Seismic Repair and Strengthening of Lap Splices inRC Columns: Carbon Fiber–Reinforced Polymer

versus Steel ConfinementAmer M. ElSouri1 and Mohamed H. Harajli2

Abstract: A design approach, developed specifically for seismic bond strengthening of the critical splice region of reinforced concretecolumns or bridge piers, is presented and discussed. The approach is based on providing adequate concrete confinement within the splicezone for allowing the spliced bars to theoretically develop enough postelastic tension strains demanded by large earthquakes before experi-encing splitting bond failure. The accuracy of the approach was validated experimentally by evaluating the seismic behavior of full-scalegravity load-designed (as-built) rectangular columns that were strengthened or repaired in accordance with the proposed approach. Threetypes of confinement were used and compared, namely, internal steel ties, external fiber polymer reinforced jackets, and a combination ofboth. The repaired/strengthened columns developed sizable postyield strains of the spliced bars, considerable increases in the lateral load anddrift capacities, and much less concrete damage within the splice zone when compared with the as-built columns. As a further support ofthe adequacy of the design strengthening approach, the backbone lateral load-drift response of the strengthened columns showed a goodagreement with the envelope response generated using nonlinear flexural analysis assuming perfect bond between the column reinforcementand concrete. DOI: 10.1061/(ASCE)CC.1943-5614.0000213. © 2011 American Society of Civil Engineers.

CE Database subject headings: Seismic effects; Concrete columns; Steel; Confinement; Fiber reinforced polymer; Nonlinear analysis;Rehabilitation.

Author keywords: Bond; Column; Concrete; Reinforced; Confinement; Cyclic; Fiber reinforced polymers; Nonlinear analysis; Repair;Strengthening; Seismic.

Introduction

It was a common practice before the 1980s to lap splice columnreinforcement in building structures or pier reinforcement in bridgestructures at the base of the column or pier. While this constructiontechnique works well in areas of low or no seismic hazard, fieldexperience has demonstrated the shortcoming of this practice whenthe structure is subjected to strong ground motions in areas of highseismic hazard. The fact that the splices are located within the criti-cal hinging region of the column or pier, large lateral drift reversalsinduced by strong earthquakes often lead to bond splitting of thespliced bars and consequently rapid bond and flexural strengthdegradation of the structure. In fact, most failures of concretebridges or buildings following major earthquakes around the worldare attributed to substandard detailing and inferior bond perfor-mance of the steel reinforcement at the critical regions of thestructures where plastic hinging is likely to develop. At present,the ACI building code (2008) specifies that lap splicing of columnreinforcement is allowed only at column midheight where steelstress reversals are theoretically minimal.

This paper presents a design procedure and application of a sys-tem for seismic bond strengthening of the critical splice region ofreinforced concrete columns or bridge piers. The strengtheningsystem is based on improving the bond capacity of the spliced barsby providing adequate confinement within the splice zone forallowing the bars to theoretically develop enough postelastic ten-sion strain demanded by large earthquakes. Different design ap-proaches for evaluating the amount of confinement needed for“adequate” seismic bond strengthening were derived and com-pared. The strengthening system was applied on a number of full-scale reinforced concrete columns before being subjected to lateralload, as well as columns that have already experienced damageunder large drift reversals. To test the accuracy of the proposeddesign approaches, several practical means of concrete confinementand their effect on the seismic response of the columns were ex-perimentally evaluated and compared. These include internal con-finement using steel ties, external confinement using carbon fiberreinforced polymer (CFRP) jackets, or a combination of both. Thedesign strengthening approach was further validated by comparingthe backbone load-deformation response of the strengthened col-umns against the envelope response generated using nonlinearanalysis of the same columns but assuming instead a perfect bondbetween the steel bars and concrete.

Design Approach for Seismic Bond Strengtheningof Lap Splices

Derivation of the Approach

Earlier analytical and experimental studies have demonstrated thatthe bond resistance in general and the response of the bond-critical

1Ph.D. Student, Dept. ofCivil andEnvironmental Engineering,AmericanUniv. of Beirut (AUB), Beirut, Lebanon. E-mail: ame58@aub.edu.lb

2Professor, Dept. of Civil and Environmental Engineering, Faculty ofEngineering and Architecture, American Univ. of Beirut (AUB), Beirut,Lebanon (corresponding author). E-mail: mharajli@aub.edu.lb

Note. This manuscript was submitted on October 3, 2010; approved onFebruary 23, 2011; published online on February 25, 2011. Discussionperiod open until March 1, 2012; separate discussions must be submittedfor individual papers. This paper is part of the Journal of Composites forConstruction, Vol. 15, No. 5, October 1, 2011. ©ASCE, ISSN 1090-0268/2011/5-721–731/$25.00.

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zones in particular under static or seismic loading can be improvedsubstantially using internal concrete confinement by steel ties(Oragun et al. 1977; Darwin et al. 2005; ACI Committee 408 1992)or external confinement by FRP laminates (Aquino and Hawkins2007; Ghosh and Sheikh 2007; Harajli et al. 2004; Harajli andDagher 2008; Harajli and Khalil 2008; Harries et al. 2006; Hawkinset al. 2000; Kim et al. 2008; Priestly et al. 1996; Seible et al. 1997).

For design application, the bond strength of developed orspliced bars in tension can be related to the concrete cover, thebar diameter, the splice length, the concrete compressive strength,and the area of transverse confining reinforcement per unit lengthwithin the development/splice zone. In this study, three differentdesign expressions for seismic bond strengthening are developedand compared. Starting with steel-confined concrete, the bondcapacity of spliced bars in tension may be calculated using oneof the following expressions:Orangun, Jirsa, and Breen (1977) (referred to as OJB):

Us

f 01=2c

¼ 0:1þ 0:25cdb

þ 4:15dbLs

þ Atrf yt41:6snsdb

ð1Þ

Darwin et al. (2005):

Us

f 01=2c

¼ 1

f 01=4c

�0:23þ 0:46

cmdb

þ 14:1dbLs

��0:1

cMcm

þ 0:9�

þ f 01=4c3:0trtdAtr

snsdbð2Þ

CEB-FIP model code (1990):

Us ¼1:64

�f 0c�2:75

10

�2=3

�1:15� 0:1 c

db

��1� K

AtrLs=s�P

Atrm

Ab

� ð3Þ

where Us (MPa) = average bond strength, at bond failure, in steel-confined concrete; f 0c (MPa) = concrete compressive strength; Ls(mm) = splice/development length; db (mm) = diameter of thespliced or developed bar; c (mm) in Eqs. (1) and (3) = smaller ofside concrete cover of the steel bars cs, bottom cover cb, or one-halfthe clear horizontal spacing between the bars, while in Eq. (2) cm =minimum and cM = maximum value (cM=cm ≤ 3:5) of cs and cb,where cs = smaller of side cover or one-half the clear distance be-tween bars þ6:0 mm; Atr (mm2) = total cross sectional area perspacing s (mm) of all transverse reinforcement crossing the poten-tial plane of splitting through the reinforcement being developed;ns = number of spliced or developed bars in tension; f yt (MPa) =yield strength of the transverse reinforcement; tr ¼ 9:6Rr þ 0:28,where Rr is the relative rib area ratio, which for conventional barscan be taken equal to 0.0727 and hence tr can be taken equal to 1.0;td ðmmÞ ¼ 0:031db þ 0:22; K ¼ 0:1 for a bar confined at a cornerbend of a stirrup or tie, 0.05 for a bar confined with a tie, and 0 forunconfined concrete; and

PAtrm ¼ 0:25Ab for beams and zero for

slabs, where Ab (mm) is the area of the largest bar being developed.Notice that in Eq. (1) the value of Atrf yt=ð41:6snsdbÞ is limited to0.25 to safeguard against pullout bond failure. For plain unconfinedconcrete, the terms Atr and K in the preceding expressions are setequal to 0.

In 2004, Harajli et al. developed the following experimentallybased expression for evaluating the increase in average bondstrength Ufrp, at bond failure, owing to external confinement byFRP jackets:

Ufrp

f 01=2c

¼ Ef nf tf4;000nsdb

ð4Þ

where nf = number of FRP wraps; tf = design thickness per wrap;and Ef = modulus of elasticity of the FRP material. Eq. (4)is applicable for both rectangular and circular sections. To safe-guard against pullout bond failure, the magnitude of Ufrp=f

01=2c was

limited to 0.4.The total bond strength Uc of concrete confined with a combi-

nation of internal steel ties and FRP jackets can be expressed bycombining Eq. (4) linearly with either of Eqs. (1) and (2) or (3), i.e.,Uc ¼ Us þ Ufrp. Using this combination, setting the equilibriumbetween the bar force and bond force (i.e., UcπdbLs ¼ Ab f s, whereAb is the area of one spliced/developed bar), and assuming for sim-plicity that the multiplier ð0:1cM=cm þ 0:9Þ in Eq. (2) is equal tounity, the following expressions are obtained for estimating theamount of confining reinforcement (steel or FRP or a combinationof both) required to develop a “design” steel stress f s in the splicedbars at bond failure (these expressions are still referred to as OJB,Darwin et al., and CEB-FIP for convenience):OJB:�

Atrf yt10:4stnsdb

þ Ef nf tf1;000nsdb

�¼ f s

f 01=2c

dbLs

��cdb

þ 16:6dbLs

þ 0:4

�

ð5ÞDarwin et al.:�

f 01=4c12:0tdAtr

stnsdbþ Ef nf tf1;000nsdb

�

¼ f s

f 01=2c

dbLs

� 1:84cm=db þ 56:4db=Ls þ 0:92

f 01=4c

ð6Þ

CEB-FIP:0B@ 1:64

�f 0c�2:75

10

�2=3

�1:15� 0:1 c

db

��1� K AtrLs=s�0:25Ab

Ab

�þ f 01=2cEf nf tf

4;000nsdb

1CA¼ f sdb

4Ls

ð7Þ

Eq. (4) and design Eqs. (5)–(7) apply for the case when the FRPis wrapped continuously along the splice length, which is prefer-able for seismic applications. For the case in which the FRP is com-posed of a number Nf of equally spaced discrete bands of width bfeach, the term nf tf (which in this case represents the total thicknessof each of the bands) should be replaced by the term αf nf tf , whereαf is equal to the ratio of the total FRP width Nf bf to the splicelength Ls. Also, when applying Eqs. (5)–(7) for a circular or rec-tangular section with uniformly distributed reinforcement along thesection perimeter, it would be reasonably conservative to considerns as the number of splices within the tension half of the crosssection.

It should be indicated that few other methods or expressionshave been developed for evaluating the thickness of FRP jacketsrequired for bond strengthening of spliced bars in tension (Priestleyet al. 1996; Seible et al. 1997; Hawkins et al. 2000; and Aquino andHawkins 2007). However, unlike Eq. (4), these methods or expres-sions are limited to circular sections. Furthermore, in support ofthe validity of Eq. (4) for both circular and rectangular sections,experimental studies (Harajli and Dagher 2008; Harajli and Khalil2008) have found that for normal size sections and for identicalbond parameters (concrete cover, bar diameter, splice length,and concrete compressive strength), the level of improvement ofbond capacity of spliced bars in rectangular sections owing toFRP confinement is almost identical to that in circular sections.

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Notice that for wide rectangular sections (e.g., walls) havingbending about their minor axis, the number of splices is relativelylarge, and hence, according to Eq. (4), the FRP jacket for a givenjacket stiffness Ef nf tf becomes much less effective in improvingthe splice bond strength when compared to normal size sectionswith a limited number of splices. For this type of sections, theuse of CFRP anchors (Kim et al. 2008) probably becomes morecost-effective.

Criterion for Seismic Bond Strengthening

To estimate the area of confining reinforcement (using steel, orFRP, or a combination of both) required for bond strengtheningin accordance with Eqs. (5), (6), or (7), a designer has to postulatea “safe” splice stress f s that needs to be developed. For static loadconditions, it is adequate to set f s ¼ 1:25f yðdesignÞ, which is in linewith the philosophy of the ACI building code (2008) for calculatingthe development or splice length of steel bars in tension. However,under seismic loading, strong earthquakes demand large postelasticsteel strains. Accordingly, since the actual yield strength of steelbars is normally larger than the design yield strength, in orderto avoid splitting bond failure under large steel strains/stresses de-manded by strong earthquakes, it is recommended to set a designdevelopment/splice stress for seismic application of at least 25%larger than the actual yield strength, i.e., f s ¼ 1:25f yðactualÞ. In theabsence of information about the actual yield strength, it is sug-gested to use a development/splice stress f s of at least 85% largerthan the design yield strength, i.e., f s ¼ 1:85f yðdesignÞ. This lastsuggestion complies with the experimentally based recommenda-tion of Hawkins et al. (2000) for using splice stress values of517 MPa and 758 MPa for grade 40 ( f y ¼ 276 MPa) and grade 60( f y ¼ 415 MPa) bars, respectively.

Experimental Program

As-Built Columns

The dimensions and reinforcement details of the as-built columnspecimens are provided in Fig. 1. These specimens constitute a partof an experimental program, the test results of which are presented

in detail elsewhere (Harajli and Dagher 2008). The columns werereinforced with eight bars of 14 mm diameter (column C14), eightbars of 16 mm diameter (C16), and six bars of 20 mm diameter(C20). A summary of the steel yield strength, concrete compressivestrength, and relevant bond design parameters for the three types ofcolumns is provided in Table 1. In each column, the longitudinalreinforcement was lap spliced at the base with starter (dowel) barsof the same diameter, projecting above the footing. The splicelength for all columns was 30db in accordance with the ACI build-ing code (2008) requirements for gravity load design, or the 1977version of the same code for seismic design.

When subjected to large drift reversals, all of the as-built col-umns suffered splitting bond failure of the starter bars and extensiveconcrete damage within the splice zone (Harajli and Dagher 2008).This damage occurred without buckling of the starter bars or rup-turing of the transverse steel ties. Relevant discussion of the testresults of the as-built columns is presented in comparison withthe results of the repaired/strengthened columns in the subsequentsections.

Repaired/Strengthened Columns

Repair ProcedureAs-built columns C14, C16, and three identical column specimensdesignated C20 were repaired and/or strengthened for improvingtheir seismic resistance. The repair procedure consisted of remov-ing both the damaged and undamaged concrete completely withinthe splice zone and casting new concrete. Notice that in real col-umns, removing all the concrete within the splice zone may not bepossible, but it was done in the current laboratory study only for thepurpose of facilitating the construction of the test specimens. Noadhesive was used to bond the old concrete to the new concrete.The repair procedure is described schematically in Fig. 2.

Design and Application of the Strengthening SystemThe strengthening system consisted of providing confinementwithin the splice zone composed of internal steel ties, externalFRP jackets, or a combination of both. The steel ties weregrade 60 ( f yt ¼ 415 MPa) of 10 mm diameter. The FRP re-inforcement consisted of commercially available unidirectionalCFRP sheets having the following design properties provided

30d b

h o=

140

0

50

200

500

8T16

42

42

Note : All dimensionsare in mm

P

Rep

air

Zon

e

φ8@200

C16

400

50

Starter Bars

1200

200

400 6T20

30

30

C20

Sections @ Column Base

StarterBars

200

8T14

27

27

400

C14

φ8@200 (C14, C16, C20)φ8@200 (C14-S20-FP2, C16-S20-FP1, C20-S20-FP2)φ10@50 (C14-S5, C16-S5, C20-S5)φ10@100 (C20-S10-FP1)None (C20-FP3)

West East

100

Fig. 1. Dimensions and reinforcement details of the column specimens

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by the manufacturer: thickness per application (including resinbinder) tf ¼ 0:13 mm; modulus of elasticity Ef ¼ 230;000 MPa;fracture tensile strength ¼ 3;500 MPa; and strain at tensile frac-ture of the fibers ¼ 1:5%.

The areas of transverse confinement (steel, FRP, or a combina-tion of both) within the splice zone required for effective seismicbond strengthening were estimated in accordance with each ofEqs. (5)–(7), respectively. Results are provided in Table 2 alongwith the actually used areas for the various specimens. Table 2(and also Table 1) includes data for three additional column spec-imens that were tested previously (Harajli and Dagher 2008). These

columns, designated C14-S20-FP2, C16-S20-FP1, and C20-S20-FP2, are identical to the as-built column specimens (C14, C16,and C20), except that they were bond strengthened before testingusing external FRP jackets of the same type used in the currentstudy. These specimens are referred to as “strengthened” columns(Tables 1 and 2) to differentiate them from the “repaired” columnstested in the current investigation. In using Eq. (7) for estimatingthe design area of confining reinforcement, because the value of Kdiffers depending on the position of the spliced bar relative to thetransverse tie or stirrup [Eq. (3)], a weighted average value of theparameter K of 0.083, 0.075, and 0.1 was used for the columns inseries C14, C16, and C20, respectively.

The design areas (or thicknesses) of confining reinforcementprovided in Table 2 were calculated using the actual concrete com-pressive strength given in Table 1 for two different values of splicestress f s: f s ¼ 1:25f yðactualÞ where f yðactualÞ is given for the differentspecimens in Table 1, and f s ¼ 1:85f yðdesignÞ where f yðdesignÞ is takenequal to 415 MPa (grade 60). For the steel-FRP confined columns,the design thickness of the FRP jacket was calculated taking intoaccount the area of internal steel confinement Atr=s provided alongthe splice length of 1:57 mm2=mm (ϕ10@100 mm) in columnC20-S10-FP1, and 0:5 mm2=mm (ϕ8@200 mm) in columns C14-S20-FP2, C16-S20-FP1, and C20-S20-FP2, respectively.

It can be observed from the results in Table 2 that Eqs. (5) and(7) overestimate the effect of steel confinement on bond strength,thus predicting relatively smaller design transverse steel areas for

Wood Form for casting newconcrete

Top View of Specimen During Repair/Strengthening

Fractured concrete totally removed, 10 mm transverse reinforcement placed (as applicable), and new concrete added

Old concrete Zone where concrete previously fractured by bond splittting due to cyclic loading

Cold joint

30d b

Fig. 2. Reinforcement details of the repaired column specimens

Table 1. Summary of Test Parameters and Test Data

Column category Specimen Col. reinf. As c=dba Ls=db

f 0c(MPa)b

f y(MPa)

Area of transverse steel(within the splice zone)

Thickness of FRPjacket nf tf (mm)

Average peakload (kN)

As-built columns C14 8ϕ14 1.4 30 39 550 ϕ8@200 — 79.0

C16 8ϕ16 2.0 30 40 528 ϕ8@200 — 96.6

C20 6ϕ20 1.0 30 32 617 ϕ8@200 — 87.3

Strengthened

columns

C14-S20-FP2 8ϕ14 1.4 30 39 550 ϕ8@200 0.26 91.5

C16-S20-FP1 8ϕ16 2.0 30 40 528 ϕ8@200 0.13 110.0

C20-S20-FP2 6ϕ20 1.0 30 32 617 ϕ8@200 0.26 140.0

Repaired columns C14-S5 8ϕ14 1.4 30 45 550 ϕ10@50 None 97.4

C16-S5 8ϕ16 2.0 30 45 528 ϕ10@50 None 107.3

C20-S5 6ϕ20 1.0 30 45 617 ϕ10@50 None 142.0

C20-S10-FP1 6ϕ20 1.0 30 45 617 ϕ10@100 0.13 129.6

C20-FP3 6ϕ20 1.0 30 45 617 None 0.39 145.8ac ¼ cm in Eqs. (2) and (6).bWithin the splice zone for the repaired columns.

Table 2. Estimated Design Area of Steel or FRP Confinement

Column category Specimen nsa

Confinementbeing calculated

Design Atr=s or nf tf (mm) Actually usedAtr=s or nf tf (mm)OJB Darwin CEB-FIP

Strengthened

columns

C14-S20-FP2 3 FRPb 0.15d/0.23e 0.24d/0.31e 0.17d/0.25e 0.26 (2 layers)

C16-S20-FP1 4 FRPb 0:06=0:22 0:16=0:32 0:18=0:34 0.13 (1 layer)

C20-S20-FP2 2 FRPb 0:36=0:36 0:40=0:40 0:37=0:37 0.26 (2 layers)

Repaired

columns

C14-S5 3 Steel 1:12=1:54 2:77=3:59 1:21=1:56 3.14 (ϕ10@50)

C16-S5 4 Steel 0:65=1:38 2:25=3:80 1:08=1:73 3.14 (ϕ10@50)

C20-S5 2 Steel 1:88=1:88 3:19=3:19 2:02=2:02 3.14 (ϕ10@50)

C20-S10-FP1 2 FRPc 0:05=0:05 0:18=0:18 0:08=0:08 0.13 (1 layer)

C20-FP3 2 FRP 0:33=0:33 0:36=0:36 0:25=0:25 0.39 (3 layers)aEqual to the number of spliced bars in tension.bTaking into account a provided area of steel reinforcement Atr=s ¼ 0:5 mm (ϕ8@200).cTaking into account a provided area of steel reinforcement Atr=s ¼ 1:57 mm (ϕ10@100).dBased on f y ¼ 1:25½f yðactualÞ�.eBased on f y ¼ 1:85½f yðdesignÞ ¼ 415 MPa� (Hawkins et al. 2000).

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the steel-confined specimens when compared with the predictionsof Eq. (6). Consequently, the design thicknesses of the FRP jacketcalculated using Eqs. (5) and (7) when FRP is used in combinationwith internal steel ties (columns C20-S10-FP1, C14-S20-FP2, C16-S20-FP1, and C20-S20-FP2) are also smaller than those calculatedusing Eq. (6) as would be expected. On the other hand, for columnC20-FP3, for which the confinement is composed exclusively ofFRP, the design thicknesses of the FRP jacket estimated usingEqs. (5) and (6) are approximately identical, but slightly larger thanthat estimated using Eq. (7). Notice that the difference in the esti-mated areas of transverse confinement (steel or FRP) between de-veloping a design splice stress f s ¼ 1:25f yðactualÞ and a stress f s ¼1:85f yðdesignÞ for the specimens in series C14 and C16 is attributedto the difference between the corresponding two magnitudes of thesplice stress. On the other hand, for the specimens in series C20, themagnitude of 1:25f yðactualÞ of 770 MPa is practically equal to that of1:85f yðdesignÞ of 768 MPa, and hence the amounts of confiningreinforcement estimated for the two respective magnitudes of f sare practically identical.

Furthermore, it can be seen in Table 2 that, with the exception ofspecimen C20-S20-FP2 and to some extent C20-S10-FP1, the areasof confining reinforcement provided are more or less consistentwith the design areas estimated using Eq. (6) for developing asplice stress f s ¼ 1:25f yðactualÞ. It should be mentioned that the cal-culated as well as the actually used areas of transverse steelreinforcement of ϕ10@50 (3:14 mm2=mm) for the steel-confinedcolumns (C14-S5, C16-S5, and C20-S5) are approximately doublethe minimum area of transverse reinforcement along the loadingdirection of about 1:6 mm2=mm required at the base of the columnsin regions of high seismic hazard, estimated using Eqs. 21-4 and21-5 of the ACI building code (2008).

All repaired specimens were cast together using one batchof Ready-Mix concrete. The concrete mix consisted of portlandcement type I, crushed limestone aggregate, and natural sand withaggregate∶sand∶cement proportions by weight of 1∶0:73∶0:35 andwater-cement ratio of 0.40. The cylindrical strength of the addedconcrete was measured at 45 MPa. The CFRP in all strengthenedand repaired specimens consisted of 600-mm-wide flexible sheetsapplied around the perimeter of the column section covering thefull height of the splice zone (Fig. 1) with fiber orientation alongthe circumference.

Loading Procedure and Measurements

All specimens were tested using a reaction wall for applying thelateral load at the top of the column (Fig. 1) and a strong floorfor supporting the column footing. The specimens were subjectedto the loading protocol given in Fig. 3. The actuator load, whichwas applied at 1.4 m above the column base, consisted of a se-quence of displacement-controlled cycles for achieving target drift

ratios DR (DR = ratio of lateral drift to one-half column height orshear span ho of 1.4 m) varying between a minimum of about 0.5%and a maximum of about 6.0%. Notice that since the bond strengthof the spliced column bars is governed by the splice tensionstresses, no axial load was applied on the column as it was deemedto have little influence on the outcome of this study. Test measure-ments included applied actuator load, lateral drift, strain in theoutermost column reinforcement (starter bars) at the column base,and transverse strain in the FRP jacket. The FRP strain wasmeasured using four electric strain gauges mounted at 200 mmabove the column base. The test and test measurements were moni-tored and controlled automatically using a data acquisition andcontrol unit.

Experimental Results

Failure Mode

Failure modes of the repaired/strengthened columns in comparisonwith the as-built ones are shown in Figs. 4 and 5, respectively.Because of the inferior bond capacity of the spliced bars associatedwith small splice length and/or small concrete covers, and moreimportantly inadequate area of transverse steel confinement, allthe as-built columns experienced splitting bond failure of the starterbars before the bars were able to develop their yield strength. Bondfailure occurred in columns C14 and C20 earlier in the responsewhen compared to column C16 because of their relatively smallerratio of concrete cover to bar diameter. Bond splitting was accom-panied with considerable damage within the splice zone, quickbond degradation, and loss of lateral load resistance.

All repaired/strengthened columns developed yielding of thestarter bars at peak lateral load and significantly less concretedamage within the splice region when compared with the as-builtcolumns. Most of the damage in columns C14-S5 and C16-S5concentrated within 50 mm to 100 mm above the column base.Bottom splitting cracks parallel to the outermost spliced bars wereobserved in these columns at large drift ratios, but they were muchless adverse than those developed in the as-built columns. On theother hand, because of the relatively small ratio of concrete cover tobar diameter, column C20-S5 experienced clear side and bottomsplitting cracks near the outermost spliced bars, resulting in a

-8

-6

-4

-2

0

2

4

6

8

0 4 8 12 16 20 24Cycle No.

Drif

t Rat

io (

DR

), %

Fig. 3. Load history of the column specimens Fig. 4. Failure modes of the columns in series C14

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relatively larger concrete damage within the column hinging regionwhen compared to columns C14-S5 and C16-S5, particularly afteryielding of the starter bars (Fig. 5). Damage in all FRP confinedcolumns (strengthened and repaired) concentrated near the columnbase where small bands of the FRP sheets tended to fracture and thedeformation to concentrate at one major crack.

Based on the experimental observation, it can be concluded thatall repaired/strengthened columns developed full flexural capacityand hence failed in a predominantly flexural mode, combined withcyclic bond degradation at large drift reversals.

Load-Splice Strain Response

Figs. 6 and 7 show typical variations of the applied lateral loadversus strain in the outermost starter bars measured at column base.Unfortunately, the steel (and FRP) strain data for column C20-S10-FP1 was not recoded owing to some measurement error.

The maximum splice strains developed in the as-built columnsbefore bond failure on the east and the west sides of the columns(Fig. 1) were 1;400 με and 1;900 με for column C14, 2;300 με and3;080 με (partial yielding) for C16, and 700 με and 1;100 μεfor C20, respectively, which are sizably lower than the yieldstrain, particularly for columns C14 and C20. On the other hand,the strengthened/repaired columns developed considerably largersplice strains, beyond yield, as originally anticipated in designingthe strengthening system. The corresponding strains on the east-west sides of the columns (developed before strain gauge damage)were, respectively, 5;350 με� 9;500 με for C14-S5, 3;300 με�8;500 με for C16-S5, 5;500 με� 12;000 με for C20-S5,6;600 με� 10;300 με for C20-FP3, 2;500 με (damaged)-5;380 μεfor C14-S20-FP2, 5;950 με� 4;200 με for C16-S20-FP1, and7;300 με� 17;000 με for C20-S20-FP2. Notice that the measuredsplice strains on the west side of the columns were generally largerthan those developed on the east side because the first flexuralcracks in all specimens occurred at the west side.

FRP Strain/Stress

A typical response of FRP strains with applied lateral load is shownfor column C20-FP3 at different drift ratios DR in Fig. 8. Resultsare given for the average of two strain gauges mounted on eitherside of the column in the position shown in the caption of Fig. 8.

The mechanism by which the FRP strains increase with lateralapplied load when used for splice bond strengthening in rectangularor circular column sections has been described in previous studies(Harajli 2008). At low drift ratios, before the initiation of splittingcracks, the increases in FRP strains or stresses with an increase inlateral drift or applied load are relatively small. As soon as bondsplitting occurs, in the process of restraining the width of the split-ting cracks, the FRP strains/stresses increase rapidly with appliedload. Unloading from a maximum drift causes the FRP strain tonaturally decrease but then to increase again slightly owing tothe dilation of the concrete in compression. The residual FRP strainat zero applied load in any one cycle after bond splitting is attrib-uted to the incomplete closure of the splitting cracks owing to

Fig. 5. Failure mode of the columns in series C20 -120

-90

-60

-30

0

30

60

90

120

-2500 0 2500 5000 7500 10000

Splice Strain (µε)

Late

ral L

oad

P (

kN)

C14-S5

-120

-90

-60

-30

0

30

60

90

120

-2500 0 2500 5000 7500 10000

Splice Strain (µε)

Late

ral L

oad

P (

kN)

Specimen C14

East Side

West Side

-120

-90

-60

-30

0

30

60

90

120

-2500 0 2500 5000 7500 10000

Splice Strain (µε)

Late

ral L

oad

P (

kN)

C14-S20-FP2

East Side

West Side

Fig. 6. Applied load versus splice strain for the as-built and repaired/strengthened columns in series C14

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the roughness of the concrete surfaces on either side of the cracks.Notice that the maximum measured FRP strain for column C20-FP3 of about 1;500 με (Fig. 8) and also those of the remainingFRP strengthened specimens of 200 με (C14-S20-FP2), 220 με(C16-S20-FP1), and 940 με (C20-S20-FP2), respectively, are con-siderably lower than the ultimate tensile strain of the CFRP used(of 15;000 με).

Load-Drift Response

Figs. 9–11 show the cyclic response of the lateral applied load(column shear) versus drift ratio, and Table 1 summarizes the aver-age peak envelope lateral load (average between the compressionand tension cycles) for all specimens in series C14, C16, and C20,respectively. Fig. 12 compares the energy dissipated per cycle atdifferent cycle number and drift ratio for the various column spec-imens. The energy (kN-m) was calculated from the area underthe lateral load (kN)–drift (m) response enclosed within one

complete cycle. Cycles 5, 8, 11,…, 20 correspond to the first cyclein the target drift ratios of 1%, 2%, 3%, and 6%, respectively.

Bond failure of the as-built columns, which generally developedbefore yielding of the spliced bars, occurred at drift ratios of about1.0% for columns C14 and C20, and about 1.2% for column C16.Bond failure was followed by quick bond degradation and lossin lateral load capacity. Notice that the area of transverse steel con-finement Atr=s in the as-built columns, of 0:5 mm2=mm (ϕ8@200),is sizably less than that estimated in Table 2 for “adequate” seismicbond strengthening, particularly for columns C14 and C20. This ineffect justifies the premature bond failure of the spliced bars and theconsequent inferior seismic performance of these columns.

Comparing the cyclic responses in Figs. 9–11, it is clear thatthe strengthened/repaired columns developed considerably largerloads and experienced much less strength and stiffness degrada-tions than the as-built columns. The confining reinforcement inthe repaired/strengthened columns limited the width or growth ofthe splitting cracks, leading to enhanced bond capacity and yieldingof the spliced reinforcement, as initially planned when designingthe strengthening system. The drift ratio at which yielding (of theoutermost starter bars) in the strengthened columns developed(taken as the average between the compression and tension cycles)was 0.85% for column C14-S20-FP2, 1.2% for C16-S20-FP1, and1.4% for C20-S20-FP2. For the repaired columns, the drift ratiosat which yielding occurred were about 0.9%, 1.3%, and 2.0% forcolumns C14-S5, C16-S5, and C20-S5; and 1.5% and 1.2%for columns C20-S10-FP1 and C20-FP3, respectively. Because ofthe improved bond strength, yielding of the spliced reinforcementwas followed by reasonably little flexural strength and stiffnessdegradations up to a sizable drift ratio of about 3.0% to 4.0%.

-150

-120

-90

-60

-30

0

30

60

90

120

150

-2500 0 2500 5000 7500 10000Splice Strain (µε)

Late

ral L

oad

P (

kN)

C20-S5

-150

-120

-90

-60

-30

0

30

60

90

120

150

-2500 0 2500 5000 7500 10000

Splice Strain (µε)Late

ral L

oad

P (

kN)

C20-FP3

-150

-120

-90

-60

-30

0

30

60

90

120

150

-2500 0 2500 5000 7500 10000Splice Strain (µε)

Late

ral L

oad

P (

kN)

Specimen C20

West Side

East Side

Fig. 7. Applied load versus splice strain for the as-built and repaired/strengthened columns in series C20

-150

-100

-50

0

50

100

150

-500 0 500 1000 1500 2000

FRP Strain (µε)Late

ral L

oad

P (

kN)

(b)

3.0% 4.0%

-150

-100

-50

0

50

100

150

-500 0 500 1000 1500 2000

FRP Strain (µε)Late

ral L

oad

P (

kN)

C20-FP3

(a)2.0% 3.0% DR = 4.0%

Fig. 8. Applied load versus average FRP strain: (a) west side ofcolumn, (b) east side of column

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The increase in load capacity of the strengthened columns incomparison with the as-built ones (see Table 1) was about 16%(C14-S20-FP2), 14% (C16-S20-FP1), and 60% (C20-S20-FP2).For the repaired columns, the corresponding increases were 23%for C14-S5, 11.0% for C16-S5, and about 63%, 45%, and 67%for C20-S5, C20-S10-FP1, and C20-FP3, respectively. Notice thatbecause the lateral load capacity of the columns is capped by steelyielding, the higher the steel strain/stress developed in the splicedbars in the as-built columns before bond failure, or the closer it is tothe yield strain/strength (owing to a relatively high c=db), the lowerwould be the expected incremental increase in the splice stressand, accordingly, the lateral load capacity owing to the addedconfinement. This justifies why the repaired/strengthened columnsin series C16 having a relatively high c=db of 2.0 developed thelowest strength gain while, on the other extreme, those in seriesC20, having the lowest c=db of 1.0 developed the highest strengthgain.

It can be observed from the load-drift responses in Figs. 9–11that the load resistance of the repaired/strengthened columns inseries C20 experienced a larger drop at drift ratios in excess of

3.5 to 4.0% when compared with the columns in series C14and C16. Consequently, the energy absorption and dissipationcapacities of the repaired and strengthened columns in seriesC20 at drift ratios in excess of about 3.5% were generally less thanthose in series C14 and C16 (Fig. 12). This behavior is attributed inone part to the quicker cyclic bond degradation associated with therelatively low ratio of concrete cover to bar diameter (of 1.0) in thecolumns of series C20 when compared to the columns in the otherseries, and in another part to the lower than “required” area ofconfining reinforcement [estimated in accordance with Eq. (6)]provided in two of these columns, namely, C20-S20-FP2 andC20-S10-FP1 (Table 2).

It should be indicated that while the FRP confined columnsperformed approximately similar to steel-confined columns in thisinvestigation, the FRP confinement limited the concrete damagewithin the critical hinging region of the column more effectivelythan steel confinement. This low concrete damage associated withthe use of external FRP confinement is attributed to the fact that theFRP wraps prevents the concrete cover from breaking off underlarge concrete strains and the column longitudinal reinforcementfrom axial buckling under large drift reversals.

-100

-75

-50

-25

0

25

50

75

100

-6 -4 -2 0 2 4 6Drift Ratio (%)

Column Shear (kN)

C14Analysis

(a)AssumingNo Bond Failure

-100

-75

-50

-25

0

25

50

75

100

-6 -4 -2 0 2 4 6Drift Ratio (%)

Column Shear(kN)

C14-S5Analysis

(b)

-100

-75

-50

-25

0

25

50

75

100

-6 -4 -2 0 2 4 6Drift Ratio (%)

Column Shear (kN)

C14-S20-FP2Analysis

(c)

Fig. 9. Shear force versus lateral drift ratio of columns C14:(a) as-built, (b) repaired, (c) strengthened

-120

-80

-40

0

40

80

120

-6 -4 -2 0 2 4 6Drift Ratio (%)

Column Shear(kN)

C16Analysis

(a)

AssumingNo Bond Failure

-120

-80

-40

0

40

80

120

-6 -4 -2 0 2 4 6Drift Ratio (%)

Column Shear(kN)

C16-S5Analysis

(b)

-120

-80

-40

0

40

80

120

-6 -4 -2 0 2 4 6Drift Ratio (%)

Column Shear(kN)

C16-S20-FP1Analysis

(c)

Fig. 10. Shear force versus later drift ratio of columns C16: (a) as-built,(b) repaired, (c) strengthened

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Proposed Expression for Seismic BondStrengthening

Based on the results of this study, it is proposed to adopt Eq. (6) forseismic bond strengthening of spliced bars whether for existingstructures using external FRP wraps or for damaged structuresusing internal steel ties, or FRP wraps, or a combination of both.This endorsement of Eq. (6) is justified directly or indirectly bythree major observations derived from the experimental resultsof this study. First, the area of confining reinforcement Atr=s, of0:5 mm2=mm (ϕ8@200), provided in the as-built columns is sig-nificantly less than that estimated using Eq. (6) for “adequate” bondstrengthening. This justifies why these columns developed prema-ture bond failure and quick bond and flexural strength degradation,particularly columns C14 and C20 (Figs. 9 and 11). Second, theareas of confining reinforcement, which were provided in thestrengthened/repaired columns (except C20-S20-FP2 and C20-S10-FP1) in accordance with the design areas estimated using

-150

-100

-50

0

50

100

150

-6 -4 -2 0 2 4 6Drift Ratio (%)

Column Shear(kN)

C20Analysis

(a)

AssumingNo Bond Failure

-150

-100

-50

0

50

100

150

-6 -4 -2 0 2 4 6Drift Ratio (%)

Column Shear (kN)

C20-S5Analysis

(b)

-150

-100

-50

0

50

100

150

-6 -4 -2 0 2 4 6Drift Ratio (%)

Column Shear (kN)

C20-FP3

Analysis

(c)

-150

-100

-50

0

50

100

150

-6 -4 -2 0 2 4 6Drift Ratio (%)

Column Shear(kN)

C20-S10-FP1Analysis

(d)

-150

-100

-50

0

50

100

150

-6 -4 -2 0 2 4 6Drift Ratio (%)

Column Shear (kN)

C20-S20-FP2

Analysis

(e)

Fig. 11. Shear force versus later drift ratio of columns C20: (a) as-built;(b), (c), (d) repaired; (e) strengthened

0.0

2.0

4.0

6.0

8.0

10.0

0 2 4 6 8 10 12 14 16 18 20 22 24Cycle Number

Ene

rgy/

Cyc

le (

kN-m

)

C14C14-S5C14-S20-FP2

DR = 1%

6%

5%

3%

2%

4%

0.0

2.0

4.0

6.0

8.0

10.0

12.0

0 2 4 6 8 10 12 14 16 18 20 22 24Cycle Number

Ene

rgy/

Cyc

le (

kN-m

)

C16C16-S5C16-S20-FP1

DR = 1%

6%5%

4%

3%

2%

0.0

2.0

4.0

6.0

8.0

10.0

0 2 4 6 8 10 12 14 16 18 20 22 24Cycle Number

Ene

rgy/

Cyc

le (

kN-m

) C20C20-S5C20-FP3C20-S10-FP1C20-S20-FP2

DR = 1%

5%4%

3%

2%

6%

Fig. 12. Variation of energy dissipated with number of cycles and driftratio

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Eq. (6) for f s ¼ 1:25f yðactualÞ, resulted in satisfactory seismic behav-ior of the columns, as illustrated previously. Third, the areas of con-fining reinforcement in columns C20-S20-FP2 and C20-S10-FP1were less than those estimated using Eq. (6), and therefore theseparticular column specimens acquired less average strength gain(Table 1) and lower energy dissipation capacity than their com-panion specimens in series C20 (Fig. 12). Notice that in adoptingEq. (6), a more conservative design area of confinement reinforce-ment (steel or FRP) can still be estimated based on developinga splice stress f s equal to the larger of 1:25f yðactualÞ as recommendedby Harajli and Khalil (2008) and 1:85f yðdesignÞ as recommended byHawkins et al. (2000).

Prediction of the Envelope Response

The adequacy of the repair and strengthening method used in thisinvestigation was further verified by analytically reproducing thebackbone load-deformation response of the various columns.The analysis is conducted assuming that the columns’ longitudinalreinforcing bars are continuous (as opposed to spliced) and the col-umns are capable of developing full plastic hinges at their base,which is identical to the nonlinear flexural analysis used previously(Harajli 2008) for reproducing the response of the as-built columns(confined and unconfined). The analysis uses constitutive modelsof the stress-strain relationships of the steel and concrete for gen-erating the moment-curvature (M � ϕ) relationship of the columnsection within the plastic region. For the steel reinforcing bars, abilinear stress-strain relationship having elastic modulus Es ¼200;000 MPa and strain hardening modulus Esh is used. The mag-nitude of the hardening modulus was estimated using the actualstress-strain curve of the steel bars at 1,050, 550, and 1,100 MPafor the 14, 16, and 20 mm bars, respectively. The stress-strainbehavior of confined concrete was simulated using a general designoriented stress-strain ( f cc � εcc) relationship derived recently byHarajli et al. (2006), which is applicable for concrete confined withsteel ties, FRP jackets, or a combination of both. Stress-strainresponses of columns C20, C20-S5, C20-S10-FP1, and C20-FP3of the current investigation, predicted using this relationship, areshown for illustration in Fig. 13. This relationship is composedof two stages. In the first stage (0 ≤ εcc ≤ εco), the stress-strainresponse is expressed as

f cc ¼ f co

�2εccεco

��εccεco

�2�

ð8Þ

where f co and εco = peak stress and corresponding strain in the firststage of the response, which also represents the intersection pointbetween the first and second stages (Fig. 13). The stress-strain re-sponses in the second stage (εcc ≥ εco), including the values of f coand εco at the intersection point between the first and second stages[to be used with Eq. (8)] corresponding to εℓ ¼ εℓo ¼ 0:002,εcc ¼ εco, and f cc ¼ f co, are evaluated using the following generalexpressions:

f cc ¼ f 0c

�1þ 1:25

�f ℓf þ f ℓsAcc=Ag

f 0c

�1=2

�ð9Þ

εcc ¼ εo�1þ

�25800e1:17h=b

ðρf Ef Þ0:83εℓ þ 2:0

��f ccf 0c

� 1

��ð10Þ

where the value of f cc=f 0c in Eqs. (9) and (10) shall not be takenless than ½1þ 2:0ðf ℓf þ f ℓsÞ=f 0c� nor greater than ½1þ 7:0ðf ℓfþf ℓsÞ=f 0c�, where

ρf ¼4nf tfD

ð11Þ

f ℓf ¼�kef kvf ρf Ef

2

�εℓ ð12Þ

f ℓs ¼�keskvsρst

2

�f yt ð13Þ

The term εℓ = lateral FRP strain; εo = strain at peak stress forunconfined concrete, taken equal to 0.002; h and b = depth andwidth of the rectangular section; D = diameter for circular columnsections, while for rectangular sections it is calculated as D ¼2bh=ðbþ hÞ in accordance with the 2002 report of ACI Committee440; Acc = area of the steel-confined concrete core, and Ag is thearea of the gross section. The terms ke (kef or kes) and kv (kvf or kvs)account for the effectiveness of the lateral reinforcement in confin-ing the concrete along the horizontal plane, and the concrete be-tween transverse ties or FRP strips, respectively. Expressions forke and kv are provided in the 2002 report of ACI Committee440 (or Harajli et al. 2006). For circular sections, the ratio h=bin Eq. (10) is set equal to 1.0.

The stress-strain curve in the second stage can be generated byincrementally increasing εℓ in Eq. (12) beyond εℓ ¼ 0:002, calcu-lating f cc for each respective value of εℓ from Eq. (9), and thencalculating the corresponding strain εcc from Eq. (10). Note thatfor unconfined concrete or steel-confined concrete in which thesecond stage (εcc ≥ εco) is known to be a descending stage,the stress-strain curve is generated using the second stage of thestress-strain relationship developed by Scott et al. (1982):

f cc ¼ f co½1� Zðεcc � εcoÞ� ≥ 0:2f co ð14Þ

Z ¼ 0:5

3þ 0:29f 0c145f 0c � 1000

þ 34ρst

ffiffiffiffiffih00

Sh

s� εco

ð15Þ

where h00 = width of concrete core measured to the outside of theperipheral tie; Sh = center-to-center spacing of ties. εco and f coare calculated using Eqs. (9) and (10) for εℓ ¼ εℓo ¼ 0:002, andρf ¼ 0:0. For unconfined concrete εco and f co in Eq. (8) becomeequal to εo and f 0c, respectively.

0

10

20

30

40

50

60

70

0 0.003 0.006 0.009 0.012 0.015εcc

f cc (

MP

a)

C20C20-S5C20-S10-FP1C20-FP3

(f co , ε co )

Fig. 13. Typical predicted stress-strain behavior for unconfined andsteel, FRP, or steel-FRP confined concrete

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The envelope lateral load-drift response, which was generatedfrom the envelopeM � ϕ relationship and the concept of equivalentplastic hinge length (Harajli 2008), is superimposed on the mea-sured backbone response for the various column specimens inFigs. 9–11. The discrepancy between the analytical and experimen-tal responses for the as-built columns, particularly C14 and C20 isattributed to the premature bond failure of these columns that is notaccounted for in the analysis. On the other hand, the comparisonsfor the repaired/strengthened columns clearly show good agree-ment between the experimental and analytical results, particularlyfor drift ratios less than about 3.0 to 4.0%. The nonlinear analysispredicted reasonably well the initial stiffness of the load-driftresponse, the lateral load at which first yielding occurred, andthe peak load capacity of the strengthened columns.

Conclusions

Based on the results of this investigation, the following conclusionscan be drawn:1. Gravity load-designed (as-built) columns in which the column

reinforcement is spliced within the critical hinging region at thecolumn base are expected to develop splitting bond failurebefore steel yielding when subjected to lateral drift reversalsinduced by strong earthquakes. Splitting bond failure leads toquick stiffness and strength degradations under cyclic loadingand considerable concrete damage within the splice region.This behavior is particularly true for columns with small ratiosof concrete cover to bar diameter.

2. Eq. (6) is recommended for designing the area of confiningreinforcement (steel, FRP, or a combination of both) requiredfor adequate seismic bond strengthening of existing unda-maged columns (using FRP) or for repairing/strengtheningdamaged ones. In using this equation, it is suggested thatone evaluate the area of confining reinforcement based onthe development of a steel stress f s equal to the larger of1:25f yðactualÞ and 1:85f yðdesignÞ.

3. Repair/strengthening of the as-built columns of the current in-vestigation by providing an area of confining reinforcement(steel, FRP, or a combination of both) within the spliced zone,determined in accordance with Eq. (6), allowed the spliced barsto develop sizable postyield strains and the columns to acquireconsiderably larger lateral loads and energy dissipation capa-cities and to experience much less damage than the as-builtcolumns.

4. In further support of the validity of the design strengtheningapproach, the nonlinear analysis conducted assuming a perfectbond between the columns’ longitudinal steel bars and con-crete reproduced accurately the envelope (backbone) load-driftresponses of the repaired/strengthened columns.

Acknowledgments

This research was supported by the Lebanese National Council forScientific Research (LNCSR). The writer is most grateful for thissupport and to the Faculty of Engineering and Architecture at the

American University of Beirut (AUB) for providing the laboratoryfacilities.

References

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Harajli, M. H., Hamad, B. H., and Rteil, A. (2004). “Effect of confinementon bond strength between steel bars and concrete.” ACI Struct. J.,101(5), 595–603.

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