced properties, so-called reduced beam section (RBS) or“dogbone” connections, appear to have the potential to provide an economical


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solution to the WMSF connection problem. These recommendations are based onlimited design configurations that have successfully been tested ing that has beenconducted of these types of connections to date. While a A large number of RBStests have been conducted, these tests have not included the effects of floor slabsor loading rates approximating those that would be produced by a building’sresponse to earthquake ground motionsincluding some tests of assemblies withfloor slabs present. Extensive additional testing of RBS connections, intended toexplore these and other factors relevant to connection performance, are currentlyplanned under funding provided by NIST and the SAC phase II program. In theinterim, designers specifying RBS connections may wish to consider provision ofdetails to minimize the participation of the slab in the flexural behavior of thebeam at the reduced section. The criteria presented in this section are partiallybased on a draft procedure developed by AISC (Iwankiw, 1996).

ReducedSection Drilled Constant Drilled Tapered

Circular

Straight Tapered

Figure 7.5.3-2 Alternative Reduced-Beam Section PatternsFigure 7.5.3-1 Reduced Beam Section Patterns

Several alternative configurations of RBS connections have also been testedto date. As indicated in Figures 7.5.3-21 and 7.9.6-1, these include constantsection, tapered section, curved section, and drilled hole patterns. It appears thatseveral of these configurations are more desirable than others. In particular, thedrilled hole section patterns have been subject to tensile failure across thereduced net section of the flange through the drill holes. A few RBS tests utilizingstraight or tapered cuts have failed within the reduced section at plastic rotationdemands less than recommended by these Guidelines. In all of these cases, thefailure occurred at locations at which there was a change in direction of the cutsin the beam flange, resulting in a geometric stress riser or notch effect. It is alsoreported that one of these tests failed at the beam flange continuity plate - tocolumn flange joint. There have been no reported failures of RBS connection


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assemblies employing the circular curved flange cuts, and therefore, this is thepattern recommended in these Guidelines. This would appear, therefore, to be amore desirable configuration, although some successful tests have beenperformed using the straight and tapered configurations.

It is important that the pattern of any cuts made in the flange be proportionedso as to avoid sharp cut corners. All corners should be rounded to minimizenotch effects and in addition, cut edges should be cut or ground in the directionof the flange length to have a surface roughness meeting the requirements of AWSC4.1-77 class 4, or smootherroughness value less than or equal to 1,000, asdefined in ANSI/ASME B46.1.

Concerns have been raised by some engineers over the strength reductioninherent in the RBS. Clearly, code requirements for strength, considering gravityloads and gravity loads in combination with wind, seismic and other loads mustbe met. For higher seismic zones, beam sizes are typically governed by elasticstiffness considerations (drift control) and this must be addressed. Also, forseismic loads, the Building Codes typically require that connections for SpecialMoment Resisting Frames must develop the “strength” or the “plastic bendingmoment” of the beam. There may be a problem of semantics where theserequirements are applied to a system using RBS connections. Is the RBS part ofthe connection or is it part of the beam, the strength of which must be developedby the connection? Clearly, the latter interpretation should be applied.

Notwithstanding the above, it must be kept in mind that, although unstated,and typically not quantified, there is inherent in design practice an impliedrelationship between the elastic behavior that we analyze and the inelasticbehavior which the building is expected to experience. Elastic drift limitationscommonly used are considered to be related to the anticipated inelastic drifts andultimate lateral stability of the framed structure in at least an intuitivelypredictable manner. It can be shown that RBS’s such as those that have beentested will reduce the elastic stiffness (increase the drift) on the order of 5%.However, because of the reduction in strength, the effect on the inelastic drift maybe more significant. Thus, it seems prudent to require that the RBS maintain areasonably high proportion of the frame inelastic strength. For the connectionstested to date, the inelastic strength of the RBS section has been in the range of70% of that of the full section. However, the moment demand at the face of thecolumn, corresponding to development of this reduced section strength, is likelyto be in the range of 85% to 90% of the strength of the full beam. This seems tobe quite reasonably high considering the accuracy of other seismic designassumptions.

Although the use of RBS designs tends to reduce the total strength demand onthe beam flange - to - column flange connection, relative to strengthened


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connections, designs utilizing RBS configurations should continue to follow therecommendations for beam flange continuity plates, weld metal and base metalnotch toughness recommended by the Interim Guidelines for strengthenedconnections.

7.5.3.1 Determine Reduced Section and Plastic Hinge Locations

The reduced beam section should be located at a sufficient distance from the face of thecolumn flange (dimension “c” in Figures 7.5.3-1 and 7.5.3.1-1) to avoid significant inelasticbehavior of the material at the beam flange - to - column flange joint. Based on testing performedto date, it appears that a value of “c” on the order of ½ to ¾ of the beam width, bf, is sufficient.d/4 (where “d” is the beam depth) is sufficient. The total length of the reduced section of beamflange (dimension “l” in Figures 7.5.3-1 and 7.5.3.1-1) should be on the order of 0.65d to 0.85d,where d is the beam depth.3d/4 to d. The location of the plastic hinge, sh,, may be taken as ½ thelength of the cut-out, l.indicated in Table 7.5.3.1-1, unless test data indicates a more appropriatevalue should be used. When tapered configurations are utilized, the slope of the tapered cut in thebeam flange should be arranged such that the variation of the plastic section modulus, Zx, withinthe reduced section approximates the moment gradient in the beam during the condition whenplastic hinges have formed within the reduced beam sections at both ends.

L

Bea

m d

epth

- d

L’Plastichinge

reducedsection

c

l

sh

Figure 7.5.3.1-1 Critical Dimensions - RBS Assemblies

7.5.3.2 Determine Strength and Probable Plastic Moment in RBS

The RBS may be proportioned to meet the following criteria:


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1. The section at the RBS should be sufficient to satisfy the strength criteria specified bythe building code for Dead, Live, Seismic, Snow, Wind, and other applicable designforces.

2. The elastic stiffness of the frame, considering the effects of the RBS, should besufficient to meet the drift requirements specified by the code, under the design seismicand other forces.

3. The expected stress in the beam flange - to - column flange weld, under the applicationof gravity forces and that seismic force that results in development of the probableplastic moment of the reduced section at both ends of the beam, should be less than orequal to the strength of the weld, as indicated in Section 7.2.2 of the InterimGuidelines.

4. The expected through-thickness stress on the face of the column flange, calculated asMf/Sc, under the application of gravity forces and that seismic force that results indevelopment of the probable plastic moment of the reduced section at both ends of thebeam, should be less than or equal to the values indicated in Section 7.5.1, where Mf isthe moment at the face of the column flange, calculated as indicated in Section 7.5.2.4,and Sc is the elastic section modulus of the beam at the connection considering weldreinforcement, bolt holes, reinforcing plates, etc. The maximum moment at the face ofthe column should be in the range of 85 percent to 100 percent of the beam’s expectedplastic moment capacity. The depth of cut-out, a, should be selected to be less than orequal to bf/4.

The plastic section modulus of the RBS may be calculated from the equation:

( )Z Z b t d tRBS x R f f= − − (7.5.3.2-1)

where:ZRBS is the plastic section modulus of the reduced beam sectionZx is the plastic modulus of the unreduced sectionbR is the total width of material removed from the beam flangetf is the thickness of the beam flanged is the depth of the beam

The probable plastic moment, Mpr, at the RBS shall be calculated from the equation:

M Z Fpr RBS y= β (7.5.3.2-2)

where:ZRBS is the plastic section modulus of the reduced beam sectionβ is as defined in Section 7.5.2.2


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The strength demand on the beam flange - to - column flange weld and on the face of thecolumn may be determined by following the procedures of Section .7.5.2.3 and 7.5.2.4 of theInterim Guidelines, using the value of Mpr determined in accordance with Eq. 7.5.3.2-2.

Commentary: Initial design procedures for RBS connections published by SACrecommended that sufficient reduction of the beam flange be made to maintainflexural stresses in the beam, at the column face, below the anticipated through-thickness yield strength of the column flange material. Since the publication ofthose recommendations, extensive testing of RBS connections has been conducted,both with and without composite slabs. The testing conducted to date on RBSspecimens This testing has typically been for configurations that would result insomewhat larger strength demands at the face of the column flange thansuggested by the criteria originally published by SAC. contained in this Advisory. Typically, the tested specimens had reductions in the beam flange area on theorder of 35% to 45% and produced moments at the face of the column thatresulted in stresses on the weld and column as large as large as 90 to 100% of theexpected material strength of the beam, which is often somewhat in excess of thethrough-thickness yield strength of the column material. The specimens in thesetests all developed acceptable levels of inelastic deformation. Recent studiesconducted for SAC at Lehigh University confirm that the significant conditions ofrestraint that exist at the beam flange to column flange joint results insubstantially elevated column through-thickness strength, negating a need toreduce flexural stresses below the anticipated column yield strength. In view ofthis evidence, SAC has elected to adopt design recommendations consistent withconfigurations that were successfully tested. The criteria contained in thisAdvisory suggest that these demands be reduced to a level which would maintainweld stresses within their normally specified values and through-thickness columnflange stresses at the same levels recommended for strengthened connections. This may require the beam flanges to be reduced by as much as 50% or more forsome frame configurations, or that supplemental reinforcement such as coverplates or vertical ribs be provided in addition to the reduced section. Thisapproach was taken to maintain consistency with the criteria recommended forstrengthened connections and with the knowledge that the factors affecting theperformance of these connections are not yet fully understood.

7.5.3.3 Strong Column - Weak Beam Condition

The adequacy of the design to meet strong column - weak beam conditions should be checkedin accordance with the procedures of Section 7.5.2.5


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7.5.3.4 Column Panel Zone

The adequacy of the column panel zone should be checked in accordance with the proceduresof Section 7.5.2.6.

7.5.3.5 Lateral Bracing

The reduced section of the beam flanges should be provided with adequate lateral support toprevent lateral-torsional buckling of the section. Lateral braces should be located within adistance equal to 1/2 the beam depth from the expected location of plastic hinging, but should notbe located within the reduced section of the flanges.

Commentary: Unbraced compression flanges of beams are subject to lateral-torsional buckling, when subjected to large flexural stresses , such as occur in theplastic hinges of beams reduced sections of RBS connections during response tostrong ground motion. To prevent such behavior lateral-torsional buckling, it isrecommended that both flanges of beams be provided with lateral support.Section 9.8 of the 1997 AISC Seismic Specification requires such bracing ingeneral, and specifically states as follows:

“Both flanges of beams shall be laterally supported directly or indirectly. The unbraced length between lateral supports shall not exceed 2500ry/Fy. Inaddition, lateral supports shall be placed near concentrated forces, changesin cross section and other locations where analysis indicates that a plastichinge will form during inelastic deformations of the SMF.”

Adequate lateral support of the top flanges of beams supporting concretefilled metal deck or formed slabs can usually be obtained through the normalwelded attachments of the deck to the beam or through shear studs. Lateralsupport of beam flanges can also be provided through the connections oftransverse framing members or by provision of special lateral braces, attacheddirectly to the flanges. Such attachments should not be made within the reducedsection of the beam flange as the welding or bolting required to make suchattachments can lead to premature fracturing in these regions of high plasticdemands.

For beams in moment-resisting frames, it has traditionally been assumed thatthe direct attachment of the beam flanges to the columns provided sufficientlateral support of both beam flanges to accommodate the plastic hingesanticipated to develop in these frames at the beam-column connection. However,connection configurations like the RBS, developed following the Northridgeearthquake, are intended to promote formation of these plastic hinges at somedistance from the beam-column interface. This brings to question the adequacyof the beam flange to column flange attachments to provide the necessary lateral


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support at the plastic hinge. While this issue is pertinent for any connectionconfiguration that promotes plastic hinge formation remote from the beam-column interface, RBS connections could be more susceptible to lateral-torsionbuckling at the plastic hinge because the reductions in the beam flange used toachieve plastic hinge formation also locally reduce the torsional resistance of thesection. For that reason, FEMA-267a recommended provision of lateral bracingadjacent to the reduced beam section.

Provision of lateral bracing does result in some additional cost. Therefore,SAC has engaged in specific investigations to evaluate the effect of lateralbracing both on the hysteretic behavior of individual connections as well asoverall frame response to large lateral displacements. Until these investigationshave concluded SAC continues to recommend provision of lateral bracing forRBS connections. It should be noted that Section 9.8 of the 1997 AISC SeismicSpecification states:

“If members with Reduced Beam Sections, tested in accordance withAppendix S are used, the placement of lateral support for the member shall beconsistent with that used in the tests.”

Most testing of RBS specimens performed as part of the SAC project haveconsisted of single beams cantilevered off a column to simulate the exteriorconnection in a multi-bay moment-resisting frame. The beams have generallybeen braced at the end of the cantilever length, typically located about 100 inchesfrom the face of the column. For the ASTM A572, Grade 50, W36x150 sectionstypically tested, this results in a nominal length between lateral supports that iscomparable to 2500ry/Fy.

The appropriate design strength for lateral bracing of compression elementshas long been a matter of debate. Most engineers have applied “rules of thumb”that suggest that the bracing element should be able to resist a small portion,perhaps on the order of 2% to 6% of the compressive force in the element beingbraced, applied normal to the line of action of the compression. A recentsuccessful test of an RBS specimen conducted at the University of Texas at Austinincorporated lateral bracing with a strength equal to 6% of the nominalcompressive yield force in the reduced section.

7.5.3.6 Welded Attachments

Headed studs for composite floor construction should not be placed on the beam flangebetween the face of the column and the extreme end of the RBS, as indicated in Figure 7.5.3.6-1.Other welded attachments should also be excluded from these regions of the beam.


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Reduced beam section

welded attachment permitted

welded attachment prohibited

Figure 7.5.3.6-1 Welded Attachments to RBS Beams

Commentary: There are two basic reasons for omitting headed studs in theregion between the reduced beam section and the column. The first of these isthat composite action of the slab and beam can effectively counteract thereduction in beam section properties achieved by the cutouts in the top beamflange. By omitting shear studs in the end region of the beam, this compositebehavior is neutralized, protecting the effectiveness of the section reduction. Thesecond reason is that the portion of the beam at the reduced section is expected toexperience large cyclic inelastic strains. If welded attachments are made to thebeam in this region, the potential for low-cycle fatigue of the beam, under theselarge cyclic inelastic strains is greatly increased. For this same reason, otherwelded attachments should also be excluded from this region.

7.6 Metallurgy and Welding


7.7 Quality Control/Quality Assurance


7.8 Guidelines on Other Connection Design Issues



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7.8.1 Design of Panel Zones

No current recommendations are made to supplement or modify the UBC-1994 {NEHRP-91}provisions for the design of panel zones, other than as indicated in Section 7.5.2.6, above. Panelzone demands should be calculated in accordance with Section 7.5.2.6. As with other elements ofthe connection, available panel zone strength should be computed using minimum specified yieldstress for the material, except when the panel zone strength is used as a limit on the requiredconnection strength, in which case Fym should be used.

Where connection design for two-sided connection assemblies is relying on test data for one-sided connection assemblies, consideration should be given to maintaining the level of panel zonedeformation in the design to a level consistent with that of the test, or at least assume that thepanel zone must remain elastic, under the maximum expected shear demands.

Commentary: At present, no changes are recommended to the code requirementsgoverning the design of panel zones, other than in the calculation of the demand. As indicated in Section 7.5.2.6, it is recommended that the formulation for panelzone demand contained in the UBC, based on 1.85 times the prescribed seismicforces, not be utilized. This formulation, which is not contained in either theAISC Seismic Provisions or the NEHRP Provisions, is felt to lead to the design ofpanel zones that are excessively flexible and weak in shear. There is evidencethat panel zone yielding may contribute to the plastic rotation capability of aconnection. However, there is also concern and some evidence that if thedeformation is excessive, a kink will develop in the column flange at the joint withthe beam flange and, if the local curvature induced in the beam and columnflanges is significant, can contribute to failure of the joint. This would suggestthat greater conservatism in column panel zone design may be warranted.

In addition to the influence of the deformation of the panel zone on theconnection performance, it should be recognized that the use of doubler platesand especially the welding associated with them is likely to be detrimental to theconnection performance. It is recommended that the Engineer consider use ofcolumn sizes which will not require addition of doubler plates, where practical.

7.8.2 Design of Web Connections to Column Flanges

Specific modifications to the code requirements for design of shear connections are not madeat this time. It should be noted that the emergency code change to the UBC-94 {NEHRP-94}deleted the former requirements for supplemental web welds on shear connections. This is felt tobe appropriate since these welds can apparently contribute to the potential for shear tab failure atlarge induced rotations.

When designing shear connections for moment-resisting assemblies, the designer shouldcalculate shear demands on the web connection in accordance with Section 7.5.2.4, above. For


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connection designs based on tested configurations, the web connection design should beconsistent with the conditions in the tested assemblies.

Commentary: Some engineers consider that it is desirable to develop as muchbending strength in the web as possible. Additionally, it has been observed insome laboratory testing that pre-mature slip of the bolted web connection canresult in large secondary flexural stresses in the beam flanges and the weldedjoints to the column flange. However, there is some evidence to suggest that ifflange connections should fail, welding of shear tabs to the beam web maypromote tearing of the tab weld to the column flange or the tab itself through thebolt holes, and some have suggested that welding be avoided and that webconnections should incorporate horizontally slotted holes to limit the momentwhich can be developed in the shear tab, thereby protecting its ability to resistgravity loads on the beam in the event of flexural connection failure.

Some recent finite element studies of typical connections by Goel, Popov andothers have suggested that even when the shear tab is welded, shear demands atthe connections tend to be resisted by a diagonal tension type behavior in the webthat tends to result in much of the shear being resisted by the flanges. Investigation of these effects is continuing.

7.8.3 Design of Continuity Plates


7.8.4 Design of Weak Column and Weak Way Connections


7.9 Moment Frame Connections for Consideration in New Construction


7.9.1 Cover Plate Connections

Figure 7-5 Figure 7.9.1-1 illustrates the basic configuration of cover plated connections. Short cover plates are added to the top and bottom flanges of the beam with fillet welds adequate totransfer the cover plate forces to the beam flanges. The bottom flange cover plate is shop welded to thecolumn flange and the beam bottom flange is field welded to the column flange and to the cover plate. The top flange and the top flange cover plate are both field welded to the column flange with acommon weld. The web connection may be either welded or high strength (slip critical) bolted. Limited testing of these connections (Engelhardt & Sabol - 1994), (Tsai & Popov -1988) has beenperformed. More than 30 tests of such connections have been performed, with data on at least 18 ofthese tests available in the public domain.


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A variation of this concept which has been tested successfully very recently (Forell/ElsesserEngineers -1995), uses cover plates sized to take the full flange force, without direct welding of thebeam flanges themselves to the column. In this version of the detail, the cover plate provides a crosssectional area at the column face about 1.7 times that of the beam flange area. In the detail which hasbeen tested, a welded shear tab is used, and is designed to resist a significant portion of the plasticbending strength of the beam web.

T&B

Figure 7-5 Figure 7.9.1-1 - Cover Plate Connection

Design Issues: Following the Northridge earthquake, the University of Texas at Austinconducted a program of research, under private funding, to develop a modified connectionconfiguration for a specific project. Following a series of unsuccessful tests on various types ofconnections, approximatelyApproximately eight connections similar to that shown in Figure 7-5 Figure 7.9.1-1 were have been recently tested (Engelhardt & Sabol - 1994), and they havedemonstrated the ability to achieve acceptable levels of plastic rotation provided that the beamflange to column flange welding wasis correctly executed and through-thickness problems in thecolumn flange were are avoided. This configuration is relatively economical, compared to someother reinforced configurations, and has limited architectural impact. As a result of thesefactors, and the significant publicity that followed the first successful tests of these connections,cover plated connections quickly became the predominant configuration used in the design ofnew buildings. As a result, a number of qualification tests have now been performed on differentvariations of cover plated connections, covering a wide range of member sizes ranging fromW16 to W36 beams, as part of the design process for individual building projects. The results ofthese tests have been somewhat mixed, with a significant number of failures reported. Althoughthis connection type appears to be significantly more reliable than the typical pre-Northridgeconnection, it should be expected that some connections in buildings incorporating this detailmay still be subjected to earthquake initiated fracture damage. Designers should consider usingalternative connection types, unless highly redundant framing systems are employed.


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Six of eight connections tested by the University of Texas at Austin were able to achieveplastic rotations of at least 0.025 radians, or better. Strength loss at the extreme levels of plasticrotation did not reduce the flexural capacity to less than the plastic moment capacity of thesection based on minimum specified yield strength. One specimen achieved plastic rotations of0.015 radians when a brittle fracture of the CJP weld (type W2 failure) occurred. This maypartially be the result of a weld that was not executed in conformance with the specified weldingprocedure specification. The second unsuccessful test specimen achieved plastic rotations of0.005 radian when a section of the column flange (type C2 failure) occurred. A similar failureoccurred in recent testing by Popov of a specimen with cover plates having a somewhat modifiedplan shape.

Quantitative Results: No. of specimens tested: 18Girder Size: W21 x 68 to W36 x 150Column Size: W12 x 106 to W14 x 455Plastic Rotation achieved-

6 13 Specimens : >0.025 radian1 3 Specimens: 0.015 0.005 < θp < 0.025 radian1 2 Specimens: 0.005 radian

Although apparently more reliable than the former prescriptive connection, thisconfiguration is subject to some of the same flaws including dependence dependent on properlyexecuted beam flange to column flange welds, and through-thickness behavior of the columnflange. Further these effects are somewhat exacerbated as the added effective thickness of thebeam flange results in a much larger groove weld at the joint, and therefore potentially moresevere problems with brittle heat affected zones and lamellar defects in the column. Indeed, asignificant percentage of connections of this configuration have failed to produce the desiredamount of plastic rotation.

One of the issues that must be faced by designers utilizing cover plated connections is thesequence of operations used to attach the cover plate and beam flange to the column. In oneapproach, the bottom cover plate is shop welded to the column, and then used as the backing forthe weld of the beam bottom flange to the column flange. This approach has the advantage ofproviding an erection seat and also results in a somewhat reduced amount of field welding forthis joint. A second approach is to attach the cover plate to the beam flange, and then weld it tothe column, in the field, as an integral part of the beam flange. There are tradeoffs to bothapproaches. The latter approach results in a relatively large field weld at the bottom flange withlarge heat input required into the column and beam. If this operation is not performed withproper preheat and control of the heat input, it can potentially result in an enlarged and brittleheat affected zone in both members. The first approach results in reduced heat input andtherefore, somewhat minimized potential for this effect. However, proper control of preheat andheat input remains as important in either case, as improper procedures can still result in brittleconditions in the heat affected zone. Further, the detail in which the cover plate is shop weldedto the column can lead to a notch effect for the column flange at the seam between the beam


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flange and cover plate. This is effect is illustrated in Figure 7.9.1-2. At least one specimenemploying this detail developed a premature fracture across the column flange that has beenrelated to this notch effect. This effect has been confirmed by recent fracture mechanicsmodeling of this condition conducted by Deierlein.

When developing cover plated connection details, designers should attempt to minimize thetotal thickness of beam flange and cover plate, so as to reduce the size of the complete jointpenetration weld of these combined elements to the column flange. For some frameconfigurations and member sizes, this combined thickness and the resulting CJP weld size canapproach or even exceed the thickness of the column flange. While there is no specific criteriain the AWS or AISC specifications that would suggest such weldments should not be made,judgementally they would not appear to be desirable from either a constructability orperformance perspective. As a rough guideline, it is recommended that for connections in whichboth the beam flange and cover plate are welded to the column flange, the combined thickness ofthese elements should not exceed twice the thickness of the beam flange nor 100% of thethickness of the column flange. For cover plated connections in which only the cover plate iswelded to the column flange, the same thickness limits should be applied to the cover plate.

columnflange(in tension)

cover plate

beam bottom flange

seam acts as notch

Figure 7.9.1-2 Notch Effect at Cover Plated Connections

7.9.2 Flange Rib Connections



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7.9.3 Bottom Haunch Connections

Figure 7.9.3-1 7-7 indicates the configuration of a connection with a haunch at the bottombeam flange.several potential configurations for single, haunched beam-column connections. Aswith the cover plated and ribbed connections, the intent is to shift the plastic hinge away from thecolumn face and to reduce the demand on the CJP weld by increasing the depth of the section. To date, the configuration incorporating the triangular haunch has been subjected to limitedtesting. Testing of configurations incorporating the straight haunch are currently planned, buthave not yet been performed. Several tests of this connection type were conducted by Uang underthe SAC phase I project (Uang, 1995). Following that work, additional research on the feasibilityof improving connection performance with welded haunches was conducted under a project thatwas jointly sponsored by NIST and AISC (NIST, 1998). That project was primarily focused onthe problem of upgrading connections in existing buildings. As indicated in the report of thatwork, the haunched modification improves connection performance by altering the basic behaviorof the connection. In essence, the haunch creates a prop type support, beneath the beam bottomflange. This both reduces the effective flexural stresses in the beam at the face of the support, andalso greatly reduces the shear that must be transmitted to the column through the beam. Acomplete procedure for the design of this modification may be found in NIST, 1998.


Figure 7.9.3-1 Bottom Haunch Connecction

Two Nine tests are known to have been performed to date, both successfully all intended toreplicate the condition of an existing connection that has been upgraded. Except for thosespecimens in which existing vulnerable welded joints were left in place at the top flange, theseconnections generally achieved large plastic rotations. Several dynamic tests have also been


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successfully conducted, although only moderate plastic deformation demands could be imposeddue to limitations of the laboratory equipment. Both tests were conducted in arepair/modification configuration. In one test, a portion of the girder top flange, adjacent to thecolumn, was replaced with a thicker plate. In addition, the bottom flange and haunch were bothwelded to the column. This specimen developed a plastic hinge within the beam span, outside thehaunched area and behaved acceptably. A second specimen did not have a thickened top flangeand the bottom girder flange was not welded to the column. Plastic behavior in this specimenoccurred outside the haunch at the bottom flange and adjacent to the column face at the topflange. Failure initiated in the girder at the juncture between the top flange and web, possiblycontributed to by buckling of the flange as well as lateral torsional buckling of the section. Fracture progressed slowly along the top fillet of the girder and eventually, traveled into theflange itself.

Design Issues: The haunch can be attached to the girder in the shop, reducing field erectioncosts. Weld sizes are smaller than in cover plated connections. The top flange is free ofobstructions.


Specimen 1 UCSD-1R:0.04 radian (w/o bottom flange weld and reinforced top flange)

Specimen 2 UCSD-3R:0.05 radian (with bottom flange weld and reinforced top flange)

Specimen UCSD-4R: 0.014 radian (dynamic- limited by test setup)Specimen UCSD-5R: 0.015 radian (dynamic- limited by test setup)

Girder Size: W36x150Column Size: W14x257Plastic Rotation achieved -

Specimen UCB-RN2: 0.014 radian (no modification of top weld)Specimen UTA-1R: 0.019 radian (partial modification of top weld)Specimen UTA-1RB: 0.028 radian (modified top weld)


Specimen UTA-NSF4: 0.015 radian (no modification of top weld)


Specimen SFCCC-8: 0.035 radian (cover plated top flange)


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Performance is dependent on properly executed complete joint penetration welds at thecolumn face. The joint can be subject to through-thickness flaws in the column flange; however,this connection may not be as sensitive to this potential problem because of the significantincrease in the effective depth of the beam section which can be achieved. Welding of the bottomhaunch requires overhead welding when relatively shallow haunches are used. The skewedgroove welds of the haunch flanges to the girder and column flanges may be difficult to execute. The increased depth of the beam, resulting from the haunch may have undesirable impact onarchitectural design. Unless the top flange is prevented from buckling at the face of the column,performance may not be adequate. For configurations incorporating straight haunches, thehaunch must be long, in order to adequately develop stress into the haunch, through the web. This tends to increase demands at the column face. Additional testing of all these configurationsis recommended.

7.9.4 Top and Bottom Haunch Connections


7.9.5 Side-Plate Connections


7.9.6 Reduced Beam Section Connections

In this connection, the cross section of the beam is intentionally reduced within a segment, toproduce an intended plastic hinge zone or fuse, located within the beam span, away from thecolumn face. Several ways of performing this cross section reduction have been proposed. Onemethod includes removal of a portion of the flanges, symmetrical about the beam centerline, in aso-called “dog bone” profile. Care should be taken with this approach to provide for smoothlycontoured transitions to avoid the creation of stress risers which could initiate fracture. It has alsobeen proposed to create the reduced section of beam by drilling a series of holes in the beamflanges. Figure 7-11 Figure 7.9.6-1 illustrates both concepts. The most successful configurationshave used reduced sections formed with circular cuts. Configurations which taper the reducedsection, through the use of unsymmetrical cut-outs, or variable size holes, to balance the crosssection and the flexural demand have also been tested with success.

Testing of this concept was first performed by a private party, and US patents were appliedfor and granted. These patents have now been released. Limited testing of both “dog-bone” anddrilled hole configurations have been performed in Taiwan (Chen and Yeh - 1995). The AmericanInstitute of Steel Construction is currently performing additional tests of this configuration(Smith-Emery - 1995), however the full results of this testing are not yet available. has performedsuccessful testing of 4 linearly tapered RBS connections. In the time since the first publication ofthe Interim Guidelines, a number of tests have been successfully conducted of RBS connectionswith circular curved cut-outs, including investigations and at the University of Texas at Austin,has successfully tested 4 circular curved RBS specimens. Others, including Popov at the


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University of California at Berkeley, and Texas A&M University., have also tested circular curvedRBS connections with success.

When this connection type was first proposed, There is a concern was expressed that thepresence of a concrete slab at the beam top flange would tend to limit the effectiveness of thereduced section of that flange, particularly when loading places the top flange into compression. It may be possible to mitigate this effect with proper detailing of the slab. Limited testing of RBSspecimens with composite slabs has recently been successfully conducted at Ecole Polytechnic, inMontreal, Canada. In these tests, shear studs were omitted from the portion of the top flangehaving a reduced section, in order to minimize the influence of the slab on flexural hinging. Inaddition, a 1 inch wide gap was placed in the slab, around the column, to reduce the influence ofthe slab on the connection at the column face. More recently, both the University of Texas atAustin and Texas A&M University have conducted successful tests of RBS connections withslabs and without such gaps present between the slab and column. This most recent testingsuggests that the presence of the slab actually enhances connection behavior by retarding bucklingof the top flange in compression and delaying strength degradation effects commonly observed inspecimens tested without slabs.

Design Issues: This connection type is potentially the most economical of the several types whichhave been suggested. The reliability of this connection type is dependent on the quality of thecomplete joint penetration weld of the beam to column flange, and the through-thicknessbehavior of the column flange. If the slab is not appropriately detailed, it may inhibit theintended “fuse” behavior of the reduced section beam segment. It is not clear at this timewhether it would be necessary to use larger beams with this detail to attain the same overallsystem strength and stiffness obtained with other configurations. In limited testing conducted todate of the unsymmetrical “dog-bone” configuration (Smith-Emery - 1995), the plastic hingingwhich occurred at the reduced section was less prone to buckling of the flanges than in some ofthe other configurations which have been tested, due to the very compact nature of the flange inthe region of the plastic hinge. However, the tendency for lateral-torsional buckling issignificantly increased suggesting the need for lateral bracing of the beam flanges, near thereduced section.

Experimental Results: A number of researchers have performed tests on RBS specimens to date.Most tests have utilized the ATC-24 loading protocol, which is similar to the protocol describedin Section 7.4.1 of the Interim Guidelines. Testing employed at Ecole Polytechnic, in Montreal,Canada utilized a series of different testing protocols including the ATC-24 procedure and adynamic excitation simulating the response of a connection in a building to an actual earthquakeaccelerogram (Tremblay, et. al., 1997). This research included two tests of connections withcomposite floor slabs. All of the reported tests with circular flange cuts have performedacceptably, however, the dynamic tests at Ecole Polytechnic only imposed 0.025 radians ofplastic rotation on the assembly.


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ReducedSection

Straight Tapered

Circular

Drilled Constant Drilled Tapered

Figure 7-11 7.9.6-1 - Reduced Beam Section Connection

Quantitative Results: No. of specimens tested: 219 published (without slabs)2Girder Size: W21 x 62W30 x 99 thru W 36 x 194Column Size: W14x120W14 x 176 thru W 14 x 426, W24 x 229Plastic Rotation achieved:- 0.03 radian

Straight: - 0.02 radianTapered - 0.027 - 0.045 radianCircular - 0.03 - 0.04 radian

No. of specimens tested: 42published (with slabs)Girder Size: W21 x 44 to W36 x 150Column Size: W14 x 90 to W14x257Plastic Rotation achieved: 0.03-0.05 radians (ATC-24 loading protocol)

0.025 radians (earthquake simulation – limitedby laboratory setup, no failure observed)

7.9.7 Slip - Friction Energy Dissipating Connection


7.9.8 Column-Tree Connection


7.9.9 Proprietary Slotted Web Connections

In the former prescriptive connection, in which the beam flanges were welded directly to thecolumn flanges, beam flexural stress was transferred into the column web through the combined


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action of direct tension across the column flange, opposite the column web, and through flexureof the column flange. This stress transfer mechanism and its resulting beam flange prying momentresults in a large stress concentration at the center of the beam flange, opposite the column web. Recent research (Allen, et. al. - 1995) indicates that the provision of continuity plates within thecolumn panel zone reduces this stress concentration somewhat, but not completely. The intent ofthe proprietary slotted web connections is to further reduce this stress concentration and toachieve a uniform distribution of flexural stress across the beam flange at the connection, and also,to promote local buckling of the beam flanges under compressive loads to limit the amount ofdemand on the beam flange to column flange weld. Claimed assets for this connection includeelimination of the vertical beam shear in the beam flange welds, elimination of the beam lateraltorsional buckling mode, and the participation of the beam web in resisting its portion of the beam moment. A number of different configurations for this connection type have been developed andtested. Figure 7.9.9-17-14 indicates one such configuration for this connection type that has beensuccessfully tested and which has been used in both new and retrofit steel moment-resistingframes. In this configuration, slots are cut into the beam web, extending from the weld accesshole adjacent to the top and bottom flanges, and extending along the beam axis a sufficient lengthto alleviate the stress concentration effects at the beam flange to column flange weld. The beamweb is welded to the column flange. vertical plates are placed between the column flanges,opposite the edges of the top and bottom beam flanges to stiffen the outstanding column flangesand draw flexural stress away from the center of the beam flange. Horizontal plates are placedbetween these vertical plates and the column web to transfer shear stresses to the panel zone. Theweb itself is softened with the cutting of a vertical slot in the column web, opposite the beamflange. High fidelity finite element models were utilized to confirm that a nearly uniformdistribution of stress occurs across the beam flange.

Slot, typ.

NOTICE OF CONFIDENTIAL INFORMATION:WARNING: The information presented in this figure is PROPRIETARY. US patents have been grantedand Foreign Patents have been applied for. Use of this information is strictly prohibited except asauthorized in writing by the developer. Violators shall be prosecuted in accordance with US and ForeignPatent Intellectual Property Laws.

Figure 7.9.9-17-14 - Proprietary Slotted Web Connection


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Design Issues: This detail is potentially quite economical, entailing somewhat more shopfabrication than the former prescriptive connection, but similar levels of field erection work. Contrary to the recommendations contained in these Interim Guidelines, this connection doesnot shift the location of plastic hinging away from the column face. However, two a number ofconnections employing details similar to that shown in Figure 7-147.9.9-1 have recently beentested successfully (Allen. - 1995). The connection detail is sensitive to the quality of weldingemployed in the critical welds, including those between the beam and column flanges., andbetween the vertical and horizontal plates and the column elements. It has been reported thatone specimen, with a known defect in the beam flange to column flange weld was informallytested and failed at low levels of loading.

The detail is also sensitive to the balance in stiffness of the various plates and flanges. Forconfigurations other than those tested, detailed finite element analyses may be necessary toconfirm that the desired uniform stress distribution is achieved. The developer of this detailindicates that for certain column profiles, it may be possible to omit the vertical slots in thecolumn web and still achieve the desired uniform beam flange stress distribution.

This detail may also be sensitive to the toughness of the column base metal at the region ofthe fillet between the web and flanges. In heavy shapes produced by some rolling processes themetal in this region may have substantially reduced toughness properties relative to the balanceof the section. This condition, coupled with local stress concentrations induced by the slot in theweb may have the potential to initiate premature fracture. The developer believes that it isessential to perform detailed analyses of the connection configuration, in order to avoid suchproblems. Popov tested one specimen incorporating a locally softened web, but without thevertical and horizontal stiffener plates contained in the detail shown in Figure 7-14. Thatspecimen failed by brittle fracture through the column flange which progressed into the holes cutinto the web. The stress patterns induced in that specimen, however, were significantly differentthan those which occur in the detail shown in the figure.

Quantitative Results: Number of specimens tested: 2Girder Size: W 27x94Column Size: W 14x176Plastic Rotation Achieved:

Specimen 1: 0.025 radianSpecimen 2: 0.030 radian

Quantitative data on connection testing may be obtained from thelicensor.

7.9.10 Bolted Bracket Connections

Framing connections employing bolted or riveted brackets have been used in structural steelconstruction since its inception. Early connections of this type were often quite flexible, and alsohad limited strength compared to the members they were connecting, resulting in partiallyrestrained type framing. However, it is possible to construct heavy bolted brackets employing


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high strength bolts to develop fully restrained moment connections. Pretensioing of the bolts orthreaded rods attaching the brackets to the column flanges and use of slip-critical connectionsbetween the brackets and beam flanges can help to provide the rigidity required to obtain fullyrestrained behavior. Reinforcement of the column flanges may be required to prevent localyielding and excessive deformation of these elements, as well. Two alternative configurations thathave been tested recently are illustrated in Figure 7.9.10-1. The developer of these configurationsoffers the brackets in the form of proprietary steel castings. Several tests of these alternativeconnections have been performed on specimens with beams ranging in size from W16 to W36sections and with large plastic rotations successfully achieved.

Design Issues: The concept of bolted bracket connections is similar to that of the riveted “windconnections” commonly installed in steel frame buildings in the early twentieth century. Theprimary difference is that the riveted wind connections were typically limited in strength eitherby flexural yielding of outstanding flanges of the brackets, or shear and tension on the rivets,rather than by flexural hinging of the connected framing. Since the old-style wind connectionscould not typically develop the flexural strength of the girders and also could be quite flexible,they would be classified either as partial strength or partially restrained connections. Followingthe Northridge earthquake, the concept of designing such connections with high strength boltsand heavy plates, to behave as fully restrained connections, was developed and tested by aprivate party who has applied for patents on the concept of using steel castings for this purpose.

PipePlate

High tensilethreaded rod

Bolts

Bracket

WARNING: The information presented in this figure is PROPRIETARY. US and ForeignPatents have been applied for. Use of this information is strictly prohibited except as authorizedin writing by the developer. Violators shall be prosecuted in accordance with US and ForeignPatent Intellectual Property Laws.

Figure 7.9.10-1 Bolted Bracket Connections

Bolted connections offer a number of potential advantages over welded connections. Since nofield welding is required for these connections, they are inherently less labor intensive during


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erection, and also less dependent on the technique of individual welders for successfulperformance. However, quality assurance should be provided for installation and tensioning ofthe bolts, as well as correction of any problems with fit-up due to fabrication tolerances.

Experimental Results: A series of tests on several different configurations of proprietary heavybolted bracket connections have been performed at Lehigh University (Ksai & Bleiman, 1996) toqualify these connections for use in repair and modification applications. To test repairapplications, brackets were placed only on the bottom beam flange to simulate installations on aconnection where the bottom flange weld in the original connection had failed. In thesespecimens, bottom flange welds were not installed, to approximate the condition of a fullyfractured weld. The top flange welds of these specimens were made with electrodes rated fornotch toughness, to preclude premature failure of the specimens at the top flange. Forspecimens in which brackets were placed at both the top and bottom beam flanges, both weldswere omitted. Acceptable plastic rotations were achieved for each of the specimens tested.

Quantitative Results: No. of specimens tested: 8Girder Size: W16x40 and W36x150Column Size: W12x65 and W14x425Plastic Rotation achieved - 0.05 radians - 0.07 radians

7.10 Other Types of Welded Connection Structures


7.10.1 Eccentrically Braced Frames (EBF)


7.10.2 Dual Systems


7.10.3 Welded Base Plate Details


7.10.4 Vierendeel Truss Systems


7.10.5 Moment Frame Tubular Systems



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7.10.6 Welded Connections of Collectors, Ties and Diaphragm Chords


7.10.7 Welded Column Splices


7.10.8 Built-up Moment Frame Members



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Metallurgy & Welding

8-1

8. METALLURGY & WELDING

8.1 Parent Materials

8.1.1 Steels

Designers should specify materials which are readily available for building construction and whichwill provide suitable ductility and weldability for seismic applications. Structural steels which may beused in the lateral-force-resisting systems for structures designed for seismic resistance without specialqualification include those contained in Table 8.1.1-1. Refer to the applicable ASTM referencestandard for detailed information.

Table 8.1.1-1 - Structural Steel Prequalified for Use in Seismic Lateral-Force-Resisting Systems

ASTM Specification DescriptionASTM A36 Carbon Structural SteelASTM A283Grade D

Low and Intermediate Tensile Strength Carbon Steel Plates

ASTM A500(Grades B & C)

Cold-Formed Welded & Seamless Carbon Steel Structural Tubing in Rounds &Shapes

ASTM A501 Hot-Formed Welded & Seamless Carbon Steel Structural TubingASTM A572(Grades 42 & 50)

High-Strength Low-Alloy Columbium-Vanadium Steels of Structural Quality

ASTM A588 High-Strength Low-Alloy Structural Steel (weathering steel)ASTM A9921 Steel for Structural Shapes for Use in Building FramingNotes:1- See Commentary

Structural steels which may be used in the lateral-force-resisting systems of structures designed forseismic resistance with special permission of the building official are those listed in Table 8.1.1-2. Steelmeeting these specifications has not been demonstrated to have adequate weldability or ductility forgeneral purpose application in seismic-force-resisting systems, although it may well possess suchcharacteristics. In order to demonstrate the acceptability of these materials for such use in WSMFconstruction it is recommended that connections be qualified by test, in accordance with the guidelinesof Chapter 7. The test specimens should be fabricated out of the steel using those welding proceduresproposed for use in the actual work.

Table 8.1.1-2 - Non-prequalified Structural Steel

ASTM Specification DescriptionASTM A242 High-Strength Low-Alloy Structural SteelASTM A709 Structural Steel for BridgesASTM A913 High-Strength Low-Alloy Steel Shapes of Structural Quality, Produced by

Quenching & Self-Tempering Process



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Commentary: Many WSMF structures designed in the last 10 years incorporatedASTM A36 steel for the beams and ASTM A572 grade 50 steel for the columns. This provided an economical way to design structures for the strong column -weak beam provisions contained in the building code. Recent studies conductedby the Structural Shape Producers Council (SSPC), however, indicate thatmaterial produced to the A36 specification has wide variation in strengthproperties with actual yield strengths that often exceed 50 ksi. This widevariation makes prediction of connection and frame behavior difficult. Somehave postulated that one of the contributing causes to damage experienced in theNorthridge earthquake was inadvertent pairing of overly strong beams withaverage strength columns.

The AISC and SSPC have been working for several years to develop a newspecification for structural steel that would have both minimum and maximumyield values defined and provide for a margin between maximum yield andminimum ultimate tensile stress. AISC recently submitted such a specification,for a material with 50 ksi specified yield strength, to ASTM for development intoa standard specification. ASTM formally adopted the new specification forstructural shapes, with a yield strength of 50 ksi, under designation A992 in 1998and It is anticipated that domestic mills will begin have begun producingstructural wide flange shapes to this specification. within a few years and thateventually, this new material will replace A36 as the standard structural materialfor incorporation into lateral-force-resisting systems.

Since the formal approval of the A992 specification by ASTM occurred afterpublication of the 1997 editions of the building codes and the AISC SeismicSpecification, it is not listed in any of these documents as a prequalified materialfor use in lateral force resisting systems. Neither is it listed as prequalified inAWS D1.1-98. However, all steel that complies with the ASTM-992 specificationwill also meet the requirements of ASTM A572, Grade 50 and should therefore bepermissible for any application for which the A572 material is approved. Seealso, the commentary to Section 8.2.2.

Under certain circumstances it may be desirable to specify steels that are notrecognized under the UBC for use in lateral-force-resisting systems. Forinstance, ASTM A709 might be specified if the designer wanted to place limits ontoughness for fracture-critical applications. In addition, designers may wish tobegin incorporating ASTM A913, Grade 65 steel, as well as other higher strengthmaterials, into projects, in order to again be able to economically design forstrong column - weak beam conditions. Designers should be aware, however, thatthese alternative steel materials may not be readily available. It is alsoimportant when using such non-prequalified steel materials, that precautions betaken to ensure adequate weldability of the material and that it has sufficientductility to perform under the severe loadings produced by earthquakes. The



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cyclic test program recommended by these Interim Guidelines for qualification ofconnection designs, by test, is believed to be an adequate approach to qualifyalternative steel material for such use as well.

Note that ASTM A709 steel, although not listed in the building code asprequalified for use in lateral-force-resisting systems, actually meets all of therequirements for ASTM A36 and ASTM A572. Consequently, specialqualification of the use of this steel should not be required.

Although the 1994 editions of the Uniform Building Code and the NEHRPProvisions do not prequalify the use of ASTM A913 steel in lateral force resistingsystems, the pending 1997 edition of the UBC does prequalify its use. Both the1997 NEHRP Provisions and the AISC Seismic Provisions prequalify the use ofthis steel in elements that do not undergo significant yielding, for example, thecolumns of moment-resisting frames designed to meet strong column - weak beamcriteria. Consequently, special approval of the Building Official should nolonger be required as a pre-condition of the use of material conforming to thisspecification, at least for columns.

8.1.2 Chemistry


Commentary: Some concern has been expressed with respect to the movement inthe steel producing industry of utilizing more recycled steel in its processes. Thisresults in added trace elements not limited by current specifications. Althoughthese have not been shown quantitatively to be detrimental to the performance ofwelding on the above steels, a the new A992specification for structural steelproposed by AISC does place more control on these trace elements. Mill testreports now include elements not limited in some or all of the specifications. They include copper, columbium, chromium, nickel, molybdenum, silicon andvanadium. The analysis and reporting of an expanded set of elements should bepossible, and could be beneficial in the preparation of welding procedurespecifications (WPSs) by the welding engineer if critical welding parameters arerequired. Modern spectrographs used by the mills are capable of automatedanalyses. When required by the engineer, a request for special supplementalrequests should be noted in the contract documents.

8.1.3 Tensile/Elongation Properties

Mechanical property test specimens are taken from rolled shapes or plates at the rolling mill in themanner and location prescribed by ASTM A6 and ASTM A370. Table 8-3 Table 8.1.3-1 gives thebasic mechanical requirements for commonly used structural steels. Properties specified, andcontrolled by the mills, in current practice include minimum yield strength or yield point, ultimate



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tensile strength and minimum elongation. However, there can be considerable variability in the actualproperties of steel meeting these specifications.

SSPC, in cooperation with SEAOC, has collected statistical data on the strength characteristics oftwo grades (ASTM A36 and ASTM A572 Grade 50) of structural steels, based on mill test reportsfrom selected domestic producers for the 1992 production year. Data were also collected for "DualGrade" material that was certified by the producers as complying with both ASTM A36 and ASTMA572 Grade 50. Table 8-4 Table 8.1.3-2 summarizes these results as well as data provided by a singleproducer for ASTM A913 material.

Unless special precautions are taken to limit the actual strength of material incorporated into thework to defined levels, new material specified as ASTM A36 should be assumed to be the dual gradefor connection demand calculations, whenever the assumption of a higher strength will result in a moreconservative design condition.

Table 8-3 Table 8.1.3-1 - Typical Tensile Requirements for Structural Shapes

ASTMMinimum YieldStrength or Yield

Point, Ksi

Ultimate TensileStrength, Ksi

Minimum Elongation%

in 2 inches

Minimum Elongation%

in 8 inchesA36 36 Min. 58-801 212 20A242 424 Min.. 63 MIN. 213 18

A572, Gr. 42 42 Min. 60 Min. 24 20A572, GR50 50 Min. 65 MIN. 212 18

A588 50 Min. 70 MIN. 213 18A709, GR36 36 Min. 58-80 212 20A709, GR50 50 Min. 65 MIN. 21 18A913, GR50 50 Min. 65 MIN. 21 18A913, GR65 65 Min. 80 MIN. 17 15

A992 50 Min. – 65 Max. 65 MIN 21 18Notes: 1- No maximum for shapes greater than 426 lb./ft.

2- Minimum is 19% for shapes greater than 426 lb. /ft.3- No limit for Shape Groups 1, 2 and 3.Minimum is 18% for shapes greater than 426 lb./ft.4. Minimum is 50 ksi for Shape Groups 1 and 2, 46 ksi for Shape Group 3, and 42 ksi for Shape Groups 4

and 5.



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Table 8-4 Table 8.1.3-2 - Statistics for Structural Shapes1,2

Statistic A 36 DualGRADE

A572GR50

A913GR65

Yield Point (ksi) Mean 49.2 55.2 57.6 75.3 Minimum 36.0 50.0 50.0 68.2 Maximum 72.4 71.1 79.5 84.1 Standard Deviation [ s ] 4.9 3.7 5.1 4.0 Mean + 1 s 54.1 58.9 62.7 79.3

Tensile Strength (ksi) Mean 68.5 73.2 75.6 89.7 Minimum 58.0 65.0 65.0 83.4 Maximum 88.5 80.0 104.0 99.6 Standard Deviation [ s ] 4.6 3.3 6.2 3.5 Mean + 1 s 73.1 76.5 81.8 93.2

Yield/Tensile Ratio Mean 0.72 0.75 0.76 0.84 Minimum 0.51 0.65 0.62 0.75 Maximum 0.93 0.92 0.95 0.90 Standard Deviation [ s ] 0.06 0.04 0.05 0.03 Mean + 1 s 0.78 0.79 0.81 0.87 Mean - 1 s 0.66 0.71 0.71 0.81

1: The data presented for ASTM A36, “Dual Grade” and ASTM A572 Grade 50 were included aspart of the SSPC study (SSPC-1994). The data for ASTM A913 were derived from a singleproducer and may not be available from all producers.

2. Statistical Data on the distribution of strength properties for material meeting ASTM A992 are notpresently available. Pending the development of such statistics, it should be assumed that A992material will have similar properties to ASTM A572, Gr. 50 material.

Commentary: The data given in Table 8-4 Table 8.1.3-2 for A36 and A572Grade 50 is somewhat weighted by the lighter, Group 1 shapes that will notordinarily be used in WSMF applications. Excluding Group 1 shapes andcombining the Dual Grade and A572 Grade 50 data results in a mean yieldstrength of 48 ksi for A36 and 57 ksi for A572 Grade 50 steel. It should also benoted that approximately 50% of the material actually incorporated in a projectwill have yield strengths that exceed these mean values. For the design offacilities with stringent requirements for limiting post-earthquake damage,consideration of more conservative estimates of the actual yield strength may bewarranted.

Until recently, In wide flange sections the tensile test coupons in wide flangesections are currently were taken from the web. The amount of reduction rolling,finish rolling temperatures and cooling conditions affect the tensile and impact



8-6

properties in different areas of the member. Typically, the web exhibits about fivepercent higher strength than the flanges due to faster cooling. In 1998 ASTM A6was revised to specify that coupons be taken from the flange of wide flangeshapes.

Design professionals should be aware of the variation in actual propertiespermitted by the ASTM specifications. This is especially important for yieldstrength. Yield strengths for ASTM A36 material have consistently increased overthe last 15 years so that several grades of steel may have the same properties orreversed properties, with respect to beams and columns, from those the designerintended. Investigations of structures damaged by the Northridge earthquakefound some WSMF connections in which beam yield strength exceeded columnyield strength despite the opposite intent of the designer.

As an example of the variations which can be found, Table 8-5 Table 8.1.3-2presents the variation in material properties found within a single buildingaffected by the Northridge earthquake. Properties shown include measured yieldstrength (Fya,), measured tensile strength (Fua ) and Charpy V-Notch energy rating(CVN).

Table 8-5Table 8.1.3-2 - Sample Steel Properties from a Building Affected by the NorthridgeEarthquake

Shape Fya1 ksi Fua, ksi CVN, ft-lb.

W36 X 182 38.0 69.3 18

W36 X 230 49.3 71.7 195Note 1 - ASTM A36 material was specified for both structures.

The practice of dual certification of A36 and A572, Grade 50 can result inmean yield strengths that are fifty percent higher than the specified yield of A36. Since there is no practical way to discern whether dual grade steel will besupplied, unless direct purchase of steel from specific suppliers is made, in theabsence of such procurement practices, the prudent action for determiningconnection requirements, where higher strengths could be detrimental to thedesign, would be to assume the dual grade material whenever A36 or A572 Grade50 is specified.

In the period since the initial publication of the Interim Guidelines, severalresearchers and engineers engaged in connection assembly prototype testing havereported that tensile tests on coupons extracted from steel members used in theprototype tests resulted in lower yield strength than reported on the mill testreport furnished with the material, and in a few cases lower yield strength thanwould be permitted by the applicable ASTM specification. This led to someconfusion and concern, as to how mill test reports should be interpreted.



8-7

The variation of the measured yield strength of coupons reported byresearchers engaged in connection prototype testing, as compared to thatindicated on the mill test reports, is not unusual and should be expected. Thesevariations are the result of a series of factors including inconsistencies betweenthe testing procedures employed as well as normal variation in the material itself. The following paragraphs describe the basis for the strengths reported byproducers on mill certificates, as well as the factors that could cause independentinvestigators to determine different strengths for the same material.

Mill tests of mechanical properties of steel are performed in accordance withthe requirements of ASTM specifications A6 and A370. ASTM A6 hadhistorically required that test specimens for rolled W shapes be taken from thewebs of the shapes, but recently was revised to require testing from the flanges ofwide flange shapes with 6 inch or wider flanges. A minimum of two tests must bemade for each heat of steel, although additional tests are required if shapes ofsignificantly different thickness are cast from the same heat. Coupon size andshape is specified based on the thickness of the material. The size of the couponused to test material strength can effect the indicated value. Under ASTM A6,material that is between 3/4 inches thick and 4 inches thick can either be tested infull thickness “straps” or in smaller 1/2” diameter round specimens. In thickmaterial, the yield strength will vary through the thickness, as a result of coolingrate effects. The material at the core of the section cools most slowly, has largergrain size and consequently lower strength. If full-thickness specimens are used,as is the practice in most mills, the recorded yield strength will be an average ofthe relatively stronger material at the edges of the thickness and the lower yieldmaterial at the center. Many independent laboratories will use the smaller 1/2”round specimens, and sometimes even sub-sized 1/4” round specimens for tensiletesting, due to limitations of their testing equipment. Use of these smallerspecimens for thick material will result in testing only of the lower yield strengthmaterial at the center of the thickness.

ASTM A370 specifies the actual protocol for tensile testing including theloading rate and method of reporting test data. Strain rate can affect the strengthand elongation values obtained for material. High strain rates result in elevatedstrength and reduced ductility. Under ASTM A370, yield values may bedetermined using any convenient strain rate, but not more than 1/16 inch perinch, per minute which corresponds to a maximum loading rate of approximately30 ksi per second. Once the yield value is determined, continued testing to obtainultimate tensile values can proceed at a more rapid rate, not to exceed 1/2 inchper inch per minute.

Under ASTM A370, there are two different ways in which the yield propertyfor structural steel can be measured and reported. These include yield point andyield strength. These are illustrated in Figure 8.1.3-1. The yield point is the peak



8-8

stress that occurs at the limit of the elastic range, while the yield strength is asomewhat lower value, typically measured at a specified offset or elongationunder load. Although a number of methods are available to determine yieldpoint, the so-called “drop of the beam” method is most commonly used forstructural steel. In this method the load at which a momentary drop-off inapplied loading occurs is recorded, and then converted to units of stress to obtainthe yield point. Yield strength may also be determined by several methods, but ismost commonly determined using the offset method. In this method, the stress -strain diagram for the test is drawn, as indicated in Figure 8.1.3-1. A specifiedoffset, typically 0.2% strain for structural steel, is laid off on the abscissa of thecurve and a line is drawn from this offset, parallel to the slope of the elasticportion of the test. The stress at the intersection of this offset line with the stress-strain curve is taken as the yield strength.

ε

σYield Point

Yield Strength

Offset

Figure 8.1.3-1 Typical Stress - Strain Curve for Structural Steel

The material specifications for structural steels typically specify minimumvalues for yield point but do not control yield strength. The SSPC has reportedthat actual practice among the mills varies, with some mills reporting yieldstrength and others reporting yield point. This practice is permissible as yieldstrength will always be a somewhat lower value than yield point, resulting in asomewhat conservative demonstration that the material meets specifiedrequirements. However, this does mean that there is inconsistency between thevalues reported by the various mills on certification reports. Similarly, theprocedures followed by independent testing laboratories may be different thanthose followed by the mill, particularly with regard to strain rate and the locationat which a coupon is obtained.



8-9

Under ASTM A6, coupons for tensile tests had historically been obtained fromthe webs of structural shapes. However, most engineers and researchers engagedin connection testing have preferred to extract material specimens from theflanges of the shape, since this is more representative of the flexural strength ofthe section. Coupons removed from the web of a rolled shape tend to exhibitsomewhat higher strength properties than do coupons removed from the flanges,due to the extra amount of working the thinner web material typically experiencesduring the rolling process and also because the thinner material cools morerapidly after rolling, resulting in finer grain size. Given these differences intesting practice, as well as the normal variation that can occur along the lengthof an individual member and between different members rolled from the sameheat, the reported differences in strength obtained by independent laboratories,as compared to that reported on the mill test reports, should not be surprising. Itis worth noting that following the recognition of these differences in testingprocedure, the SSPC in coordination with AISC and ASTM developed andproposed a revision to the A6 specification to require test specimens to be takenfrom the flanges of rolled shapes when the flanges are 6 inches or more wide. Itis anticipated that mills will begin to alter practice to conform to a revisedspecification in early 1997 This has since become the standard practice.

The discovery of the somewhat varied practice for reporting material strengthcalls into question both the validity of statistics on the yield strength of structuralsteel obtained from the SSPC study, and its relevance to the determination of theexpected strength of the material for use in design calculations. Although theyield point is the quantity controlled by the ASTM material specifications, it haslittle relevance to the plastic moment capacity of a beam section. Plastic sectioncapacity is more closely related to the stress along the lower yield plateau of thetypical stress-strain curve for structural steel. This strength may often besomewhat lower than that determined by the offset drop-of-the-beam method. Since the database of material test reports on which the SSPC study was basedappears to contain test data based on both the offset and drop-of-the-beammethods, it is difficult to place great significance in the statistics derived from itand to draw a direct parallel between this data and the expected flexural strengthof rolled shapes. It would appear that the statistics reported in the SSPC studyprovide estimates of the probable material strength that are somewhat high. Thus, the recommended design strengths presented in Tables 6.6.6.3-1 and 7.5.1-1 of the Interim Guidelines would appear to be conservative with regard to designof welds, panel zones and other elements with demands limited by the beam yield strength.

Under the phase II program of investigation, SAC, together with the shapeproducers, is engaged in additional study of the statistical distribution of yieldstrength of various materials produced by the mills. This study is intended toprovide an improved understanding of the statistical distribution of the lower



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yield plateau strength of material extracted from section flanges, measured in aconsistent manner. In addition, it will provide correlation with yield strengthsdetermined by other methods such that the data provided on mill test certificatescan be properly interpreted and utilized. In addition, the possibility of revisingthe ASTM specifications to provide for more consistent reporting of strength dataas well as the reporting of strength statistics that are directly useful in the designprocess will be evaluated. In the interim period, the data reported in Table 8-1.3-2, extracted from the SSPC study, remain the best currently availableinformation.

8.1.4 Toughness Properties


8.1.5 Lamellar Discontinuities


8.1.6 K-Area Fractures

Recently, there have beenIn the period 1995-96 there were several reports of fractures initiating inthe webs of column sections during the fabrication process, as flange continuity plates and/or doublerplates were welded into the sections. This fracturing typically initiated in the region near the filletbetween the flange and web. This region has been commonly termed the “k-area” because the AISCManual of Steel Construction indicates the dimension of the fillet between the web and flange with thesymbol “k”. The k-area may be considered to extend from mid-point of the radius of the fillet into theweb, approximately 1 to 1-1/2 inches beyond the point of tangency between the fillet and web. Thefractures typically extended into, and sometimes across, the webs of the columns in a characteristic“half-moon” or “smiley face” pattern.

Investigations of materials extracted from fractured members have indicated that the material in thisregion of the shapes had elevated yield strength, high yield/tensile ratio, high hardness and very lowtoughness, on the order of a few foot-pounds at 70oF. Material with these properties can behave in abrittle manner. Fracture can be induced by thermal stresses from the welding process or by subsequentweld shrinkage, as apparently occurred in the reported cases. There have been no reported cases of in-service k-line fracture from externally applied loading, as in beam-column connections, although such apossibility is perceived to exist under large inelastic demand.

It appears that this local embrittling of sections can be attributed to the rotary straightening processused by some mills to bring the rolled shapes within the permissible tolerances under ASTM A6. Thestraightening process results in local cold working of the sections, which strain hardens the material. The amount of cold working that occurs depends on the initial straightness of the section andconsequently, the extent that mechanical properties are effected is likely to vary along the length of amember. The actual process used to straighten the section can also affect the amount of local coldworking that occurs.



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Engineers can reduce the potential for weld-induced fracture in the k-area by avoiding weldingwithin the k-area region. This can be accomplished by detailing doubler plates and continuity platessuch that they do not contact the section in this region. The use of large corner clips on beam flangecontinuity plates can permit this. Selection of column sections with thicker webs, to eliminate the needfor doubler plates; the use of fillet welds rather than full penetration groove welds to attach doublerplates to columns, when acceptable for stress transfer; and detailing of column web doubler plates suchthat they are offset from the face of the column web can also help to avoid these fabrication-inducedfracture problems.

Commentary: It appears that detailing and fabrication practice can be adjustedto reduce the potential for k-area fracture during fabrication. However, theacceptability of having low-toughness material in the k-area region for service isa question that remains. It is not clear at this time what percentage of thematerial incorporated in projects is adversely affected, or even if a problem withregard to serviceability exists. SAC recently placed a public call, asking forreports of fabrication-induced fractures at the k-area, but only received limitedresponse. However, in one of the projects that did report this problem, asignificant number of columns were affected. This may have been contributed toby the detailing and fabrication practices applied on that project.

Other than detailing structures to minimize the use of doubler plates, and toavoid large weldments in the potentially sensitive k-area of the shape, it is notclear at this time, what approach, if any, engineers should take with regard to thisissue. There are several methods available to identify possible low notchtoughness in structural carbon steels, including Charpy V-Notch testing andhardness testing of samples extracted from the members. However, both of theseapproaches are quite costly for application as a routine measure on projects andthe need for such measures has not yet been established.

Following publication of advisories on the k-line problem by AISC, and thepublication of similar advisory information in FEMA-267a,reports on thisproblem diminished. It is not clear whether this is due to revised detailingpractice on the part of engineers and fabricators, revised mill rolling practice, ora combination of both. SAC, AISC and SSPC are continuing to research thisissue in order to identify if a significant problem exists, and if it does, todetermine its basic causes, and to develop appropriate recommendations for mill,design, detailing, and fabrication practices to mitigate the problem.

8.2 Welding

8.2.1 Welding Process




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8.2.2 Welding Procedures

Welding should be performed within the parameters established by the electrode manufacturer andthe Welding Procedure Specification (WPS), required under AWS D1.1.

Commentary: A welding procedure specification identifies all the importantparameters for making a welded joint including the material specifications of thebase and filler metals, joint geometry, welding process, requirements for pre- andpost-weld heat treatment, welding position, electrical characteristics, voltage,amperage, and travel speed. Two types of welding procedure specifications arerecognized by AWS D1.1. These are prequalified procedures and qualified-by-test procedures. Prequalified procedures are those for which the importantparameters are specified within the D1.1 specification. If a prequalifiedprocedure is to be used for a joint, all of the variables for the joint must fallwithin the limits indicated in the D1.1 specification for the specific procedure. Ifone or more variables are outside the limits specified for the prequalifiedprocedures, then the fabricator must demonstrate the adequacy of the proposedprocedure through a series of tests and submit documentation (procedurequalification records) demonstrating that acceptable properties were obtained. Regardless of whether or not a prequalified or qualified-by-test procedure isemployed, the fabricator should prepare a welding procedure specification, whichshould be submitted to the engineer of record for review and be maintained at thework location for reference by the welders and inspectors. The followinginformation is presented to help the engineer understand some of the issuessurrounding the parameters controlled by the welding procedure specification.

For example, the position (if applicable), electrode diameter, amperage orwire feed speed range, voltage range, travel speed range and electrode stickout(e.g. all passes, 0.072 in. diameter, 248 to 302 amps, 19 to 23 volts, 6 to 10inches/minute travel speed, 170 to 245 inches/minute wire feed speed, 1/2" to 1"electrode stickout) should be established. This information is generally submittedby the fabricator as part of the Welding Procedure Specification. Its importancein producing a high quality weld is essential. The following information ispresented to help the engineer understand some of the issues surrounding theseparameters.

The amperage, voltage, travel speed, electrical stickout and wire feed speedare functions of each electrode. If prequalified WPSs are utilized, theseparameters must be in compliance with the AWS D1.1 requirements. For FCAWand SMAW, the parameters required for an individual electrode vary frommanufacturer to manufacturer. Therefore, for these processes, it is essential thatthe fabricator/erector utilize parameters that are within the range ofrecommended operation published by the filler metal manufacturer. Alternately,the fabricator/erector could qualify the welding procedure by test in accordance



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with the provisions of AWS D1.1 and base the WPS parameters on the test results.For submerged arc welding, the AWS D1.1 code provides specific amperagelimitations since the solid steel electrodes used by this process operate essentiallythe same regardless of manufacture. The filler metal manufacturer’s guidelineshould supply data on amperage or wire feed speed, voltage, polarity, andelectrical stickout. The guidelines will not, however, include information ontravel speed which is a function of the joint detail. The contractor should select abalanced combination of parameters, including travel speed, that will ensure thatthe code mandated weld-bead sizes (width and height) are not exceeded.

Recently, ASTM approved a new material specification for structural steelshape, ASTM A992. This specification is very similar to the ASTM A572, Grade50 specification except that it includes additional limitations on yield and tensilestrengths and chemical composition. Although material conforming to A992 isexpected to have very similar welding characteristics to A572 material, it wasadopted too late to be included as a prequalified base material in AWS D1.1-98. Although the D1 committee has evaluated A992 and has taken measures toincorporate it as a prequalified material in AWS D1.1-2000, technically, underAWS D1.1-98, welded joints made with this material should follow qualified-by-test procedures.

In reality, structural steel conforming to ASTM A992 may actually havesomewhat better weldability than material conforming to the A572 specification.This is because A992 includes limits on carbon equivalent, precluding thedelivery of steels where all alloys simultaneously approach the maximumspecified limits. Therefore, it should be permissible to utilize prequalifiedprocedures for joint with base metal conforming to this specification.

8.2.3 Welding Filler Metals


Commentary: Currently, there are no notch toughness requirements for weldmetal used in welding ASTM A 36 or A 572, Grade 50, steel in AWS D1.1. Thistopic has been extensively discussed by the Welding Group at the JointSAC/AISC/AISI/NIST Invitational Workshop on September 8 and 9, 1994, and byall participants of the SAC Invitational Workshop on October 28 and 29, 1994.The topic was also considered by the AWS Presidential Task Group, whichdecided that additional research was required to determine the need fortoughness in weld metal. There is general agreement that adding a toughnessrequirement for filler metal would be desirable and easily achievable. Most fillermetals are fairly tough, but some will not achieve even a modest requirement suchas 5 ft-lb. at + 70? F. What is not in unanimous agreement is what level oftoughness should be required. The recommendation from the Joint Workshop was



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20 ft-lb. at -20? F per Charpy V-Notch [CVN] testing. The recommendationfrom the SAC Workshop was 20 ft-lb. at 30? F lower than the Lowest AmbientService Temperature (LAST) and not above 0? F. The AWS Presidential TaskGroup provided an interim recommendation for different toughness valuesdepending on the climatic zone, referenced to ASTM A709. Specifically, therecommendation was for 20 ft-lb. at temperatures of 70 degrees F for Zone 1, 40degrees F for Zone 2, and 10 degrees F for Zone 3. The AWS also suggestedtoughness values for base metals used in these applications.

Some fractured surfaces in the Northridge and Kobe Earthquakes revealedevidence of improper use of electrodes and welding procedures. Prominentamong the misuses were high production deposition rates. Pass widths of up to 1-1/2 inches and pass heights of 1/2 inch were common. The kind of heat inputassociated with such large passes promotes grain growth in the HAZ andattendant low notch toughness. In evaluation of welds in buildings affected by theNorthridge earthquake, the parameters found to be most likely to result indamage-susceptible welds included root gap, access capability, electrodediameter, stick-out, pass thickness, pass width, travel speed, wire feed rate,current and voltage were found to be the significant problems in evaluation ofwelds in buildings affected by the Northridge earthquake.

Welding electrodes for common welding processes include:

AWS A5.20: Carbon Steel Electrodes for FCAWAWS A5.29: Low Alloy Steel Electrodes for FCAWAWS A5.1: Carbon Steel Electrodes for SMAWAWS A5.5: Low Alloy Steel Covered Arc Welding Electrodes (for SMAW)AWS A5.17: Carbon Steel Electrodes and Fluxes for SAWAWS A5.23: Low Alloy Steel Electrodes and Fluxes for SAWAWS A5.25: Carbon and Low Alloy Steel Electrodes and Fluxes for Electroslag

Welding

In flux cored arc welding, one would expect the use of electrodes that meeteither AWS A5.20 or AWS A5.29 provided they meet the toughness requirementsspecified below.

Except to the extent that one requires Charpy V-Notch toughness andminimum yield strength, the filler metal classification is typically selected by theFabricator. Compatibility between different filler metals must be confirmed bythe Fabricator, particularly when SMAW and FCAW-SS processes are mixed. Generally speaking, SMAW-type filler metals may not be applied to FCAW-SStype filler metals (e.g. when a weld has been partially removed) while FCAW-typefiller metals may be applied to SMAW-type filler metals. This recommendationconsiders the use of aluminum as a killing agent in FCAW-SS electrodes that can



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be incorporated into the SMAW filler metal with a reduction in impact toughnessproperties.

As an aid to the engineer, the following interpretation of filler metalclassifications is provided below:

E1X2X3T4X5 For electrodes specified under AWS A5.20 (e.g. E71T1)E1X2X3T4X5X6 For electrodes specified under AWS A5.29 (e.g. E70TGK2)E1XX7X8X9X10 For electrodes specified under AWS A5.1 or AWS A5.5. (e.g. E7018)

NOTES:

1. Indicates an electrode.

2. Indicates minimum tensile strength of deposited weld metal (in tens of ksi, e.g., 7 = 70ksi).

3. Indicates primary welding position for which the electrode is designed (0 = flat andhorizontal and 1 = all positions).

4. Indicates a flux cored electrode. Absence of a letter indicates a "stick" electrode forSMAW.

5. Describes usability and performance capabilities. For our purposes, it conveys whetheror not Charpy V-Notch toughness is required (1, 5, 6 and 8 have impact strengthrequirements while 2, 4, 7, 10 and 11 do not). A "G" signifies that the properties are notdefined by AWS and are to be agreed upon between the manufacturer and the specifier. Impact strength is specified in terms of the number of foot-pounds at a given temperature(e.g., 20 ft-lb. at 0 degrees F). Note that for electrodes specified under AWS A5.20, theformat for usage is "T-X".

6. Designates the chemical composition of deposited metal for electrodes specified underAWS A5.29. Note that there is no equivalent format for chemical composition forelectrodes specified under AWS A5.20.

7. The first two digits (or three digits in a five digit number) designate the minimum tensilestrength in ksi.

8. The third digit (or fourth digit in a five digit number) indicates the primary weldingposition for which the electrode is designed (1 = all positions, 2 = flat position and filletwelds in the horizontal position, 4 = vertical welding with downward progression and forother positions.)

9. The last two digits, taken together, indicate the type of current with which the electrodecan be used and the type of covering on the electrode.



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10. Indicates a suffix (e.g., A1, A2, B1, etc.) designating the chemical composition of thedeposited metal.

Electrode Diameter: (See AWS D1.1 Section 4.14.1.2) The issue of maximumelectrode diameter has not been studied sufficiently to determine whether or notelectrode diameter is a critical variable. Recent tests have produced modifiedframe joints with acceptable test results using the previous standard-of-practice 0.120 in. diameter wire. The use of smaller diameter electrodes will slow the rateof deposition (as measured by volume) but will not, in and of itself, produce anacceptable weld. The following lists the maximum allowable electrode diametersfor prequalified FCAW WPS’s according to D1.1:

• Horizontal, complete or partial penetration welds: 1/8 inch (0.125")*• Vertical, complete or partial penetration welds: 5/64 inch (0.078")• Horizontal, fillet welds: 1/8 inch (0.125")• Vertical, fillet welds: 5/64 inch (0.078")• Overhead, reinforcing fillet welds: 5/64 inch (0.078")

* This value is not part of D1.1-94, but will be part of D1.1-96.

For a given electrode diameter, there is an optimum range of weld bead sizesthat may be deposited. Weld bead sizes that are outside the acceptable size range(either too large or too small) may result in unacceptable weld quality. The D1.1code controls both maximum electrode diameters and maximum bead sizes (widthand thickness). Prequalified WPS’s are required to meet these coderequirements. Further restrictions on suitable electrode diameters are notrecommended.

Low-hydrogen electrodes. Low hydrogen electrodes should be used to minimizethe risk of hydrogen assisted cracking (HAC) when conditions of high restraintand the potential for high hardness microstructures exist. Hydrogen assistedcracking can occur in the heat affected zone or weld metal whenever sufficientconcentrations of diffusible hydrogen and sufficient stresses are present alongwith a hard microstructure at a temperature between 100 C and –100 C. Hydrogen is soluble in steel at high temperatures and is introduced into the weldpool from a variety of sources including but not limited to: moisture from coatingor core ingredients, drawing lubricants, hydrogenous compounds on the basematerial, and moisture from the atmosphere.

At the present time, the term “low hydrogen” is not well defined by AWS. Thedegree of hydrogen control required to reduce the risk of hydrogen assistedcracking will depend on the material being welded, level of restraint,preheat/interpass temperature, and heat input level. When a controlled level ofdiffusible hydrogen is required, electrodes can be purchased with a supplementaldesignator that indicates a diffusible hydrogen concentration below 16, 8, or 4 ml



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H2/100g in the weld metal can be maintained (H16, H8, and H4 respectively)under most welding conditions .

The diffusible hydrogen potential (measured in ml/100g deposited weld metal)will depend on the type of consumable, welding process, plate/joint cleanliness,and atmospheric conditions in the area of welding. Some consumables mayabsorb moisture after exposure to the atmosphere. Depending on the type ofconsumable, this may result in a significant increase in the weld metal diffusiblehydrogen concentration. In situations where control of diffusible hydrogenconcentrations is important, the manufacturer should be consulted for advice onproper storage and handling conditions required to limit moisture absorption.

Hydrogen assisted cracking may be avoided through the selection andmaintenance of an adequate preheat /interpass temperature and/or minimum heatinput. Depending on the type of steel and restraint level, a trade-off between aneconomic preheat/interpass temperature and the diffusible hydrogen potential ofa given process exists. There have been several empirical approaches developedto determine safe preheat levels for a given application that include considerationof carbon equivalent, restraint level, electrode type, and preheat. When followed,the guidelines for preheat that have been established in AWS D1.1 and D1.5 aregenerally sufficient to reduce the risk of hydrogen assisted cracking in most mildsteel weldments.

Hydrogen assisted cracking will typically occur up to 72 hours after completionof welding. For the strength of materials currently used in moment frameconstruction, inspection of completed welds should be conducted no sooner than24 hours following weld completion.

8.2.4 Preheat and Interpass Temperatures


8.2.5 Postheat


8.2.6 Controlled Cooling


8.2.7 Metallurgical Stress Risers




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8.2.8 Welding Preparation & Fit-up
