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SEISMIC RESILIENCE OF TALL BUILDINGS - BENCHMARKING
PERFORMANCE AND QUANTIFYING IMPROVEMENTS
A THESIS
SUBMITTED TO THE DEPARTMENT OF CIVIL AND ENVIRONMENTAL ENGINEERING
AND THE COMMITTEE ON GRADUATE STUDIES
OF STANFORD UNIVERSITY
IN PARTIAL FULFILLMENT OF THE REQUIREMENTS
FOR THE DEGREE OF
ENGINEER
JENNISIE TIPLER
DECEMBER 2014
http://creativecommons.org/licenses/by-nc/3.0/us/
This dissertation is online at: http://purl.stanford.edu/xh842sm8488
© 2014 by Jennisie Frances Tipler. All Rights Reserved.
Re-distributed by Stanford University under license with the author.
This work is licensed under a Creative Commons Attribution-Noncommercial 3.0 United States License.
ii
Approved for the department.
Gregory Deierlein, Adviser
Approved for the Stanford University Committee on Graduate Studies.
Patricia J. Gumport, Vice Provost Graduate Education
This signature page was generated electronically upon submission of this thesis in electronicformat. An original signed hard copy of the signature page is on file in University Archives.
iii
1
SEISMIC RESILIENCE OF TALL BUILDINGS - BENCHMARKING PERFORMANCE AND
QUANTIFYING IMPROVEMENTS
ABSTRACT
Modern tall buildings are generally not considered to be a large contributor to the seismic risk of
cities, based on the presumption that they are designed and built with sufficient safeguards to
ensure good performance. This is in spite of the fact that current building code provisions have few,
if any, provisions to ensure that tall buildings have better performance than other low-rise
structures. This implies that a 40-story building is not expected, or designed, to perform any better
than a one-story building following a large seismic event, despite the huge differences in the
consequences of collapse and/or damage to these type of structures. The performance of a 42-
story couple core wall building located in downtown San Francisco, designed using a state-of-the-
practice performance-based approach, is evaluated. Two additional structural schemes, damped
outriggers and base isolation, and one additional non-structural scheme are investigated. Non-linear
response history analysis is conducted on each of the three structural building designs in order to
assess the structural performance at five different seismic hazard levels. Subsequently, the expected
building repair cost and downtime are estimated for each scheme; there are six schemes in total
when considering the additional non-structural design scheme.
The baseline building is expected to suffer financial losses exceeding 15% of the total building cost
and functional downtime of almost 2 years (84 weeks) following a design level earthquake. The
damped outrigger and base isolation schemes are found to reduce financial losses and downtime,
with an expected loss of 14% and 10% of the building cost, respectively and an expected functional
downtime of 62 weeks and 43 weeks, respectively following a design-level earthquake. The non-
structural design alternative, which also includes provisions to reduce building downtime, was found
2
to reduce loss and downtime in all cases. The best performing building is the base-isolated building
with enhanced non-structural design, expected to experience losses of 2.4% of the building value
and functional downtime of only 6 weeks following a design-level earthquake.
A cost-benefit analysis reveals that all schemes are preferable to the baseline building. The payback
period for the two structural design alternatives is found to be 4.6 years and 6.6 years for the
damped outrigger and the base isolation schemes, respectively, and the payback period for the
non-structural design alternatives are 5.3 years, 9.0 years and 8.7 years for the fixed base, damped
outrigger and the base isolation schemes, respectively.
3
ACKNOWLEDGEMENTS
The author would like to sincerely thank her supervisor, Professor Greg Deierlein, for his support,
patience and incredible wealth of knowledge. She also acknowledges Ibrahim Almufti for
collaborating with her on this work and thanks Ibrahim for his mentoring, sharing of ideas, donating
his time and, most importantly, sharing his passion for resilient building design. Several other people
who have contributed significantly to this work are Professor Eduardo Miranda, Michael Willford,
Sean Merrifield and Brian Carey.
Financial support for this research was provided by the John. A. Blume Earthquake Engineering
Center, the Fulbright Foundation and National Science Foundation Grant CMMI-NEES Grant No.
1135029.
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TABLE OF CONTENTS
Abstract ........................................................................................................................................... 1 Acknowledgements ......................................................................................................................... 3 Table of Contents ............................................................................................................................ 4 List of Tables ................................................................................................................................... 6 List of Illustrations ............................................................................................................................ 8 1 Introduction ............................................................................................................................. 10
1.1 The need to Quantify Seismic Resilience of Tall Buildings ............................................. 12 1.2 Previous Studies of Tall Buildings in Los Angeles ......................................................... 13 1.3 Overview of the REDi Rating System ............................................................................ 15
2 Building Design ........................................................................................................................ 17 2.1 Fixed Base Building Design .......................................................................................... 17 2.2 Damped Outrigger Building Design .............................................................................. 26 2.3 Base-isolated Building Design ...................................................................................... 27 2.4 Design of Non-Structural components ......................................................................... 29 2.5 Summary of Building Costs .......................................................................................... 32
3 NLRHA .................................................................................................................................... 35 3.1 Ground Motion Selection and Scaling ............................................................................... 35 3.2 Non-Linear Building Model ............................................................................................... 37 3.3 Results ............................................................................ Error! Bookmark not def ined. 3.4 Collapse Safety ................................................................................................................ 52
4 Loss Assessment .................................................................................................................... 54 4.1 Methodology .................................................................................................................... 54 4.2 User-Defined Fragilities ..................................................................................................... 54 4.3 Loss Assessment Results ................................................................................................. 59 4.4 Expected Annual Loss ...................................................................................................... 61
5 Downtime Assessment ............................................................................................................ 63 5.1 Methodology .................................................................................................................... 63 5.2 Downtime Assessment Results ........................................................................................ 66 5.3 Expected Annual Downtime ............................................................................................. 67
6 Cost-Benefit Analysis ............................................................................................................... 68 6.1 Methodology ................................................................................................................ 68
6.2 Results ............................................................................................................................. 72 6.3 Sensitivity Analysis ............................................................................................................ 76
7 Conclusions ............................................................................................................................. 82
5
References ................................................................................................................................ 84 8 .................................................................................................................................................... 84
6
LIST OF TABLES
Table 1 Performance-Based Design Criteria ................................................................................... 17
Table 2 Gravity Loading Criteria ..................................................................................................... 18
Table 3 Code-level Seismic Design Criteria .................................................................................... 19
Table 4 Schedule of Vertical Reinforcement in Core Walls .............................................................. 22
Table 5 Coupling Beam Reinforcement .......................................................................................... 23
Table 6 Concrete Material Properties ............................................................................................. 24
Table 7 Steel Material Properties .................................................................................................... 24
Table 8 Summary of the Dynamic Properties of the Fixed Base Building ........................................ 25
Table 9 Stiffness Assumptions in the Elastic Building Model ........................................................... 26
Table 10 Summary of Isolator Properties ........................................................................................ 28
Table 11 Summary of Non-Structural Component Quantities ......................................................... 30
Table 12 Building Costs Excluding Superstructure ......................................................................... 32
Table 13 Summary of Building Costs ............................................................................................. 33
Table 14 Ground Motion Selection Parameters .............................................................................. 36
Table 15 Ground Motion NGA Reference Number and Scale Factor .............................................. 37
Table 16 Median Peak Overturning Moment .................................................................................. 44
Table 17 Median Peak Story Drift (%) of Alternative Designs at Five Ground Motion Intensities ....... 47
Table 18 Median Peak Story Racking Deformation (%) ................................................................... 48
Table 19 Median Peak Floor Acceleration (g) .................................................................................. 50
Table 20 Median Peak Coupling Beam Rotation ............................................................................ 51
Table 21 Median Base Shear Force ............................................................................................... 51
Table 22 Median Peak Base Overturning Moment .......................................................................... 52
Table 23 Number of Ground Motions where SDR exceeds 5% ...................................................... 52
Table 24 Annual Rate of Mortality ................................................................................................... 53
Table 24 Summary of Loss – % Building Value .............................................................................. 59
Table 25 Expected Annual Loss ..................................................................................................... 61
Table 26 Median Repair Time (Weeks) to Achieve Functionality (Time including impeding factors in
brackets) ............................................................................................................................... 66
Table 27 Median Repair Time (Weeks) to Achieve Full Recovery (Time including impeding factors in
brackets) ............................................................................................................................... 66
Table 28 Expected Annual Time that Building is not functional (days) ............................................. 67
Table 29 Summary of Costs and Benefits ...................................................................................... 70
Table 30 Construction Costs .......................................................................................................... 73
7
Table 31 Summary of Cost Premium Above the Benchmark Fixed Based Design .......................... 73
Table 32 Expected annual loss from building repair ........................................................................ 73
Table 33 Expected loss for each design alternative ........................................................................ 73
Table 34 Summary of Benefits ....................................................................................................... 74
Table 35 Summary of Benefit-Cost Ratios ...................................................................................... 74
Table 37 Expected loss (including downtime only) for each design alternative ................................ 75
Table 38 Expected loss (including downtime and mortality/morbidity) for each design alternative ... 75
Table 39 Summary of Benefits ....................................................................................................... 76
Table 40 Summary of Benefit-Cost Ratios (including downtime and mortality/morbidity) ................ 76
Table 41 Summary of the range of Annualized Loss ....................................................................... 79
Table 42 Summary of the range of Annualized Downtime (days) ..................................................... 79
8
LIST OF ILLUSTRATIONS
Figure 1 Typical Tower Floor Plan used in the PEER Study (Moehle et al., 2011) ............................ 14
Figure 2 Tower Isometric View (Moehle et al., 2011) ....................................................................... 14
Figure 3 Seismic Response Spectra ............................................................................................... 20
Figure 4 Core Wall Layout, Dimensions and Reference Scheme ..................................................... 21
Figure 5 Coupling Beam Orientation ............................................................................................... 23
Figure 6 Screenshot of the Elastic Building Model .......................................................................... 25
Figure 7 Outrigger Schematic ......................................................................................................... 27
Figure 8 Base Isolated Building Schematic ..................................................................................... 28
Figure 9 Example Floor Plan from Planned San Francisco Building (location undisclosed) .............. 30
Figure 10 Response Spectra of Selected Ground Motions (compared to 475 year design basis
spectrum shown in Red) ........................................................................................................ 36
Figure 11 Wall Pier Elements (Black) and Rigid Body Constraints (Colored) .................................... 38
Figure 12 Illustration of location of integration points in wall pier ..................................................... 38
Figure 13 Verification of Shear Behaviour of Coupling Beams (ARUP, 2013) .................................. 39
Figure 14 Coupling Beam Moment-Curvature Relationship ............................................................ 40
Figure 15 Illustration of Racking Deformation ................................................................................. 41
Figure 16 Principal Building Directions ............................................................................................ 42
Figure 17 RP475 Fixed Base Story Drift Ratios (mean shown in black) ........................................... 42
Figure 18 RP475 Fixed Base Racking Drift Ratio (mean shown in black) ........................................ 42
Figure 19 Comparison of SDR and Racking Drifts .......................................................................... 43
Figure 20 RP475 Fixed Base Peak Floor Acceleration .................................................................... 43
Figure 21 RP475 Fixed Base Peak Coupling Beam Rotation .......................................................... 44
Figure 22 RP475 Fixed Base Story Shear Force ............................................................................. 44
Figure 23 Fixed Base Design - Peak Story Drifts at Five Ground Motion Intensities (rad) ................. 45
Figure 24 Outrigger Design - Peak Story Drifts at Five Ground Motion Intensities (rad) ................... 46
Figure 25 Isolated Design - Peak Story Drifts at Five Ground Motion Intensities (rad) ...................... 46
Figure 26 SDR versus Spectral Acceleration .................................................................................. 46
Figure 27 Fixed Base Peak Racking Drifts (rad) at Five Ground Motion Intensities .......................... 47
Figure 28 Outrigger Peak Racking Drifts (rad) at Five Ground Motion Intensities ............................. 48
Figure 29 Base Isolated Peak Racking Drifts (rad) at Five Ground Motion Intensities ....................... 48
Figure 30 Fixed Base Peak Floor Acceleration (g) ........................................................................... 49
Figure 31 Outrigger Peak Floor Acceleration (g) .............................................................................. 49
Figure 32 Base Isolated Peak Floor Acceleration (g) ....................................................................... 50
9
Figure 33 Fixed Base Peak Coupling Beam Rotation (rad) .............................................................. 50
Figure 34 Outrigger Peak Coupling Beam Rotation (rad) ................................................................ 51
Figure 35 Base Isolated Peak Coupling Beam Rotation (rad) .......................................................... 51
Figure 36 Enhanced Curtain Wall Fragility ....................................................................................... 55
Figure 37 Enhanced Partition Fragility ............................................................................................. 59
Figure 38 Fixed Base: Breakdown of Financial Losses ................................................................... 60
Figure 39 Outrigger: Breakdown of Financial Losses ...................................................................... 60
Figure 40 Base Isolated: Breakdown of Financial Losses ............................................................... 61
Figure 41 Illustration of Payback period .......................................................................................... 75
Figure 42 Lognormal Curves Fit to Loss Data ................................................................................. 77
Figure 43 Lognormal Curves Fit to Downtime Data ........................................................................ 78
Figure 44 Annual loss and Downtime Distribution ........................................................................... 78
Figure 45 Upper Bound of Payback ............................................................................................... 80
Figure 46 Lower bound of payback ................................................................................................ 80
10
1 INTRODUCTION
As urban populations rise, the demand for working and living space within cities increases. In
response to heightened demand, cities tend to promote the construction of new tall buildings, which
minimizes urban sprawl and transportation inefficiency. Urban populations, often living in highly
seismic regions, are therefore becoming more susceptible to the seismic risk in tall buildings, both
residential and commercial. Current design guidelines do not always reflect the additional relative risk
that comes from having large numbers of people living in a single structure. Unless the impact of
building failure is considered to be large enough to trigger a classification of ASCE - 7 risk category
III (almost all tall buildings are considered to be in risk category II), the intended safety
margin/probability of collapse for tall buildings is equivalent to that of low to mid-rise structures,
despite the catastrophic social and financial impacts of tall building closure, collapse or demolition.
With knowledge of how our most densely populated buildings are likely to function following large
seismic events we can better predict the overall seismic resilience of a city and inform intelligent
decision processes that optimize the benefits of increased seismic performance.
The primary objective of this research is to benchmark the seismic performance of code-conforming
tall buildings in San Francisco. Secondary objectives are to investigate the structural and non-
structural performance of enhanced building design alternatives and to determine whether the
additional cost of enhanced performance can be financially justified by the reduction in seismic
losses and downtime. The key indicators of seismic performance used in this research are financial
repair cost, building downtime (in particular, the time before the building regains function) and
collapse safety.
On the West Coast of the United States, many recent tall building designs utilize a reinforced
concrete core as the lateral force resisting system, sometimes with outriggers to perimeter columns.
The baseline case study we present is a 42-story coupled core wall residential building located in
downtown San Francisco and designed using a non-prescriptive performance-based approach. We
do not intend to presume that the performance of this particular building model is representative of
11
the performance of all modern tall buildings, however, it is useful to examine the performance of this
building in order to gauge the magnitude of expected direct loss and downtime and to quantify the
relative performance improvement of enhanced designs.
In addition to the baseline case, two enhanced structural schemes were designed with the aim
towards improving performance by reducing the earthquake demands (and damage) on the building.
Both designs retained the general core and coupling beam arrangement; one scheme incorporated
damped outriggers at the mid-height of the building and the other scheme incorporated triple-
pendulum friction bearings beneath the columns and core walls at ground level.
Resilience-based design criteria were implemented on all structural designs, as described in the
REDi™ (ARUP, 2013) guidelines. The purpose of these criteria is to minimize financial losses and
building downtime. The criteria include guidelines for enhanced non-structural component design
and detailing, as well as contingency planning measures to reduce as much as possible the
downtime due to ‘impeding factors’ which delay the commencement of repairs.
The resilience-based building designs (referred to as RBDs) satisfy the criteria of a REDi ‘Gold’ rated
building. In a design level earthquake, a Gold rated building should suffer less than 5% direct
financial loss, be available for immediate re-occupancy and be capable of functional recovery within
one month. This means that any damage that would hinder re-occupancy or prevent a speedy
return to functionality must not occur. For this reason the structure and non-structural components
must not suffer damage requiring more than cosmetic repair. Thus, the building structure should be
designed to remain essentially elastic in the design level earthquake and non-structural components
should be designed and detailed so that they can accommodate the anticipated building
displacements and accelerations with very minor damage. Although the REDi guidelines for a Gold
rated building were followed, it is still necessary to conduct a loss and downtime analysis to show
that buildings meet the Gold level resilience objectives.
12
1.1 The need to Quantify Seismic Resilience of Tall Buildings
In the San Francisco Planning for Urban Resilience (SPUR) report, ‘Safe Enough to Stay,’ the issues
of building downtime and whether buildings are occupiable (and the implications on sheltering in
place) are directly acknowledged as an important factor to the city’s seismic resilience (SPUR,
2012). The performance of tall buildings was not emphasized in the SPUR study, since, in contrast
to vulnerable older concrete or soft first-story buildings, there is a presumption that tall buildings will
not suffer significant damage following a moderate earthquake. This impression is encouraged by
the generally good seismic performance of tall buildings in past earthquake events in the United
States, although evidence to support this impression is limited. The perceived history of satisfactory
performance from tall buildings has deflected attention from the fact that building code provisions
are targeted toward uniform building performance, regardless of the type of structure.
Well-engineered tall buildings may indeed be safer than shorter, stiffer buildings since the probability
that seismic damage will pose a threat to life-safety is less in a tall building (Naeim & Graves, 2005).
However, as Mander and Huang (Mander & Huang, 2012) point out, even well designed structures
may need to be closed for long periods of time following an earthquake, as has recently been seen
in the Canterbury earthquake sequence (Stevenson, 2011).
A survey of many various stakeholders, undertaken as part of the Pacific Earthquake Engineering
Research Center (PEER) Tall Buildings Initiative, found that almost all interviewees believed that the
stated building-code performance expectations for tall buildings is inadequate and that tall buildings
should be classed as special structures, owing to the “devastating” consequence of damage
causing long-term closure (Holmes, Kircher, Petak, & Youssef, 2008). Stakeholder responses
indicated that building cost premiums for better performance on the order of 5-10% of the building
cost would be acceptable, and all agreed that seismic risk (regardless of how small it is perceived to
be) should be communicated to building owners (Holmes et al., 2008).
13
1.2 Previous Studies of Tall Buildings in Los Angeles
The archetype building used for this case study was originally designed by Magnusson Klemencic
Associates (MKA) as part of the Pacific Earthquake Engineering Research Center (PEER) Tall
Buildings Initiative (Moehle et al., 2011). The structure has 42 levels above grade containing
residential apartments and 4 basement levels below grade for parking. The lateral resisting system is
a coupled core wall in the center of the building, housing 4 elevators and 2 stairwells. The building
was originally designed for Los Angeles seismic hazard. For this study, the structure was
redesigned for the seismic hazard in downtown San Francisco at a site near the new Transbay
Terminal (Lat 37.79, Long -122.39). The configuration of the building, including column sizing/layout,
floor height and core dimensions, is identical to that used in the PEER study, with the exception of
an additional four corner columns to support a rectangular floor plate (108’ x 107’).
The floor slabs are 8” thick post-tensioned concrete. Perimeter columns range from 36” square at
the base to 18” square at the top. The thickness of the cores and depth of coupling beams vary
based on which design guideline is used. Since core-only lateral systems are subject to a peer
review process which requires performance-based design and assessment, the designs for
Buildings 1B (to LATBSDC) and 1C (to PEER TBI Guidelines) are the most likely to be similar to the
building design presented herein, owing to the performance-based approach used. The core wall
thickness for Building 1B varied from 28” to 32” at the base, reducing to 21” at the roof. The core
wall thickness for Building 1C varied from 32” to 36” at the base, reducing to 21” at the roof. For
more information, see (Moehle et al., 2011).
14
Figure 1 Typical Tower Floor Plan used in the PEER Study (Moehle et al., 2011)
Figure 2 Tower Isometric View (Moehle et al., 2011)
The goal of the PEER study was to assess the performance of tall buildings designed using three
different guidelines: a) the 2006 International Building Code (IBC, 2006), b) the Los Angeles Tall
Buildings Seismic Design Guideline (LATBSDC, 2008) with some modifications and c) the PEER Tall
Building Initiative (PEER TBI, 2010) which were in draft at the time of the study. In the PEER study, a
time-based loss assessment was carried out by analyzing the performance of the building at 5
different hazard levels. Two methods were used to estimate the direct financial losses caused by
earthquake damage. The first, considered to be state-of-the-practice, was developed by Risk
Management Solutions Inc. (RMS). The second method used to estimate the direct financial losses
was based on the ATC-58 framework, still in progress at the time (Moehle et al., 2011). The median
15
repair cost of the building, calculated using the ATC-58 method, is estimated to be between 8.4%
(for Building 1C) and 10.8% (Building 1B) of the initial building cost following a design basis
earthquake (DBE) intensity level. The RMS study only derived losses relative to the code-designed
building (Building 1A), and is not presented here.
1.3 Overview of the REDi Rating System
The REDi™ Rating System (REDi, 2013) is an actionable framework for owners, engineers,
architects, and other design team members to implement resilience-based earthquake design. The
framework is intended to reduce the extent of building damage and the associated financial losses,
as well as to reduce the time in which the building cannot be occupied or is unable to perform it’s
primary function. The REDi™ resilience objectives exceed code-intended performance objectives
and, by corollary, typical performance objectives in the PEER and LATBDC designs and are
intended to be equivalent to code-based designs.
The REDi™ framework has mandatory requirements and non-mandatory recommendations based
on the rating tier desired (Platinum, Gold, or Silver). To qualify for a REDi™ rating, it is necessary to
satisfy the mandatory requirements for that tier in a number of design and planning categories. A
Loss Assessment must be performed to verify that a building meets the REDi™ resilience objectives
- measured in terms of downtime and financial loss. The REDi Downtime methodology (described in
subsequent chapters) may be used to determine the expected time before three distinct recovery
states are achieved; re-occupancy, functional recovery and full recovery. The reader is referred to
the website where the REDi™ Rating System is available for download (REDi, 2013).
1.3.1 Utility disruption
To achieve a REDi Gold rating, the estimated functional recovery time of the building must be less
than one month following a design level earthquake event. It is assumed that all utilities will be
restored to a minimal level of function within this timeframe, so that the disruption to utilities does not
preclude attainment of the functional recovery objective. The REDi™ guidelines require the owner to
be informed if there is evidence that any of the utilities would be disrupted for longer than 1 month,
16
however, in the absence of contraindications, the assumption that utilities will be restored within one
month is acceptable. Back-up systems should be able to support building occupancy, such that
residents are not required to vacate the building, but a minimal level of functionality is sufficient.
Based on Section A4.3 in the REDi™ guidelines, the estimated repair rates for pipes for a design
level earthquake in San Francisco (based on peak ground velocity) is greater than 0.2 breaks/km.
Therefore, according to the REDi™ guidelines, the median estimates for electricity, water, and
natural gas disruption is 3 days, 21 days, and 42 days, respectively. The disruption time for natural
gas slightly exceeds the functionality objective of 1 month, but by the logic described above, the
building still qualifies for a Gold rating.
17
2 BUILDING DESIGN
2.1 Fixed Base Building Design
2.1.1 Design Objectives
The fixed base building was designed using a non-prescriptive seismic design approach intended to
represent the state of practice for tall building design in San Francisco. The general performance
objectives for the structure are to provide “collapse-prevention” in the maximum considered
earthquake (MCE), “life-safety” in a design level earthquake (DBE), and minimal damage in
serviceability level earthquakes (SLE). The design guidelines set out in PEER TBI Guidelines (2010)
require that the structure be designed to remain essentially elastic under serviceability earthquake
demands and evaluated through a non-linear response history analysis (NLRHA) to explicitly verify
that the “collapse prevention” objective is satisfied for the MCE. Although the PEER TBI Guidelines
do not specify a performance objective for the DBE, the local jurisdiction requires that the design
meet minimum code requirements under the DBE hazard level, generally associated with a “life-
safety” objective.
Table 1 Performance-Based Design Criteria
SLE DBE MCER
Overall Objective Minimal structural damage Code compliance implies Life-Safety
Collapse Prevention
Method used to generate demands
Response Spectrum Analysis
Response Spectrum Analysis
NLRHA
Force Reduction Factor Un-factored R = 6 Un-factored Acceptance Criteria Story drift ratio < 0.5%
Demand<1.5ΦNominal Capacity
Demand< Φ Nominal Capacity
Mean story drift ratio < 3.0% Coupling beam rotation < 0.06 1.2V*<V Concrete compression strain < 0.015 Rebar tensile strain < 0.05 Rebar compression strain < 0.02
Material Properties Nominal Strength Nominal Strength Expected Strength Strength reduction factor
Code defined Code defined N/A*
* Strength reduction factors were set to unity, which is consistent with the original MKA designs
18
The core-only lateral system is designed such that energy is dissipated through two flexural yield
mechanisms: plastic hinges at the base of each wall pier and the ends of the coupling beams up the
height of the building. Capacity design principles are used to design the flexural reinforcement in the
core walls. Flexural yielding of reinforcement is limited to the bases of the wall piers by designing for
substantially lower demand to capacity ratios (DCRs) in the mid/upper levels where flexural yielding
is undesirable and higher DCRs where yielding of reinforcement is desirable (Priestley et al., 2007).
Table 1 provides a summary of the design criteria followed to satisfy the performance objectives and
intended yield locations.
2.1.2 Loading Criteria
In general, the loading criteria used to design the building is the same as that used in the PEER TBI
study (PEER, 2012). A summary of the gravity loads applied to the structure, in addition to self-
weight is presented in Table 2. Gravity loads are the minimum loads required by ASCE 7-10 and the
PEER guidelines.
Table 2 Gravity Loading Criteria
Superimposed Dead Loads (kN/m2)
Live Loads (kN/m2)
Car Parking 0.14 1.9 Level 1 Retail (under tower footprint) 5.3 4.8 Level 1 Plaza 16.8 4.8 Residential Areas 1.3 1.9 Exit Areas (inside core walls) 1.3 4.8 Roof 1.3 1.2
Seismic design loads, in accordance with ASCE 7 requirements, were calculated to ensure that the
building design would, at least, meet code criteria. The parameters used to determine the code level
design loads are described in Table 3. The code-level base shear was compared to the service level
base shear obtained using response spectrum analysis (see Section 2.1.3) to confirm that the
service level analysis would control the structural member design checks.
19
Table 3 Code-level Seismic Design Criteria
Parameter Value Occupancy Category II Importance Factor 1.0 Spectral Acceleration Ss = 1.5 ; S1=0.6 Site Class D Site Class Coefficients Fa = 1.0 ; Fv = 1.5 Spectral Response Coefficients Sds = 1.0 ; Sd1 = 0.6 Seismic Design Category D Lateral System Building frame, special reinforced concrete shear walls Building Period 2.55s (based on H=124m) Cs (Eq 12.8.2) 0.167 Csmax (Eq 12.8.3) 0.039 Csmax (Eq 12.8.5) 0.044 Csmin (Eq 12.8.6) 0.05 Seismic response Coefficient 0.05 Seismic Weight (Above ground) 400 MN Design Base Shear (Φ=0.85) V = 0.85CsW = 0.041W = 16.9 MN
2.1.3 Response Spectrum Analysis (RSA)
Response spectrum analysis was conducted at the service level hazard and the design level hazard.
The service level spectrum is based on a 50% probability of exceedance within a 30-year period (43-
year return period), considering 2.5% damping. The service level spectrum used to design the
building was developed by ARUP, specifically for the Transbay terminal site. ARUP performed a site-
specific seismic hazard analysis per Chapter 21 of ASCE 7-10 to construct the response spectrum.
They used four of the Next Generation Attenuation (NGA) relationships, to predict response spectra
based on earthquake magnitude, distance, and site class. The maximum horizontal response was
obtained by multiplying the geomean estimate of the hazard by period-dependent NEHRP Maximum
Demand factors (NEHRP, 2009). Figure 3 shows the SLE response spectrum compared to the
code-level design spectrum and the MCE response spectrum. The load case used for the modal
response spectrum analysis is shown in Equation 1.
Load Combination = 1.0D + 0.25L + 1.0E (1)
Eight modes were considered, giving an effective mass equal to 92% of the building weight. The
modal responses were combined using the complete quadratic combination (CQC) method. The
20
elastic base shear under the SLE analysis was found to be 6.2% of the building weight, whereas the
factored base shear (using an R-factor of 6) under the DBE level analysis was found to be 4.1%. The
results of the RSA, including a description of the linear elastic building model, are discussed in
Section 2.1.7.
Figure 3 Seismic Response Spectra
2.1.4 Design of Core Walls
In addition to the design criteria in Table 1 the thickness of each wall was checked to limit the shear
stress demand expected at the MCER level to 8√f’cAc, from ACI 318 (ACI, 2011). In recognition that
higher mode shear demands may not be significantly reduced by flexural ductility in shear wall
structures, the shear demands in the walls were initially estimated for design using a modified modal
superposition in each of the primary directions (Priestley et al., 2007) given by,
𝑉!"#$%& =!!!+ 𝑉! + 𝑉!…+ 𝑉! (2)
where Vn corresponds to the elastic base shear in each mode (in each direction), R is assumed to
be 6 and n is the number of modes required to achieve at least 90% mass participation. Essentially,
this approach reduces the shear in first-mode response by the R-value, assuming that the flexural
0
2
4
6
8
10
12
14
16
0 1 2 3 4 5 6 7 8 9 10
Spec
tral R
espo
nse
Acce
lera
tion
(m/s
^2)
Period (s)
Sesimic Response Spectra
Code Level Design Spectra
SLE
MCE
21
wall hinging will limit the first mode response. Shears from higher modes are kept at their unreduced
elastic value. The shear demands from the NLRHA are later evaluated to confirm that the ACI limit is
satisfied.
The thickness of the core walls is also checked to ensure that the average axial stress under
expected gravity loads does not exceed 0.15f’c. This is a rule of thumb, intended to ensure ductility
in the hinge mechanism at the base of the wall piers (i.e. higher compressive loads may lead to
brittle compression failure in nonlinearly responding walls). This is more conservative than PEER
(2012) which used an allowable axial stress of 0.30f’c. For this building, it turns out that the
thickness of the wall is generally governed by the MCER shear strength limit. An illustration of the
dimensions of the core wall is presented in Figure 4. Table 4 shows the schedule of vertical
reinforcement in the core wall piers.
Figure 4 Core Wall Layout, Dimensions and Reference Scheme
22
Table 4 Schedule of Vertical Reinforcement in Core Walls
Pier Height Vertical Reinforcement Schedule Pier 1 Basement 22.2 mm diameter bars @ 152mm c/c L1-4 22.2 mm diameter bars @ 152mm c/c L5-12 22.2 mm diameter bars @ 152mm c/c L13-21 28.7 mm diameter bars @ 152mm c/c L22-30 25.4 mm diameter bars @ 152mm c/c L31-42 19.1 mm diameter bars @ 152mm c/c Roof 22.2 mm diameter bars @ 305mm c/c Pier 2 L1-4 35.8 mm diameter bars @ 102mm c/c L5-12 25.4 mm diameter bars @ 102mm c/c L13-21 28.7 mm diameter bars @ 102mm c/c L22-30 28.7 mm diameter bars @ 102mm c/c L31-42 25.4 mm diameter bars @ 102mm c/c Pier 3 Basement 22.2 mm diameter bars @ 152mm c/c L1-4 28.7 mm diameter bars @ 305mm c/c L5-12 22.2 mm diameter bars @ 152mm c/c L13-21 22.2 mm diameter bars @ 152mm c/c L22-30 25.4 mm diameter bars @ 152mm c/c L31-42 19.1 mm diameter bars @ 152mm c/c Roof 25.4 mm diameter bars @ 305mm c/c Pier 4 Basement 22.2 mm diameter bars @ 203mm c/c L1-4 32.3 mm diameter bars @ 203mm c/c L5-12 25.4 mm diameter bars @ 203mm c/c L13-21 25.4 mm diameter bars @ 203mm c/c L22-30 22.2 mm diameter bars @ 203mm c/c L31-42 22.2 mm diameter bars @ 203mm c/c Roof 22.2 mm diameter bars @ 305mm c/c Pier 5 L1-4 25.4 mm diameter bars @ 203mm c/c L5-12 25.4 mm diameter bars @ 203mm c/c L13-21 22.2 mm diameter bars @ 203mm c/c L22-30 22.2 mm diameter bars @ 203mm c/c L31-42 19.1 mm diameter bars @ 203mm c/c
2.1.5 Design of Coupling Beams
Coupling beams were modeled using expected stiffness properties and designed using expected
material strength properties. Coupling beams were designed for shear and flexure in accordance
with ACI 318 with a demand/capacity ratio for flexure under SLE limited to 1.5. A schedule of
coupling beam reinforcement is presented in Table 5.
23
Figure 5 Coupling Beam Orientation
Table 5 Coupling Beam Reinforcement
Diagonally Reinforced Coupling Beam Schedule (Bar Sizes Reference Nominal Diameter in mm)
Diagonals Cross-Ties Horizontals Levels Depth Width Bar
Size Columns Rows Bar Size
X Direction
Y Direction Longitudinal Bar
Size Notes
(mm) (mm) # Of Bars
# Of Bars
# Of Bars
# Of Bars
Spacing (mm)
Mp
(Kn-M) Basement 864 813 19.1 3 3 15.9 6 6 127 15.9 558
31-roof NS 762 533 19.1 3 2 15.9 8 5 127 15.9 357
31-roof EW 762 533 19.1 3 3 15.9 8 5 127 15.9 465
1 to 4 NS 762 813 28.7 3 3 15.9 6 6 127 15.9 1036
1 to 4 EW 762 813 28.7 3 3 15.9 6 6 127 15.9 1036
5 to 12 NS 762 813 28.7 3 3 15.9 6 6 127 15.9 1036
5 to 12 EW 762 813 28.7 3 3 15.9 6 6 127 15.9 1036
13 to 21 NS 762 610 22.2 2 3 15.9 7 5 127 15.9 420
13 to 21 EW 762 610 22.2 3 3 15.9 7 5 127 15.9 630
13 to 30 NS 762 610 19.1 3 3 15.9 7 5 127 15.9 465
13 to 30 EW 762 610 19.1 3 4 15.9 7 5 127 15.9 522
24
2.1.6 Building Materials
The material properties used for the design are summarized in Table 6 and Table 7. Note that the
material properties used are the same as those used in the original PEER design.
Table 6 Concrete Material Properties
Description Nominal Strength (MPa)
Expected Strength (MPa)
Nominal Stiffness (GPa)
Expected Stiffness (GPa)
Basement Walls 34 45 26.4 29.1 Non-PT Beams and Slabs 38 50 27.4 30.3 PT Floor Slabs 38 50 27.4 30.3 Columns 55 72 31.6 35.0 Shear Walls 55 72 31.6 35.0
Table 7 Steel Material Properties
Description Nominal Yield Strength (MPa)
Expected Yield Strength (MPa)
Expected Ultimate Strength (MPa)
Shear Wall Reinforcement 410 480 720 Coupling Beam Reinforcement 520 590 900
2.1.7 Elastic Building Model
The elastic building models were developed in 3D and analyzed using the software Oasys GSA V8.6.
The model included the shear walls, floor slabs, coupling beams, gravity columns and basement
walls. The shear walls, coupling beams and columns were modeled using beam elements and the
floor slabs and basement walls were modeled using 2D elastic shell elements. The model included
all structural elements down to the foundation, not including the foundation piles. Soil springs were
not incorporated in the model and the lateral stiffness of the soil surrounding the basement walls
was neglected.
Modal analysis and response spectrum analysis on the elastic building model (fixed at the bottom of
basement) yielded the information summarized in Table 8. The base shear obtained from the
response spectrum analysis was compared to that obtained from a code-level evaluation to ensure
that the design-level base shear is, at minimum, 85% of the code base shear.
25
Figure 6 Screenshot of the Elastic Building Model
Table 8 Summary of the Dynamic Properties of the Fixed Base Building
Strong Direction (Y-axis) Weak Direction (X-axis) Period (DBE) T1 = 4.37 s
T2 = 0.93 s T1 = 5.27 s T2 = 1.10 s
SLE Base Shear 25 MN 19 MN DBE Base Shear (R-Factor Included)
13 MN 10 MN
Code Level Design Base Shear 16.9 MN 2.1.8 Stiffness Assumptions
Cracked section properties were used to represent the stiffness of all elements, following the
guidelines set out by ACI (ACI, 2008) and PEER TBI (2010). These properties are summarized in
Table 9. ATC 72 presents results (from the Naish tests) which indicate that an effective stiffness
approximately equal to 15% of EIg is appropriate for coupling beams, as this accounts for the added
flexibility at the beam-wall interface due to slip/extension of the reinforcement (ATC 72, 2010).
Therefore, an effective stiffness of 0.15EIg is used for the coupling beams.
26
Table 9 Stiffness Assumptions in the Elastic Building Model
Element Stiffness Assumptions SLE DBE
Shear Walls 0.6 Ig 0.35 Ig Basement Walls 1.0 Ig 0.8 Ig Coupling Beams 0.5 Ig 0.15 Ig Tower floor slabs 0.5 Ig 0.35 Ig Basement Floor Slabs 0.5 Ig 0.25 Ig
2.2 Damped Outrigger Building Design
2.2.1 Outrigger Design
The damped outrigger (Smith and Willford, 2007) design adopted the same core wall and coupling
beam design as the fixed base case, but incorporated linear viscous dampers between concrete
wall outriggers and perimeter columns at mid-height of the building. These are intended to dissipate
additional energy and reduce the deformation demands on the structure. The concrete wall
outriggers were designed to remain essentially elastic. The width and depth of the columns receiving
the damped outriggers were increased by 50mm (the size was increased proportionally to the
increase in force) to ensure that their compressive capacity was adequate under the MCE demands
delivered by the dampers. An effective damping coefficient for the viscous dampers was obtained
using a frequency ratio method developed by Smith and Willford (2007). In the x and y directions,
the optimal properties of the dampers were estimated to be 32 MN s/m and 26 MN s/m,
respectively. Since no iterative process was employed to find the optimum properties for the
dampers, it is possible that the damped outrigger building performance, shown below, can be
improved. A schematic of the damped outrigger design is shown in Figure 7.
27
Figure 7 Outrigger Schematic
2.3 Base-isolated Building Design
2.3.1 Isolator Design
The base isolated building was designed to remain essentially elastic in the DBE. During initial
design, triple friction pendulum bearings were modeled with equivalent linear springs assuming a
representative isolation period of T = 5 seconds. The properties of the triple friction pendulum
isolators are those of Earthquake Protection Systems (EPS) FPT15670. The properties for each
sliding surface are summarized in Table 3. The lowest friction value is approximately half of the wind
base shear, which indicates that there may be some movement (less than 30 mm) in a design level
wind storm, based on the isolator hysteretic curve.
The thickness of the core walls was determined using the same methodology described in the
design of the fixed base building; however, for the base isolated case the governing factor was axial
stress under gravity loads, since the isolators greatly reduce the shear forces transferred to the
structure. In general, this design permits reduced wall thickness and reduced reinforcement ratio. As
noted above an axial stress ratio of 0.15f’c was used, which is conservative compared to PEER
(2012) and likely conservative for this case since significant ductility is not expected in the walls.
28
Therefore, it is likely that the walls thicknesses could be reduced further as long as the loads from
gravity, wind, and earthquake shear demands in the MCE are lower than the wall capacity.
The isolation scheme, specifically the location of the isolation plane, was created to allow the
elevators to run un-interrupted from ground level to the highest floor level. At ground level, the gravity
columns were isolated from the basement levels. The core wall was isolated one floor below ground
level in order to provide space for the elevator pit in the ‘above isolation’ structure. A schematic of
the design is shown in Figure 5. Other solutions, possibly more cost-effective, were not investigated.
Another bank of elevators, which transfers passengers from the basement car parks to the ground
level, could be located outside the footprint of the superstructure and therefore does not need to
cross the isolation plane. A grid of beams is added above the isolation plane and existing beams
below the isolation plane were designed to carry the eccentric axial forces in the columns and core
walls caused by the displacement across the isolators. Preliminary design of these beams was
undertaken so that the concrete and steel material quantities could be included in the building cost
estimate.
Figure 8 Base Isolated Building Schematic
Table 10 Summary of Isolator Properties
Upper Bound Friction Properties
Reff d
Surface 1 0.015 3.7592 0.508508 Surface 2 and 3 0.030 0.5842 0.05207 Surface 4 0.050 0.5842 0.05207
29
2.4 Design of Non-Structural components
The PEER and LATBDC tall building performance-based design guidelines do not explicitly consider
the design or performance of non-structural components. Instead, they recommend that non-
structural systems be designed to meet requirements outlined in building codes like ASCE 7-10
(ASCE, 2010). In general, the code intends that the performance of non-structural components does
not pose a “life-safety” hazard under design level shaking (NEHRP, 2009). In current design
practice, it is rare for the design and detailing of non-structural components to exceed the minimum
code requirements.
Two non-structural design alternatives were considered. The first of these will be referred to as the
code-compliant non-structural design, intended to be representative of current practice. The second
non-structural design is the REDi non-structural design, which includes design enhancements
specified in the REDi Guidelines (REDi, 2013). The REDi design is intended to reduce the losses
associated with seismic damage by providing enhanced partitions, curtain walls and elevator rails,
which delay the onset of damage to these components. Further information on the design of the
enhanced components can be found in Section 4.2.
2.4.1 Non-Structural Component Quantities
The quantity of non-structural components does not differ between the two design alternatives. A
summary of the component quantities used to calculate the non-structural costs is presented in
Table 11 (imperial Units are presented as these are consistent with the component fragility curves
used in the loss estimation software). The values were taken from three different sources; a quantity
estimator from Arup, San Francisco; the values listed in the TBI case studies (PEER, 2012); and the
normative quantities recommended in FEMA P58 (2012). Normative quantities were only used where
additional information was unavailable. The quantity of partitions is taken from the normative
spreadsheet but has been additionally verified using example floor plans (see Figure 9) of high-rise
residential buildings.
30
Figure 9 Example Floor Plan from Planned San Francisco Building (location undisclosed)
Table 11 Summary of Non-Structural Component Quantities
Floor Component Quantity Units Source B1 Chiller 2 Each Arup Estimator Air Handling Unit - Capacity: 5000 to
<10000 CFM 1 Each Arup Estimator
Air Handling Unit - Capacity: 10000 to <25000 CFM
1 Each Arup Estimator
Fire Sprinkler Water Piping 5.69 1000 LF Normative Value Fire Sprinkler Drop 4.14 100
Each Normative Value
Transformer - Capacity: 750 to 1500 kVA 1 Each Arup Estimator Low Voltage Switchgear - Capacity: 750
to <1200 Amp 1 Each Arup Estimator
Low Voltage Switchgear - Capacity: 1200 to 2000
2 Each Arup Estimator
Distribution Panel - Capacity: 1200 to 2000
1 Each Arup Estimator
B2-B3 Fire Sprinkler Water Piping 5.69 1000 LF Normative Value Fire Sprinkler Drop 4.14 100
Each Normative Value
L1 Suspended Ceiling 41.64 250 LF Normative Value Independent Pendant Lighting 174 Each Normative Value Traction Elevator 4 Each PEER Cold Water Piping (dia. > 2.5 inches) 0.02 1000 LF PEER Hot Water Piping (dia. < 2.5 inches) 0.88 1000 LF PEER Hot Water Piping (dia. > 2.5 inches) 0.06 1000 LF PEER Sanitary Waste Piping 0.54 1000 LF PEER Chilled Water Piping 0.45 1000 LF PEER
31
Steam Piping 0.06 1000 LF Normative Value HVAC Galvanized Sheet Metal Ducting
(dia. < 6 ft2) 0.06 1000 LF Normative Value
HVAC Galvanized Sheet Metal Ducting (dia. > 6 ft2) 0.58
1000 LF Normative Value
HVAC Drops / Diffusers 8.2
100 Each
PEER
Fire Sprinkler Water Piping 2.08 1000 LF Normative Value
Fire Sprinkler Drop 0.93 100 Each
Normative Value
Fire Sprinkler Drop - No Ceiling 0.93
100 Each
Normative Value
Curtain Walls 131.7 30 SF PEER Wall Partition 5.24 100 LF Normative Value Prefabricated steel stairs 2 Each PEER Wall Partition with Wallpaper Finish 1.58 100 LF Normative Value L2-L42 Cold Water Piping (dia. > 2.5 inches) 0.02 1000 LF PEER Hot Water Piping (dia. < 2.5 inches) 1.66 1000 LF PEER Hot Water Piping (dia. > 2.5 inches) 0.06 1000 LF PEER Sanitary Waste Piping 1.43 1000 LF PEER Chilled Water Piping 1.47 1000 LF PEER HVAC Fan 0.6 10 Each Arup Estimator HVAC Galvanized Sheet Metal Ducting 0.58 1000 LF Normative Value
HVAC Drops / Diffusers 8.2 100 Each
PEER
Variable Air Volume (VAV) box 0.6 10 Each Arup Estimator Fire Sprinkler Water Piping 2.55 1000 LF Normative Value
Fire Sprinkler Drop 1.39 100 Each
Normative Value
Curtain Walls 131.7 30 SF PEER Wall Partition 15.88 100 LF Normative Value Prefabricated steel stairs 2 Each PEER Wall Partition with Wallpaper Finish 4.76 100 LF Normative Value Roof Cooling Tower - Capacity: 350 to <750
Ton 3 Each Arup Estimator
HVAC Fan 4 Each Arup Estimator Air Handling Unit - Capacity: 10000 to
<25000 CFM 1 Each Arup Estimator
The quantities listed in Table 11 are used as inputs to the loss estimation tool, described in Section
4. Damage to non-structural components usually dominates seismic losses; therefore the quantity of
non-structural components is likely to have a large impact on these losses. For this reason, the non-
structural building quantities were crosschecked using multiple sources, including building plans, for
accuracy.
32
2.5 Summary of Building Costs
2.5.1 Building Costs Excluding Superstructure
The cost of most components is based on the original cost estimates from Davis Langdon (PEER,
2012), adjusted for inflation and for location. The cost of interior partitions and doors were provided
by an experienced cost estimator and cost estimates for the elevators and unitized façade systems
were obtained from vendors. Excavation, foundation work, and similar site preparation cost
estimates were taken from an Arup feasibility study, which assumed that a piled mat foundation
would be the most economic for the site in San Francisco (no piles were used in the PEER study).
The costs associated with non-structural enhancements (for the REDi™ designs) were based on
conversations with vendors (elevators and facades) or otherwise based on judgment. Table 12
shows building costs for the code compliant design and the REDiTM non-structural design.
Table 12 Building Costs Excluding Superstructure
Category Code Compliant Design Additional Costs for REDi™ Design
Comments
Excavation for basement $6,809,094 $- Foundations - Mat and Piling $7,630,961 $- Landscaping / plantings $96,611 $- Exterior paving $257,630 $- Prefabricated Steel Stairs $2,376,763 $- Provide slip joints Carpentry $1,096,460 $- Roof Waterproofing $1,724,589 $- Roof Insulation $133,988 $- Roofing - Balconies $1,242,435 $- Curtain Walls - Façade Unitized System
$21,254,400 $- Accommodate higher drifts
Exterior Doors, Frames, and Hardware
$606,467 $25,000 Hardware and keys for tenants
Standard Partitions $5,719,087 $285,954 5% cost premium for slip connections
CMU in Basement $660,730 $- Interior Doors $5,063,950 $- Floor Finishes Incl. Basement $8,794,210 $- Wall Finishes $4,357,374 $- Ceiling Finishes $4,686,578 $- Kitchen Cabinets $1,381,692 $- Kitchen Appliances $2,587,921 $- Signs $87,410 $- Toilet Fixtures $1,445,700 $-
33
Fire Protection Systems $4,503,990 $- Plumbing $15,978,643 $- HVAC $11,632,708 $- Trash Chute $214,691 $- Blinds / Drapes / WT $323,571 $- Elevators $4,256,000 $1,600,000 4 x $400k for OSHPD
enhancements Electrical $17,906,440 $- Fire Extinguishers Cabinets and Parking Garage Equipment
$113,559 $-
Built in Equipment $509,224 $- Washing Machine and Dryer $913,384 $- Window Washing Equipment $552,450 $- Utilities On-Site - Site Utilities $613,833 $- Total Costs $135,532,543 $137,443,498
2.5.2 Total Building Costs
A summary of the total building costs, including both structural and non-structural components, is
presented in Table 13. The cost premium for enhanced design ranges between 1.0%, for the
performance-based outrigger building, and 2.47% for the REDi Isolated building.
Table 13 Summary of Building Costs
TOTAL BUILDING COSTS Fixed Base Outrigger Base Isolated PBD REDi™ PBD REDi™ PBD REDi™ Superstructure Hard Costs $40,513,922 $40,513,922 $42,269,988 $42,269,988 $42,943,400 $42,943,400
Non-Superstructure Hard Costs $135,532,543 $137,443,498 $135,532,543 $137,443,498 $135,532,543 $137,443,498
TOTAL COSTS $176,046,466 $177,957,420 $177,802,532 $179,713,486 $178,475,943 $180,386,898
Cost per square-foot $271 $273 $273 $276 $274 $277
Cost Premium $- $1,910,954 $1,756,066 $3,667,020 $2,429,478 $4,340,433
Cost Premium % - 1.09% 1.00% 2.08% 1.38% 2.47%
For the damped outrigger building, the cost premium is attributed to the viscous dampers
($720,000) and the additional materials required for the outrigger walls and the perimeter columns
($775,139 for concrete and $260,927 for steel). For the base isolated building, the cost premium is
34
attributed to the double framing at the isolation level ($101,549 for concrete and $27,929 for steel),
the isolators themselves ($1,900,000) and the addition of flexible connections at the isolator level
($400,000).
35
3 NLRHA
Nonlinear Response History Analysis (NLRHA) was undertaken on each of the building models.
Gravity loads were applied to the structure (ramped up over the first 0.5 seconds of the analysis) and
then held constant throughout the seismic response simulation. At the end of the earthquake record,
the structure was allowed to oscillate for an additional 15 seconds, during which time additional
viscous damping was applied (~5% critical damping) so that the residual drifts could be accurately
gauged without excessive additional runtime.
3.1 Ground Motion Selection and Scaling
The archetype building was assumed to be located in downtown San Francisco, near the new
Transbay Terminal (Lat 37.79, Long -122.39). This is approximately 14 km from both the San
Andreas fault and Hayward fault. The average shear wave velocity (Vs30) of the site “at-grade” was
taken as 215m/s in the upper 30m. The average shear wave velocity in the 30m below the bottom of
the mat foundation is 259 m/s and referred to as “at-foundation” conditions. In either case, ASCE 7-
10 defines the site as Class D.
Five hazard levels were considered, having mean probabilities of exceedence in 50 years
corresponding to 80%, 31%, 10%, 2% and 1% (with corresponding mean return periods of 31, 135,
475, 2475 and 4975 years). For each hazard level, probabilistic seismic hazard deaggregation for
the site was undertaken using the USGS hazard deaggregation tool (USGS, 2014). The tool was
used to determine the spectral acceleration (Sa) at a period of 5 seconds, and the mean magnitude
and distance of the earthquakes contributing to the seismic hazard.
Twenty ground motions were selected at each hazard level using Matlab scripts developed by Jack
Baker, which are publicly available online (Baker, 2014). The Sa, mean magnitude and distance
values obtained from USGS are used as inputs to the Matlab scripts to generate a target geometric
mean Conditional Mean Spectrum (using a conditioning period of 5s). A summary of the ground
motion selection parameters is presented in Table 14.
36
Ground motions are selected using a computationally efficient algorithm developed by Jayaram et al
(2011), which selects ground motions whose response spectra match the target response spectrum
mean and variance. The ground motions selected herein are intended to be representative of the
seismic hazard at the site for the purpose of loss assessment, and do no necessarily conform to
ASCE 7-10 (2010). The response spectra of the ground motions are compared to the code-level
475-year return period (design basis) spectrum in Figure 10.
Table 14 Ground Motion Selection Parameters
Mean from Deaggregation
ε (from Baker script)
Prob. of ex. Time (years)
Return P (years)
Distance (kM) Magnitude ε Sa (g)
0.500 21 31 44.8 7.02 -0.5 0.0184 -0.30
0.200 30 135 22.9 7.39 0.18 0.0663 0.20
0.100 50 475 17.1 7.63 0.71 0.1479 0.69
0.020 50 2475 14.9 7.79 1.36 0.3058 1.37
0.010 50 4975 14.4 7.83 1.59 0.3897 1.63
Figure 10 Response Spectra of Selected Ground Motions (compared to 475 year design basis spectrum shown in Red)
1 2 3 4 5 6 7 8 9 100
0.5
1
1.5Response Spectra of Ground Motions Selected
Sa (g
)
Period (s)
37
Table 15 Ground Motion NGA Reference Number and Scale Factor
RP31 RP135 RP475 RP2475 RP4975 Record Number
NGA Ref
Scale Factor
NGA Ref
Scale Factor
NGA Ref
Scale Factor
NGA Ref
Scale Factor
NGA Ref
Scale Factor
1 2748 0.61 1553 0.41 2502 2.39 1343 1.38 1343 1.75 2 1626 1.17 1312 3.72 729 2.71 1045 3.86 1504 3.47 3 140 1.45 1782 3 1489 1.67 1238 1.53 1195 2.33 4 1259 1.66 1177 2.75 1505 0.4 1538 3.06 1505 1.06 5 1197 0.25 1118 1.15 1551 1.93 1531 2.28 1240 3.75 6 1837 2.64 169 0.8 1491 1.44 1499 3.29 1550 3.4 7 3285 1.49 729 1.21 1492 0.57 1492 1.18 1492 1.49 8 2112 1.77 1798 2.54 1826 2.34 1553 1.89 1180 1.95 9 721 0.33 1436 1.8 1493 1.87 1505 0.84 1538 3.88 10 436 3.56 1210 3.05 1470 3.67 1482 2.15 1553 2.39 11 804 1.73 36 2.63 1465 3.71 1629 3.3 1537 3 12 832 0.44 1546 0.77 192 2.39 1477 1.94 1238 1.94 13 1809 1.47 1489 0.75 1499 1.59 1554 3.52 1531 2.9 14 1521 0.27 163 1.61 1545 1.33 1527 2.39 1527 3.03 15 882 0.89 832 1.6 1430 3.79 1550 2.68 1529 1.98 16 1588 1.09 1771 3.19 2899 3.39 1542 2.3 1542 2.91 17 1816 1.03 170 0.79 1453 3.74 1497 2.44 1548 2.05 18 1489 0.21 1491 0.64 1525 2.21 1502 2.45 1477 2.46 19 791 1.22 1441 3.19 1494 1.43 1514 3.66 1502 3.11 20 1153 1.38 1223 2.56 1475 2.6 1472 3.87 1497 3.09
3.2 Non-Linear Building Model
The 3D non-linear building models were modified from the elastic models using the Oasys suite of
pre-processors for analysis in LS-Dyna (LSTC). The concrete core walls were modeled using fiber
beam elements with multiple integration points representing the reinforcing steel, the confined
concrete and the unconfined concrete. The fiber beam elements were connected using rigid body
constraints (see Figure 11) in order to account for the connection of wall piers and to enable the
coupling beams to be connected to the wall piers at the pier edge. 40-90 integration points were
used per wall section. In general, each wall pier was divided into 5-10 smaller sections, and fibers
representing the concrete and steel were defined for each section (see Figure 12). For smaller wall
piers, individual integration points were used to define each bar of reinforcement.
38
Figure 11 Wall Pier Elements (Black) and Rigid Body Constraints (Colored)
Figure 12 Illustration of location of integration points in wall pier
The unconfined concrete, confined concrete and reinforcing steel were modeled using
‘MAT_CONCRETE_EC2’ (see LSTC, 2007). ‘The moment curvature (M-k) relationships of the core
wall pier sections defined in LS-Dyna were found to compare very closely with those developed
independently using XTRACT (TRC, 2007). The intrinsic damping of the building (damping not
39
associated with non-linear hysteresis) is assumed to be 2.5% of critical, modeled as constant
damping (frequency independent) over the period range 0.10 sec to 10.0 sec.
3.2.1 Coupling Beam Elements
The coupling beams were modeled as lumped plasticity beam elements, defined between nodes at
the extreme edges of the wall piers. A plastic hinge may form at each end of the coupling beam
following a moment-rotation behavior defined by Park and Ang which accounts for both strength
and stiffness degradation (Park and Ang, 1985). The strength and ductility parameters for the
coupling beam were determined by modeling the section in XTRACT. The behavior of the coupling
beam in LS-Dyna was then verified by comparing the moment-rotation relationship with that
obtained in XTRACT. ARUP has investigated whether the shear behavior of the ‘MAT_PARK_ANG’
material model in LS-DYNA is consistent with experimental results. See Figure 13 for a comparison
of these results.
Figure 13 Verification of Shear Behaviour of Coupling Beams (ARUP, 2013)
40
Figure 14 Coupling Beam Moment-Curvature Relationship
3.2.2 Elastic Elements
Elements expected (or required) not to undergo inelastic deformations were modeled as elastic. This
includes the gravity columns, the shear resistance of the core walls, the floor slabs and basement
walls. Effective stiffness properties were adopted using values from the DBE elastic building model
(see Table 9). The assumption of elastic behavior was verified by checking the demand in these
elements.
3.2.3 Interstory drifts in tall buildings
When considering the damage experienced by a structure during an earthquake, it has been shown
that rigid body rotation of structural elements does not result in damage (Bertero et al., 1991 and
Willford et al., 2008). Instead, shear/racking deformation of the structure is responsible for causing
damage and often becomes significant for non-structural elements in tall, flexible buildings.
A non-linear static pushover analyses was performed on the 3D representation of the fixed-base
superstructure to approximate the first mode shape, which is related to the largest story drift
demands imposed on the building. Lateral load was applied to the structure incrementally and the
-2500000
-2000000
-1500000
-1000000
-500000
0
500000
1000000
1500000
2000000
2500000
-0.03 -0.02 -0.01 0 0.01 0.02 0.03
Mxx
, S
tro
ng
Axi
s (N
-m)
Theta (rad)
Moment-rotation of EW coupling beam L5-12
DYNA output XTRACT output Design Mp
41
panel shear deformations were determined at several limit states. The shear deformations
experienced by the “leading” panel were approximately equal to the story drift. The “trailing” panel,
on the other hand, distorts more as the trailing edge of the core wall lengthens in tension. We found
that the interior panels experience shear distortions equal to 1.5 to 2 times the story drift. This varies
along the height of the building and depends on the location of the panel and the direction of the
lateral load. The racking drifts were explicitly calculated at each time step and these amplified story
drifts were used to assess the damage to all components except for the façade and core
walls/coupling beams since these experienced deformations equal to the story drift of the building.
Figure 15 shows an illustration of the racking deformations experienced by a floor.
Figure 15 Illustration of Racking Deformation
3.3 Results
3.3.1 Fixed Base Building Results – 475 year return period
A design level earthquake shaking analysis (475 year return period) is undertaken on the fixed base
building in order to benchmark the performance of tall buildings in San Francisco designed to the
PEER TBI (2010) guidelines. The response of the fixed-base building to the 20 ground motion pairs
is shown in Figure 17 - Figure 22 and Table 16. X and Y building directions are defined in Figure 16.
42
Figure 16 Principal Building Directions
Figure 17 RP475 Fixed Base Story Drift Ratios (mean shown in black)
Figure 18 RP475 Fixed Base Racking Drift Ratio (mean shown in black)
0 0.005 0.01 0.015 0.02 0.025 0.03−20
0
20
40
60
80
100
120
RP475: X−Direction Story Drift Ratio
SDR (%)
Hei
ght (
m)
0 0.005 0.01 0.015 0.02 0.025 0.03−20
0
20
40
60
80
100
120
RP475: Y−Direction Story Drift Ratio
SDR (%)
0 0.01 0.02 0.03 0.04 0.05−20
0
20
40
60
80
100
120
RP475: X−Direction Racking Drift Ratio
SDR (%)
Hei
ght (
m)
0 0.01 0.02 0.03 0.04 0.05−20
0
20
40
60
80
100
120
RP475: Y−Direction Racking Drift Ratio
SDR (%)RDR (%)
43
Figure 19 Comparison of SDR and Racking Drifts
For the fixed base building at the 475-year return period, the racking drifts were found to be up to
1.85 times larger in the x-direction and over 2 times larger in the y-direction. This highlights the
importance of considering the racking deformations in a building if a loss analysis is to be
undertaken, since some of the building components, such as the wall partitions, will experience
larger shear deformations than those indicated by the story drift. The peak floor accelerations in the
fixed base building were found to be on the order of 0.3g, which is consistent with the response
spectrum of the 475-year ground motions and consistent with the peak floor accelerations reported
in the PEER TBI study (Moehle et al., 2011).
Figure 20 RP475 Fixed Base Peak Floor Acceleration
0 0.005 0.01 0.015 0.02 0.025 0.03−20
0
20
40
60
80
100
120
RP475: X−Direction Drift Ratios
SDR (%)
Hei
ght (
m)
0 0.005 0.01 0.015 0.02 0.025−20
0
20
40
60
80
100
120
RP475: Y−Direction Drift Ratios
SDR (%)
Median Story DriftMedian Racking Drift
0 0.2 0.4 0.6 0.8 1−20
0
20
40
60
80
100
120
RP475: X−Direction Peak Floor Acceleration
PFA (g)
Hei
ght (
m)
0 0.2 0.4 0.6 0.8 1−20
0
20
40
60
80
100
120
RP475: Y−Direction Peak Floor Acceleration
PFA (g)
44
Figure 21 RP475 Fixed Base Peak Coupling Beam Rotation
Figure 22 RP475 Fixed Base Story Shear Force
Table 16 Median Peak Overturning Moment
X-Dir Y-Dir Peak Overturning Moment (GN.m) 2.13 2.15
The peak overturning moment report herein is approximately double the overturning moment
reported in the PEER report (Moehle et al., 2011). However, the report does not undertake NLRHA
at DBE level and therefore, the results are the product of a response spectrum analysis using an R-
factor of 6. The peak overturning moment at the MCE-level analysis is consistent with the
0 1 2 3 4 5 6−20
0
20
40
60
80
100
120
RP475: X−Direction Peak Coupling Beam Rotation
Coupling Beam Rotation (rad)
Hei
ght (
m)
0 1 2 3 4 5 6−20
0
20
40
60
80
100
120
RP475: Y−Direction Peak Coupling Beam Rotation
Coupling Beam Rotation (rad)
0 0.05 0.1 0.15 0.20
20
40
60
80
100
120
RP475: X−Direction Peak Story Shear Force
Story Shear Force (Fraction of Building Weight)
Hei
ght (
m)
0 0.05 0.1 0.15 0.20
20
40
60
80
100
120
RP475: Y−Direction Peak Story Shear Force
Story Shear Force (Fraction of Building Weight)
45
overturning moment reported for the NLRHA in the PEER report. The effective height of the building
(M/V) is calculated to be 53m, approximately 40% of the building height, which is in line with what
would be expected.
The median peak story drift ratio in the weak direction was approximately 1.5% in the flexible
direction and about 1% in the stiff direction. The ratio of peak x and y direction story drift ratios are
consistent with those reported in the PEER report for the original building design (~1.9% and 1.1%)
(PEER, 2012). The rotation in the coupling beams at the 475-year return period are greater than the
rotations found in the PEER report for the 2475-year return period (which are ~3.0%). The coupling
beam design in the present study involved several iterations. It was found that, when the
demand/capacity ratio of the coupling beams was as close as possible to the limit of 1.5 (at the SLE
hazard level), the core walls were less likely to develop net tension forces at the base. For this
reason, the coupling beams were deliberately designed with a demand/capacity ratio close to 1.5,
which explains why the fixed base building described herein shows larger coupling beam rotations.
The results of the NLRHA for each of the structural schemes for each of the five ground motion
intensities are presented in Figure 23 to Figure 35.
3.3.2 Peak Story Drift Ratio
Figure 23 Fixed Base Design - Peak Story Drifts at Five Ground Motion Intensities (rad)
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 5 10x 107
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 5 10x 107
0
20
40
60
80
100
120
0 0.01 0.02 0.03−20
0
20
40
60
80
100
120
0 0.01 0.02 0.03−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 5 10x 107
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 5 10−20
0
20
40
60
80
100
120
0 5 10x 107
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 5 10−20
0
20
40
60
80
100
120
0 5 10x 107
0
20
40
60
80
100
120
Hei
ght (
m)
46
Figure 24 Outrigger Design - Peak Story Drifts at Five Ground Motion Intensities (rad)
Figure 25 Isolated Design - Peak Story Drifts at Five Ground Motion Intensities (rad)
Figure 26 SDR versus Spectral Acceleration
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 5 10x 107
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 5 10x 107
0
20
40
60
80
100
120
0 0.01 0.02 0.03−20
0
20
40
60
80
100
120
0 0.01 0.02 0.03−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 5 10x 107
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 5 10−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 5 10−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 0.01 0.02 0.03−20
0
20
40
60
80
100
120
0 0.02 0.04−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.40
0.005
0.01
0.015
0.02
0.025
0.03
0.035
0.04
0.045
0.05
Sa(T1) (g)
SDR
(rad
)
Maximum Story Drift Ratio vs Sa(T1)
Fixed BaseBase IsolatedOutrigger
Hei
ght (
m)
Hei
ght (
m)
47
Table 17 Median Peak Story Drift (%) of Alternative Designs at Five Ground Motion Intensities
FLE SLE DBE MCE VRE Fixed Base 0.21 0.50 1.49 2.87 3.47 Damped Outrigger
0.18 0.50 1.39 2.07 3.01
Base Isolated 0.12 0.34 1.04 1.90 2.65
The interstory drifts experienced by a structure are good indicators of building damage – since most
of the building value is made up of drift-sensitive components. The fixed base building is found to
experience the largest story drifts at all hazard levels. At the MCE level, the peak median story drift of
the fixed base building is 2.87% and at the SLE the peak median story drift is 0.5%. This indicates
that the fixed base building has met the drift criteria in the PEER Tall Building Guidelines (PEER,
2012). The outrigger building and the base isolated building show superior drift performance at all
hazard levels, although the base isolated building exhibits consistently lower drifts that the other two
alternatives.
3.3.3 Peak Story Racking Deformation
Figure 27 Fixed Base Peak Racking Drifts (rad) at Five Ground Motion Intensities
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
Hei
ght (
m)
48
Figure 28 Outrigger Peak Racking Drifts (rad) at Five Ground Motion Intensities
The large spike in racking drift ratio is an artifact of the cross-section chosen to calculate the racking
drifts in this building. At the basement level, the vertical displacements of the section are somewhat
distorted by movement in the floor slab. Ideally, the nodes in cross-section should be chosen such
that the nodes coincide with a column.
Figure 29 Base Isolated Peak Racking Drifts (rad) at Five Ground Motion Intensities
Table 18 Median Peak Story Racking Deformation (%)
FLE SLE DBE MCE VRE Fixed Base 0.42 0.74 2.77 5.10 5.99 Damped Outrigger
0.19 0.71 2.66 3.98 4.63
Base Isolated 0.18 0.41 1.98 2.96 3.21
0 0.02 0.04−20
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04−20
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 5 10−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 5 10−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.005 0.01−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 0.01 0.02 0.03−20
0
20
40
60
80
100
120
0 0.02 0.04−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
Hei
ght (
m)
Hei
ght (
m)
49
The racking drifts represent the shear deformations experienced by several non-structural
components in the building, such as the partitions and the floor slabs. For all building models, the
racking drifts were found to be significantly larger than the story drift ratios (up to double). In general,
this amplification increases with building height, which is to be expected, due to the elongation of
the core walls.
3.3.4 Peak Story Floor Acceleration
Figure 30 Fixed Base Peak Floor Acceleration (g)
Figure 31 Outrigger Peak Floor Acceleration (g)
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.5 1−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
40
60
80
100
120
0 1 2 3x 108
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 0.02 0.04 0.06−20
0
20
40
60
80
100
120
0 1 2−20
0
20
40
60
80
100
120
0 2 4 6−20
0
20
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Figure 32 Base Isolated Peak Floor Acceleration (g)
Table 19 Median Peak Floor Acceleration (g)
FLE SLE DBE MCE VRE Fixed Base 0.19 0.34 0.43 0.88 1.11 Damped Outrigger
0.18 0.31 0.41 0.79 1.02
Base Isolated 0.12 0.21 0.25 0.52 0.69 3.3.5 Coupling Beam Rotation
Figure 33 Fixed Base Peak Coupling Beam Rotation (rad)
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51
Figure 34 Outrigger Peak Coupling Beam Rotation (rad)
Figure 35 Base Isolated Peak Coupling Beam Rotation (rad)
Table 20 Median Peak Coupling Beam Rotation
FLE SLE DBE MCE VRE Fixed Base 0.19 1.06 3.42 5.74 6.30 Damped Outrigger
0.18 0.89 2.34 5.55 6.11
Base Isolated 0.12 0.50 2.10 4.16 4.98
3.3.6 Story Shear Force
Table 21 Median Base Shear Force
FLE SLE DBE MCE VRE Fixed Base 20.4 MN 34.8 MN 46.5 MN 50.3 MN 52.4 MN Damped Outrigger
20.7 MN 33.4 MN 43.2 MN 48.9 MN 51.7 MN
Base Isolated 16.2 MN 26.2 MN 27.3 MN 32.4 MN 38.5 MN
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52
3.3.7 Peak Base Overturning Moment
Table 22 Median Peak Base Overturning Moment
FLE SLE DBE MCE VRE Fixed Base 0.86 GN.m 1.55 GN.m 2.13 GN.m 3.01 GN.m 3.26 GN.m Damped Outrigger
0.72 GN.m 1.50 GN.m 2.11 GN.m 2.78 GN.m 3.46 GN.m
Base Isolated 0.54 GN.m 0.95 GN.m 1.76 GN.m 2.34 GN.m 2.75 GN.m
3.4 Collapse Safety
Building collapse is the leading cause of earthquake related casualties. Code provisions accept that
a 10% probability of collapse during the MCE event is acceptable, which was met by all building
designs. In fact, no building was found to collapse, even during OVE analysis. This may be due to
limitations of the analysis model to simulate deterioration leading to collapse and to the fact that the
analyses were all done assuming median model parameters without consideration of modeling
uncertainty. In the cost-benefit analyses, collapse is assume to occur for cases that exceeded 5%
peak story drift, presuming that large degradation may occur by that point which is not captured in
the median analysis model. The PEER tall building guidelines indicate that the maximum transient
drift should be less than 4.5% in order to prevent collapse. A value of 5% has been used in this
study as a conservative estimate so as not to overestimate the cost of casualties. Using this 5% drift
criteria, the numbers of collapsed cases are shown in Table 23. Considering that there are Y ground
motions at each intensity, the one collapse for the MCE case is within the 10% limit, and the three
collapses at VRE represent a collapse rate of ~15 %.
Table 23 Number of Ground Motions where SDR exceeds 5%
FLE SLE DBE MCE VRE Fixed Base - - - 1 3 Damped Outrigger
- - - - 3
Base Isolated - - - - -
For each building collapse, the number of casualties is expected to be 10% of the total occupant
load. The total occupant load is based on an average occupancy of 2.11/1000sf (ATC58, 2012).
53
Therefore, the total occupant load for the building is assumed to be 1027 persons. The expected
annual loss of life for the fixed base building was determined by fitting a piecewise cubic spline
through the median hazard-based losses and numerically integrating the curve with the hazard curve
for the site. The expected annual loss of life for the Outrigger building was assumed to be 75% of
the loss associated with the fixed base building because only one data point was available. The
annualized mortality rates are presented in Table 24.
Table 24 Annual Rate of Mortality
Annual Rate of Mortality
Fixed Base 0.37 Damped Outrigger 0.28 Base Isolated Negligible
54
4 LOSS ASSESSMENT
4.1 Methodology
Direct financial losses were estimated using PACT (FEMA P-58). Residual drifts were explicitly
considered in the loss estimates. Based on the REDi™ guidelines, a median residual drift of 0.5%
was assumed to result in the building being demolished. Smaller residual drifts (0.2%) may cause
elevators to stop functioning, however, all median residual drifts were smaller than 0.2%.
4.2 User-Defined Fragilities
4.2.1 Curtain Wall
The REDi™ guidelines require that facades and curtain walls are designed and tested to
accommodate relative displacements such that connections remain elastic and the building
envelope remains effective in preventing air and water intrusion. This is in recognition that
compromising the weather-tightness of the building could affect occupant comfort and thermal
conditions and that functionality of the building could be impaired because of leaking (which could
cause mold).
The information in this section is based on conversations with various curtain wall vendors (personal
comm. John Fulton (Enclos), Vincent Polhemus (Walters and Wolf) and Pat Murray (Kawneer). The
façade system for the archetype building is assumed to be a custom-designed wet-glazed unitized
curtain wall; it is normal for the system to be custom designed for a project of this size. Wet glazing
refers to the silicone-based sealant that is used to provide air/water tightness. The sealant also
functions as a structural adhesive, bonding the glass panel to an aluminum frame. The aluminum
frame would most likely be manufactured from custom dies created for the project. The standard
cost for this type of curtain wall is roughly $110-$120/sf.
The performance of the unitized system depends on the design specifications. The baseline façade
system will experience median water/air intrusion damage around a 1% interstory drift, which is
based on consultation with the vendors on best-practice, and the guidelines in AAMA 501.4 (2009).
55
It was assumed, in consultation with vendors, that an enhanced design could accommodate 2%
interstory drift and would incur no additional cost since it is custom-designed anyway. It would
incorporate larger panel spacing, slip anchors, and deeper pockets to allow for additional
movement. Tearing of the structural silicone sealant is the damage mode limiting the water/air
intrusion performance. After an earthquake, the facade will need to be checked for the water/air
intrusion damage state. Should this damage state occur, the units themselves will most likely require
re-alignment. Re-aligning panels would roughly take 8 man hours per panel, at a labor cost of
approximately $115/hour. This has been reflected in our repair time and the repair cost
consequence functions in PACT. The enhanced curtain wall fragility is shown in Figure 36.
Compared to conventional curtain wall designs, the enhanced design is expected to delay the
initiation of damage state 1 (gasket seal failure) by 1% and damage state 2 (glass dislodgement) by
1%.
Figure 36 Enhanced Curtain Wall Fragility
4.2.2 Core Walls
Fragility functions for reinforced concrete coupled core walls are under-researched and the limited
data available – mostly based on scale tests of planar uncoupled walls- shows large variability. PACT
relates damage states to various measures of peak drift ratio or “hinge rotation” at the base of a wall.
Design practice places reliance on the performance of compression reinforcement and ACI-
0 0.01 0.02 0.03 0.04 0.05 0.06 0.07 0.08 0.09 0.10
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56
compliant confined concrete to control compression failure, but recent (unpublished) analytical and
experimental studies have indicated serious concerns over the performance of gravity supporting
walls after the point the cover concrete starts to spall in gravity plus seismic flexural compression.
Since high-rise towers often employ high-strength concrete, which often exhibit less ductility than
normal-strength concrete after reaching the peak unconfined stress, spalling may occur directly
beyond the strain at peak strength of unconfined concrete. Once this state is reached, rebar
buckling (particularly in cases where only every other bar is tied) and extensive compression damage
requiring extensive repair is likely to occur. It therefore seems prudent to keep concrete core walls
elastic and/or employ reinforcing detailing which minimizes rebar buckling and maintains the integrity
of the confined core by reducing the maximum spacing between horizontal and vertical ties.
As noted above, the assessment of core wall damage using fragility curves is likely unreliable and
improved methods should be considered. The core walls in our study were modeled in PACT using
the default fragility curves only in the bottom story where a plastic hinge was expected to develop.
Since story drift is not an appropriate damage indicator of tall slender walls, we used the roof drift
(i.e. roof displacement divided by building height) as the demand parameter for core wall damage
We compared the results of the estimated damage from this methodology to the wall strains output
from our analysis and found that they were consistent for each case. We also modified the
consequence functions to reflect the actual core wall thicknesses used.
4.2.3 Coupling Beams
Diagonally reinforced confined coupling beams can accept significant inelastic deformation without
damage that would hinder re-occupancy. The resilience of diagonally reinforced beams is
substantially better than that of conventionally reinforced beams, and damage is usually cosmetic
(crack widths less than 1/16 in) until hinge rotations exceed 2% (when damage state 2 is initiated).
We used the default fragility curves in PACT as they provided a good indicator of the damage
expected.
57
4.2.4 Traction Elevator
The information in this section is based on conversations with various vendors and manufacturers -
personal communication with Michael Garceau (ThyssenKrupp), Ron Eddington (Eddington
Engineering), Richard Blaska (RCB Elevator Consulting) and Andrew Betzina (KONE Corporation).
Based on these conversations, typical elevators for a 42-story building will operate between 700 and
800 feet per minute and an installed cost of around $1 million each. Elevator design is regulated by
ASME A17; and, in high seismic zones, the rails are designed for a lateral acceleration of 0.5g.
The default elevator fragility in PACT, which has three simultaneous damage states, was used
except that the median acceleration causing damage (since damage states are simultaneous, one
fragility describes three possible damage states) was modified to 0.5g. The repair cost and repair
time per elevator, based on our conversations with vendors, were modified as follows:
• DS1 –Rail Distortion- $350,000 and 13.6 ‘man-days’ (79% probability)
• DS2 – Cab stabilizers bent, cab walls or doors damaged - $19,300 and 13.1 ‘man-days’ (68% probability)
• DS3 – Cab ceiling damaged - $15,000 and 5.9 ‘man-days’ (17% probability)
An additional fragility was employed to incorporate our understanding that elevators would cease to
operate if the tolerance between rail guides exceeded 0.25”. This relates to a residual drift of
approximately 0.2% (H/500 for story heights of 9’-8”), which is a commonly defined serviceability
limit state for tall buildings under wind. If rails have permanent deformation beyond the operational
tolerance, replacement/realignment of the entire rail guidance system is required. The assumed
repair cost is $1.2 million per elevator, with a replacement time of 180 days.
The enhanced elevator design of the REDi™ guidelines was evaluated using the California Office of
Statewide Health and Planning (OSHPD) requirements, found in the California Building Code (CBC,
2010). To meet these requirements, elevator could be expected to cost an additional $400,000. In
an OSHPD compliant system, the bracket spacing is halved, and the maximum allowable rail
deflections are reduced. Therefore, the damage state corresponding to rail distortion was removed
58
from our PACT model. OSHPD also requires seismic switches, which automatically shut elevators
down in the event of an earthquake, even if the elevator is undamaged. The elevator can only
resume operations once a reset button is pressed. Since this should only be done by a skilled
technician after a complete inspection of the elevator, it may be prudent for the building owner to
have the facility manager (or other personnel responsible for maintenance) certified to perform this
inspection.
4.2.5 Wall Partitions
Partitions typically represent one of the biggest contributors to financial loss in buildings (Araya-
Letelier and Miranda, 2012). This is because they are damaged at low drift levels, often represent a
good portion of the initial construction cost, and take considerably more time and cost to repair than
the initial installation cost itself, depending on the level of damage. If severely damaged, repair costs
and labor times must account for removing and re-installing or replacing wall-mounted electrical
fixtures, thermostats, telephone/internet ports, cabinets, plumbing pipes and fixtures, doors and
frames, trim, and wall finishes (personal communication with Peter Morris and Eduardo Miranda).
Carpets must be pulled back, baseboards removed, and ceilings cut to access the partition framing
and tracks. This can hinder re-occupancy as could falling hazards from heavy wall-mounted fixtures
(e.g. TVs or cabinets) hanging from distorted partitions.
For the conventionally designed buildings, default PACT fragilities were used. The default repair
costs for each damage state were left unchanged but the repair times were reduced based on the
judgment that the default repair times were too high. For the REDi™-designed buildings, a
sliding/frictional connection developed by Araya-Letelier and Miranda (2012) was used. The fragility
function in PACT was modified to delay the initiation of each damage state by 1.0% interstory drift to
reflect the enhanced performance as indicated in the Araya-Letelier and Miranda (2012) study.
59
Figure 37 Enhanced Partition Fragility
4.3 Loss Assessment Results
Table 25 shows the median estimated direct financial losses obtained from PACT for each of the
building designs studied. Figure 38 - Figure 40 provides a breakdown of the losses, comparing the
absolute losses attributed to specific component types for each building design. Components
contributing significantly to the financial loss are the interior partitions, coupling beams, floor slab
connections and curtain walls.
4.3.1 Summary of Losses
Table 25 Summary of Loss – % Building Value
PBD FLE SLE DBE MCE VRE Fixed Base 2.7 6.2 15.2 30.9 34.3 Damped Outrigger
0.5 4.2 13.6 23.9 28.1
Base Isolated 0.4 2.3 10.1 17.9 20.9 REDi
FLE SLE DBE MCE VRE Fixed Base 0.6 1.2 8.1 27.5 32.0 Damped Outrigger
<0.1 <0.1 7.0 19.1 23.9
Base Isolated <0.1 <0.1 2.4 10.5 13.0
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60
4.3.2 Breakdown of Losses
Figure 38 Fixed Base: Breakdown of Financial Losses
Figure 39 Outrigger: Breakdown of Financial Losses
!$#!!
!$10,000,000!!
!$20,000,000!!
!$30,000,000!!
!$40,000,000!!
!$50,000,000!!
!$60,000,000!!
!$70,000,000!!
PBD! REDi! PBD! REDi! PBD! REDi! PBD! REDi! PBD! REDi!
RP31! RP135! RP475! RP2475! RP4975!
Other! Par::ons!
Curtain!Wall! Coupling!Beams!
Floor!Slabs! Core!Walls!
!$#!!
!$10,000,000!!
!$20,000,000!!
!$30,000,000!!
!$40,000,000!!
!$50,000,000!!
!$60,000,000!!
!$70,000,000!!
PBD! REDi! PBD! REDi! PBD! REDi! PBD! REDi! PBD! REDi!
RP31! RP135! RP475! RP2475! RP4975!
Other! Par::ons!
Curtain!Wall! Coupling!Beams!
Slabs! Core!Walls!
61
Figure 40 Base Isolated: Breakdown of Financial Losses
4.4 Expected Annual Loss
The expected annual loss was determined by fitting a piecewise cubic spline through the median
hazard-based losses and numerically integrating the curve with the hazard curve for the site. The
annualized losses are presented in Figure 25.
Table 26 Expected Annual Loss
PBD REDi Fixed Base $660,000 $1.01/sf $252,000 $0.39/sf Damped Outrigger $236,000 $0.36/sf $155,000 $0.24/sf Base Isolated $221,000 $0.34/sf $48,000 $0.07/sf
4.5 Conclusions
The results of the loss analysis highlight a number of important conclusions. Firstly, it is interesting to
note that the losses are more divergent at lower hazard levels, and begin to converge at higher
!$#!!
!$10,000,000!!
!$20,000,000!!
!$30,000,000!!
!$40,000,000!!
!$50,000,000!!
!$60,000,000!!
!$70,000,000!!
PBD! REDi! PBD! REDi! PBD! REDi! PBD! REDi! PBD! REDi!
RP31! RP135! RP475! RP2475! RP4975!
Other! Par::ons!
Curtain!Wall! Coupling!Beams!
Slabs! Core!Walls!
62
levels as the damage to the building saturates. The large divergence at lower hazard levels leads to
the vast range of annualized losses (from $48,000 to $660,000).
Compared to the results of the PEER study, the annualized losses of the baseline case presented
herein are approximately double those reported by PEER. This is likely due to the difference is loss
estimation methodology, since the PEER study uses component group fragilities rather than
individual component fragilities.
The differences in the losses for the PBD building and the REDi building can largely be attributed to
the damage to partitions. This is consistent with the findings of other researchers (Miranda, 2010)
and highlights the role of partition design in damage avoidance design. It is important to note that
the partition fragilities included in the REDi designed buildings have not been tested extensively
enough for their use to be permitted by local regulations.
63
5 DOWNTIME ASSESSMENT
5.1 Methodology
5.1.1 Summary of REDi Downtime Methodology
Although FEMA P-58 provides estimates of repair time due to earthquake damage, it does not
calculate the total building downtime, which may be much longer than the repair time. There are
several significant limitations to the FEMA P-58-based assessment in relation to calculating
downtime that must be addressed:
- The repair time estimates are based on potentially unrealistic labor allocation and repair
sequence logic.
- Repair time estimates are associated with the time required to achieve full recovery.
However, most owners are primarily concerned with the time required to re-occupy the
building and/or the time required to regain functionality.
- FEMA P-58 does not account for delays that prevent the initiation of repairs, (‘impeding
factors’ such as the time it takes to inspect the building, access financing, find and mobilize
contractors/engineers, and obtain permitting) which could represent the largest contributor
to downtime.
- FEMA P-58 does not account for the disruption to utilities.
The methodology described in the REDi™ guidelines attempts to address these limitations by
building on the FEMA P-58 damage state and repair time estimates as a basis for predicting
downtime. Specifically, the enhancements provided by the REDi™ guidelines include:
- Definition of “Repair Classes” which describe whether the extent of damage to and criticality
of various building components will hinder achievement of specific recovery states like re-
occupancy, functional recovery, and full recovery.
64
- A modified approach for allocating labor and sequencing repairs based on data from RS
Means and anecdotal evidence from contractors and cost estimators.
- Estimates of delays to initiation of repairs (“impeding factors”) based on lessons from past
natural disasters and expert opinion.
- Estimates of utility disruption for electricity, water, and gas based on data from past
earthquakes and predicted regional disruptions for hypothetical future earthquake scenarios
published by experts.
- Sequential logic for calculating the time required to achieve re-occupancy, functional
recovery and/or full recovery due to “impeding factors”, utility disruption, and building
repairs (i.e. these must be considered in the order they will be initiated and completed).
REDi™ downtime estimates are the sum of the delay time (before repairs begin) due to impeding
factors and the building repair time. The impeding factors are controllable in the sense that they can
be mitigated by having in place contingency planning measures such as having qualified
professionals to inspect the building on retainer, quick access to financing, and contractors and
engineers on retainer. The REDi™ guidelines provide estimates of the delay time for each of the
impeding factors, depending of the level of structural/non-structural damage that is predicted from
PACT.
PACT assumes that building repairs are either made one floor at a time, or all floors at the same
time. Clearly this is not realistic for tall buildings. The REDi™ downtime repair time estimates are
determined by creating a logical schedule of repairs. Repairs are broken up into 7 categories;
structural, interior repairs, exterior repairs, mechanical, electrical, elevator and stair repairs where
each category has limits on the number of workers allocated to the repairs in that category. The
structural repair sequence begins first, starting at the ground floor and moving upward. All other
categories of repair follow in parallel but must be at least three floors behind the structural repairs.
65
The repair sequence for each of the categories is created from the number of man-hours (estimated
by PACT) required to conduct repairs on each floor.
Repair Classes are used to describe the severity of damage to each of the building’s components
and reflect the criticality of those components. Structural or non-structural components that suffer
only minor damage that would only hinder full recovery (e.g. cracked partitions or minor cracking of
concrete elements) are in Repair Class 1. Damaged non-structural components that do not pose a
“life-safety” risk (e.g. inoperable mechanical, electrical, and plumbing equipment and other MEP
services) but would hinder functionality are in Repair Class 2. Heavily damaged structural or non-
structural components that pose a “life-safety” hazard and would prevent re-occupancy are in
Repair Class 3. To achieve full recovery, all components in Repair Class 1, 2, and 3 must be
repaired. To achieve functional recovery, all components in Repair Class 2 and 3 only must be
repaired. And to achieve re-occupancy, all components in Repair Class 3 only must be repaired.
Thus the downtime corresponding to each of these distinct recovery states may be calculated.
Repair classes are determined based on the average damage state of the components on a floor. If
less than 10% of the components on a floor are damaged, the Repair class is set to zero. There are
two reasons behind this logic. Firstly, due to the lognormal distribution of damage fragilities,
component types with a large number of constituents are likely to have at least one damaged
component (giving an average damage state > 0), even at low demands. If the repair class definition
were not truncated, the median repair time would almost always be non-zero for a building with 40
stories that is comprised of many components. Secondly, when small percentages of components
are damaged and the damage is obscured by architectural components, it is possible that nothing
would be identified during an ATC20 evaluation, and therefore the damage would have no bearing
on the building damage placard (i.e., green, yellow or red).
The methodology does not attempt to quantify downtime caused by some “uncontrollable”
externalities, which include hazards from adjacent buildings, restricted site access, and availability of
66
employees to return to work. The reader is referred to the REDi™ (2013) guidelines to view the full
methodology for calculating the time required to achieve re-occupancy and functional recovery.
5.2 Downtime Assessment Results
5.2.1 Summary of Downtime
Table 27 and Table 28 show the median estimated functional and full downtime for each of the
building designs studied. In all cases, the median functional repair time was the same as the median
re-occupancy repair time (i.e. in the median case, no functionally dependent mechanical equipment
took longer to repair than the repair time to achieve re-occupancy).
Table 27 Median Repair Time (Weeks) to Achieve Functionality (Time including impeding factors in
brackets)
PBD FLE SLE DBE MCE VRE Fixed Base 0 (1) 5 (17) 72 (84) 113 (137) 125 (149) Damped Outrigger 0 (1) 4 (16) 50 (62) 86 (110) 101 (125) Base Isolated 0 (1) 0 (1) 31 (43) 65 (89) 72 (96)
REDi FLE SLE DBE MCE VRE Fixed Base 0 (1 day*) 0 (1 day*) 26 (29) 97 (104) 113 (121) Damped Outrigger 0 (1 day*) 0 (1 day*) 17 (20) 77 (84) 91 (99)
Base Isolated 0 (1 day*) 0 (1 day*) 3 (6) 44 (51) 52 (60) *Time to inspect the building
Table 28 Median Repair Time (Weeks) to Achieve Full Recovery (Time including impeding factors in brackets)
PBD FLE SLE DBE MCE VRE Fixed Base 3 (15) 12 (24) 74 (86) 126 (150) 138 (162) Damped Outrigger 2 (14) 9 (21) 65 (77) 97 (121) 112 (136)
Base Isolated 0 (1) 6 (18) 39 (51) 83 (107) 89 (113) REDi
FLE SLE DBE MCE VRE Fixed Base 1 (4) 6 (9) 47 (50) 114 (121) 128 (136) Damped Outrigger 0 (1 day*) 0 (1 day*) 39 (42) 87 (94) 103 (112)
Base Isolated 0 (1 day*) 0 (1 day*) 20 (23) 47 (54) 67 (75) *Time to inspect the building
67
The fixed based PBD building performed the worst with 1.5 years of downtime following a design
level event. Following the MCE and VRE, the downtime for the PBD fixed base building is well over 2
years, which may be long enough to favor rebuild over repair for some building owners. Even the
PBD base isolated building, which suffered the least damage, is not expected to regain functionality
for almost one year after the DBE. The REDi buildings, with enhanced non-structural design,
showed significant reductions in building downtime, especially for low-level hazards. This is primarily
due to the enhanced partition design, as the partitions in the PBD building became damaged at low
levels of drift, hindering immediate functional recovery.
These results highlight the importance of limiting both structural and non-structural damage if timely
recovery is desired. Given that the financial losses and building downtime are driven by a small
number of components, the effect of enhancing these components is significant.
5.3 Expected Annual Downtime
The expected annual downtime was determined using the same method as for the expected annual
loss. A piecewise cubic spline was fit through the median hazard-based functional downtime and
numerically integrating the curve with the hazard curve for the site. The annualized downtime values
are presented in Table 29. The expected annual downtime, like the expected annual loss, is useful in
estimating the cost of insurance for business interruption.
Table 29 Expected Annual Time that Building is not functional (days)
PBD REDi Fixed Base 4.3 2.1 Damped Outrigger 3.9 1.3 Base Isolated 2.8 0.1
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6 COST-BENEFIT ANALYSIS
Cost benefit analysis provides a methodology to evaluate whether or not enhanced seismic design
of buildings makes financial sense to building stakeholders. The expected benefits of enhanced
seismic design include reduced repair cost, reduced closure time and reduced incidents of injury
and/or death caused by building damage. The realization of these benefits will only occur if a
damaging earthquake strikes the building, which may never happen in the life of the building, or may
occur in the near future. In contrast, the additional cost of design improvements must be paid up
front, without any certainty that a return on the investment will eventuate.
In the San Francisco Planning for Urban Resilience (SPUR) report, ‘Building it right the first time’
(SPUR, 2009), the authors note that, although older buildings are more likely to be damaged in an
earthquake, the cost of retrofit is significantly larger than the marginal cost of improving seismic
safety and “thus has a relatively good cost-benefit ratio’. From the perspective of building owners,
considering the potential for enormous loss of profit due to building closure, it is hypothesized that a
relatively small investment in improved seismic performance will be outweighed by the potential
future benefits. The size of the investment deemed ‘worthwhile’ will depend on the time-horizon
considered and that rate at which future benefits are discounted. For a building owner who intends
to sell the building shortly after it is built, the size of the time-horizon is likely to preclude justification
of any investment in seismic safety. In contrast, when the entire lifetime of the building is considered,
larger margins of investment may be justified.
6.1 Methodology
Cost benefit analysis has a number of discrete steps, which enable the analyst to convey
assumptions in a transparent manner. The major steps of a CBA (Boardman, 2011) are as follows:
1. Specify the set of alternative projects
2. Decide whose benefits and costs count
3. Identify the impact categories, catalogue them, and select measurement indicators
69
4. Predict the impacts quantitatively over the life of the project
5. Monetize all impacts
6. Discount benefits and costs to obtain present values
7. Compute the net present value of each alternative
8. Perform Sensitivity analysis
9. Make a recommendation
6.1.1 Alternatives Considered
Three different structural design alternatives and two non-structural design alternatives are
considered. For details of the design of these buildings refer to chapter 2. Each design was
compared to the fixed base building with code conforming non-structural design (the status quo). A
total of five design alternatives will be compared to the status quo (fixed base design).
6.1.2 Perspective of the Analysis
The perspective of the analysis will be that of the building owner. In the absence of more stringent
regulations or financial incentives from a governing body, it is likely that the adoption of enhanced
performance objectives will only occur if a compelling financial argument can be made to the
person/s funding the project. The cost-benefit analysis is intended to determine the conditions under
which it makes economic sense to provide enhanced performance, and how robust the benefit-cost
(B-C) ratio is to variations in these conditions.
6.1.3 Costs and Benefits Considered
The benefits of the enhanced designs are quantified in terms of the reduction in expected building
repair cost, downtime and mortalities/fatalities. The expected annual loss is calculated as the sum of
the expected annual losses from deaths, damage and downtime as follows;
𝐸𝑥𝑝𝑒𝑐𝑡𝑒𝑑 𝐴𝑛𝑛𝑢𝑎𝑙 𝐿𝑜𝑠𝑠 = 𝐸 𝑑𝑖𝑟𝑒𝑐𝑡 𝑙𝑜𝑠𝑠 + 𝐸 𝑑𝑜𝑤𝑛𝑡𝑖𝑚𝑒 𝑙𝑜𝑠𝑠 + 𝐸(𝑚𝑜𝑟𝑏𝑖𝑑𝑖𝑡𝑦 𝑎𝑛𝑑 𝑚𝑜𝑟𝑡𝑎𝑙𝑖𝑡𝑦 𝑙𝑜𝑠𝑠)
70
Where each component of the expected annual loss is measured in dollar terms (see 6.1.4). The
expected annual loss of future years must be discounted to present day value, reflecting the fact
that money today is more valuable than the same amount of money next year (the time value of
money). The rate at which money is discounted may be calculated from interest rates, the
government’s borrowing rate or a weighted average, depending on the relative contribution of the
sources. The expected loss is then calculated from the annual loss, for a given time horizon and
discount rate, r;
𝐸𝑥𝑝𝑒𝑐𝑡𝑒𝑑 𝐿𝑜𝑠𝑠 = 𝐸𝐴𝐿1 + 𝑟 !
!"#$ !"#$%"&
!!!
The benefit of each alternative is calculated as the difference in the expected loss, relative to the
status quo.
𝐵𝑒𝑛𝑒𝑓𝑖𝑡 = 𝐸𝑥𝑝𝑒𝑐𝑡𝑒𝑑 𝐿𝑜𝑠𝑠!"#$% !"#$ − 𝐸𝑥𝑝𝑒𝑐𝑡𝑒𝑑 𝐿𝑜𝑠𝑠!"!!"#$% !"#$%&
The cost of the each alternative is quantified in terms of the additional cost of construction relative to
the fixed base case. These costs must be paid up front and therefore needn’t be discounted to
present day values. A summary of the cost and benefits considered is shown in Table 30.
𝐶𝑜𝑠𝑡 = 𝐼𝑛𝑖𝑡𝑖𝑎𝑙 𝑐𝑜𝑛𝑠𝑡𝑟𝑢𝑐𝑡𝑖𝑜𝑛 𝑐𝑜𝑠𝑡!"#$% !"#$ − 𝐼𝑛𝑖𝑡𝑖𝑎𝑙 𝑐𝑜𝑛𝑠𝑡𝑟𝑢𝑐𝑡𝑖𝑜𝑛 𝑐𝑜𝑠𝑡!"!!"!"# !"#$%&
Table 30 Summary of Costs and Benefits
Costs Benefits Status Quo No net costs/benefits Alternative Designs
Increased construction costs
Reduction in repair costs Reduction in downtime Reduction in mortality/fatality
𝐵𝑒𝑛𝑒𝑓𝑖𝑡 − 𝐶𝑜𝑠𝑡 𝑅𝑎𝑡𝑖𝑜 (𝐵𝐶𝑅) = 𝐵𝑒𝑛𝑒𝑓𝑖𝑡𝐶𝑜𝑠𝑡
Whereby a scheme is determined to be worth the additional expenditure if the BCR is greater than
1.
71
6.1.4 Monetizing Costs and Benefits
Cost-benefit analysis requires that all costs and benefits be quantified in monetary terms, even if the
impact isn’t tangible. As a first pass, the CBA was conducted without consideration of the intangible
costs (such as mortality, morbidity and downtime). In order to refine the analysis, the CBA was re-
run, accounting for estimations of intangible costs.
Valuing human life is often a point of contention among analysts, firstly because of people who argue
that a life should not be measured in monetary terms and secondly, because the way in which the
value of a life is measured will have a large effect on the magnitude of the value. The US Department
of Transportation currently uses a value of statistical life (VSL) at $9.1 million dollars and the
Environmental Protection Agency (EPA) currently uses a VSL of $6.3 million dollars (White House,
2013). This study will use a VSL of $7 million USD, reflecting the range of values used in US
Departments between $5-9 million.
The value of morbidity or injury is arguably even harder to define than the VSL because the methods
used to calculate the rate of injuries is uncertain and nature of the injury is unknown. A study on the
cost of non-fatal injuries in the 1994 Northridge Earthquake estimate the cost of a non-fatal injury to
be, on average, between $8,050 and $13,850 when converted to 2014 dollars (Porter, 2005). The
expected cost of an injury used in this study is $10,950. The number of injuries was found to be
7,460 times the number of deaths, although, for this study, the use of a multiplication factor of 7,460
is unreasonable, given the total number of occupants in the building. Instead, it is assumed that the
number of injuries is 5x the number of deaths (if the building collapses, 10% of occupants are
assumed to die and 50% are assumed to be injured).
Also considered in this analysis is the cost of downtime. Typically, downtime is not converted into
monetary terms, because it is building specific. In this CBA, the cost of downtime will be however, in
order to complete a CBA on building design, the cost of downtime must be included, regardless of
how difficult it may be to quantify. The expected cost of downtime is based on the loss of rental
income per square foot, per month. As of November, 2014, the average rental price for properties in
72
San Francisco is $3.55/sf (Trovit, 2014). This is likely a conservative estimate, given that the building
used in this study is located in an area where rental prices are above average for the city. The
sensitivity analysis will consider a range of costs, from $2/sf to $5/sf.
6.1.5 Time Horizon
The time horizon to be used in the analysis has a significant effect on the outcome. Perhaps the
most straightforward estimate of a time horizon is 50-years, since this is the design life of a modern
building, however, for tall buildings, it is likely that they will not be replaced until well over 50-years
has past. While this supports the use of a larger time horizon, it is also important to consider the
perspective of the analysis. From a societal point of view, a time horizon of 100-years may be
appropriate, but for building owners/investors, the time horizon that is relevant is simply the time for
which a building is owned. The sensitivity analysis will consider time horizons ranging from 5 years to
100 years. This way, it is possible to calculate the time horizon for which the B-C ratio transitions
from below 1 to above 1. An expected time-horizon of 25 years is considered.
6.1.6 Discounting Future Costs and Benefits
Typically, the social discount rate (SDR) used in CBAs is between 3% and 6% (White House, 2013).
This is referred to as the real discount rate because it has already been adjusted to account for
inflation. The expected discount rate considered for the analysis is 4.5% and the range of discount
rates considered in the sensitivity analysis is 3-6%.
6.2 Results
6.2.1 Summary of Costs
As previously mentioned, the cost of each alternative is calculated to be the additional cost of
construction, relative to the code conforming, fixed base case. The construction costs for each of
the alternatives is summarized in Table 31 and the difference in these costs is summarized in Table
32.
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Table 31 Construction Costs
PBD REDi Fixed Base $176,046,466 $177,957,420 Damped Outrigger $177,802,532 $179,713,486 Base Isolated $178,475,943 $180,386,898
Table 32 Summary of Cost Premium Above the Benchmark Fixed Based Design
PBD REDi Fixed Base $- $1,910,954 Damped Outrigger $1,756,066 $3,667,020 Base Isolated $2,429,478 $4,340,433
6.2.2 Summary of Benefits, Considering Repair Cost Only
Using the expected values of 4.5% discount rate and 25-year time horizon, the benefits of each
design alternative is calculated. The expected loss is then calculated from the expected annual loss.
A summary of the expected losses is shown in Table 34. The benefit of each of the design
alternatives is the decrease in expected loss relative to the code conforming fixed base case (see
Table 35).
Table 33 Expected annual loss from building repair
PBD REDi Fixed Base $660,000 $252,000 Damped Outrigger $236,000 $155,000 Base Isolated $221,000 $48,000
Table 34 Expected loss for each design alternative
PBD REDi Fixed Base $9,780,000 $3,730,000 Damped Outrigger $3,500,000 $2,290,000 Base Isolated $3,280,000 $710,000
74
Table 35 Summary of Benefits
PBD REDi Fixed Base - $6,050,000 Damped Outrigger $6,290,000 $7,490,000 Base Isolated $6,510,000 $9,080,000
6.2.3 Summary of Benefit-Cost Ratios Considering Repair Costs Only
The benefit-cost ratio is a measure of the relative size of costs and benefits. If the BC ratio is greater
than 1, the benefits of an alternative are found to outweigh the additional costs. Conversely, a BC
ratio less than 1 represents an alternative where the benefits are not significant enough to make up
for the additional cost. A negative BC ratio indicates that either the alternative has positive costs and
negative benefits (an alternative not worth pursuing) or negative costs and positive benefits (a very
attractive alternative). A summary of the BC ratios are presented in Table 36.
Table 36 Summary of Benefit-Cost Ratios
PBD REDi Fixed Base 3.2 Damped Outrigger 3.6 2.0 Base Isolated 2.7 2.1
An illustration of the how the benefit-cost ratios change with the time horizon is presented in Figure
41. The payback period is the time at which the benefit-cost ratio exceeds 1, found to be between 4
and 9 years for the alternatives considered.
75
Figure 41 Illustration of Payback period
6.2.4 Benefit-Cost Analysis Including Downtime and Morbidity/Mortality
A summary of the expected annual downtime is presented in Table 29. The expected annual
casualties are presented in Table 24. As mentioned in a previous section, the expected number of
annual injuries is estimated to be five times the number of casualties. At a cost of $10,950 per injury,
the cost per casualty including injuries is $7.05 million USD.
The expected losses for each alternative, considering the cost of downtime and morbidity/mortality
is presented in Table 37. As for the previous section, a time horizon of 25 years is considered and a
discount rate of 4.5% is applied.
Table 37 Expected loss (including downtime only) for each design alternative
PBD REDi Fixed Base $13,773,000 $5,159,000 Damped Outrigger $7,597,000 $3,439,000 Base Isolated $6,104,000 $742,000
Table 38 Expected loss (including downtime and mortality/morbidity) for each design alternative
PBD REDi Fixed Base $54,542,000 $45,929,000 Damped Outrigger $35,822,000 $31,664,000 Base Isolated $6,104,000 $742,000
0 5 10 15 20 250
0.5
1
1.5
2
2.5
3
3.5
4Illustration of Payback Period
Bene
fit−C
ost R
atio
Year
Fixed Base REDiOutrigger PBDOutrigger REDiIsolated PBDIsolated REDi
76
Table 39 Summary of Benefits
PBD REDi Fixed Base - $8,614,000 Damped Outrigger $18,720,000 $22,878,000 Base Isolated $48,438,000 $53,801,000
The results highlight the enormous impact of mortality and morbidity on the expected losses. Given
that the number of collapses was gauged in an approximate manner, based on the exceedence of a
given drift limit, morbidity and mortality will not be included in the sensitivity analysis.
Table 40 Summary of Benefit-Cost Ratios (including downtime and mortality/morbidity)
PBD REDi Fixed Base - 4.5 Damped Outrigger 10.7 6.2 Base Isolated 19.9 12.4
6.3 Sensitivity Analysis
The purpose of the sensitivity analysis is to determine how robust the BC ratios are to the
assumption in the analysis. It is important to recognize the uncertainty in the analysis and to quantify
the effect of this uncertainty on the conclusions of the analysis.
Three methods of sensitivity analysis are typical in CBAs. The first of these is a one-way (partial)
analysis whereby the upper and lower bound of each of the parameter is estimated and the
sensitivity of the results to each parameter is estimated by varying the parameter from its upper to
lower bound, while all other parameters remain constant. This approach does not account for
simultaneous variability of parameters, however it highlights the parameters significantly affecting the
results. The second approach to sensitivity analysis is a best and worst case scenario method. Each
of the parameters is set to the bound corresponding to the worst case and then to the bound
corresponding to the best case. This method successfully binds the results of the analysis but does
not indicate the likelihood of either scenario. The third approach is probabilistic sensitivity analysis
(PSA). Here, the distribution of each of the parameters is estimated and Monte Carlo simulation is
77
used to determine the distribution of the BC ratios. This approach enables the analyst to estimate
the probability with which a BC ratio will be greater than/less than 1. Given that the distribution of
relevant parameters is very difficult to gauge, the second approach will be used in this study,
whereby the maximum and minimum parameters will be used to bound the analysis.
6.3.1 Range of Relevant Parameters
The sensitivity analysis will consider a range of discount rates, between 3% and 6%, and a range of
the cost of downtime, from $2-$5 per square foot per month. The annualized loss and downtime
values are modeled using a lognormal distribution. In order to obtain the parameters of the
lognormal distribution, the entire range of loss and downtime values obtained in Sections 4 and 5
was considered. For each hazard level, 1000 simulations were run, each yielding a loss and
downtime estimate. A lognormal distribution was fit to the loss and downtime values at each hazard
level. Figure 42 and Figure 43 show the distribution of the fixed-base results. In each case, the left-
most curves represent the 31-year return-period hazard and the right-most curves represent the
4975-year return period hazard.
Figure 42 Lognormal Curves Fit to Loss Data
0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9 10
0.1
0.2
0.3
0.4
0.5
0.6
0.7
0.8
0.9
1REDi Loss
Loss (Proportion of building value)
Prob
abilit
y of
non−e
xcee
denc
e
0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9 10
0.1
0.2
0.3
0.4
0.5
0.6
0.7
0.8
0.9
1PBD Loss
Loss (Proportion of building value)
Prob
abilit
y of
non−e
xcee
denc
e
78
Figure 43 Lognormal Curves Fit to Downtime Data
The distribution of annualized loss and downtime is calculated, as previously mentioned, by fitting a
curve through 5 points of loss/downtime data. Instead of sampling randomly from the distribution of
hazard-specific loss and downtime curves, which would result in an exaggerated dispersion of
annualized values, the values are assumed to be fully correlated. Therefore, a ‘low’ loss estimate at
one hazard level, will result in an equally ‘low’ estimate at the other hazard levels. The distribution of
annualized loss and downtime for the fixed base building is shown in Figure 44. The 50% probability
of non-exceedence values are consistent with the median results presented earlier.
Figure 44 Annual loss and Downtime Distribution
0 100 200 300 400 500 600 700 800 900 10000
0.1
0.2
0.3
0.4
0.5
0.6
0.7
0.8
0.9
1PBD Downtime
Downtime (days)
Prob
abilit
y of
non−e
xcee
denc
e
0 100 200 300 400 500 600 700 800 900 10000
0.1
0.2
0.3
0.4
0.5
0.6
0.7
0.8
0.9
1REDi Downtime
Downtime (days)
Prob
abilit
y of
non−e
xcee
denc
e
0 0.01 0.02 0.03 0.04 0.05 0.06 0.07 0.08 0.09 0.10
0.1
0.2
0.3
0.4
0.5
0.6
0.7
0.8
0.9
1Fixed Base Annual Loss
Annual Loss (fraction of building value)
Prob
abilit
y of
non−e
xcee
denc
e
PBDREDi
0 5 10 15 20 25 300
0.1
0.2
0.3
0.4
0.5
0.6
0.7
0.8
0.9
1Fixed Base Annual Downtime
Annual Downtime (days)
Prob
abilit
y of
non−e
xcee
denc
e
PBDREDi
79
In the sensitivity analysis values corresponding to the 10% and 90% annualized values are used.
These values are presented in Table 41.
Table 41 Summary of the range of Annualized Loss
PBD REDi %10 %90 %10 %90 Fixed Base $172,500 $3,320,000 $28,700 $960,000 Damped Outrigger $61,700 $1,187,000 $17,700 $590,500 Base Isolated $57,800 $1,112,000 $5,500 $182,900
Table 42 Summary of the range of Annualized Downtime (days)
PBD REDi %10 %90 %10 %90 Fixed Base 0.66 29.26 0.47 7.49 Damped Outrigger 0.60 26.53 0.29 4.63 Base Isolated 0.43 19.05 0.02 0.36
6.3.2 Time Horizon
The time horizon is not a random variable, however it will vary depending on the specific building
owner. For this reason, the sensitivity analysis is presented in terms of payback period, rather than
using distinct time horizons.
6.3.3 Results
The results of the sensitivity analysis are presented in Figure 45 and Figure 46. Looking at the lower
bound of payback, it appears that all design alternatives will be ‘paid back’ in the first year, which is
likely to be unreasonable. However, the upper-bound results indicate that, in the worst-case
scenario, all design alternatives reach a benefit-cost ratio of 1 within 35 years. Given that the design
life of a building is usually considered to be 50 years, this indicates that a building owner who
intends to keep the building for it’s useable life would be better off adopting an enhanced design
alternative.
80
Figure 45 Upper Bound of Payback
Figure 46 Lower bound of payback
6.3.4 Conclusions
The results of the CBA suggest that all design alternatives are preferential to the status quo.
Surprisingly, the B/C ratio is higher for the structural design alternatives, compared to the enhanced
non-structural design alternatives. When considering the hazard-specific losses, the REDi designs
are shown to significantly reduce losses and downtime compared to the performance-based
approach, yet the results of the CBA indicate that the investment is slower to be paid back. This
indicates that it is wise to invest small amounts of money for small improvements in performance,
0 5 10 15 20 25 30 35 40 45 500
0.5
1
1.5
2
2.5Illustration of Payback Period
Bene
fit−C
ost R
atio
Year
Fixed Base REDiOutrigger PBDOutrigger REDiIsolated PBDIsolated REDi
0 0.5 1 1.5 2 2.5 3 3.5 4 4.5 51
2
3
4
5
6
7
8
9Illustration of Payback Period
Bene
fit−C
ost R
atio
Year
Fixed Base REDiOutrigger PBDOutrigger REDiIsolated PBDIsolated REDi
81
but that there are diminishing returns as the investment increases. Future studies should investigate
the optimal level of investment in enhanced performance and a means for assessing whether a
design alternative is optimal from a financial standpoint.
82
7 CONCLUSIONS
Earthquake resilience of tall buildings is important because damage causing downtime could
disproportionately affect the local economy and potentially displace thousands of residents. The
findings of this research suggest that the seismic performance of tall buildings in San Francisco,
though unlikely to cause fatalities, may lead to significant financial losses and building downtime that
could have a profound impact on the overall resilience of the city.
The expected performance of an archetype, 42-story coupled core wall residential building, located
in downtown San Francisco is evaluated. The expected direct financial loss caused by seismic
damage following a 475-year return period earthquake event is estimated to be more than 15% of
the total cost of the building. The expected time required before the building restores functionality is
on the order of 1.5-years, which includes the time required to inspect the building, obtain financing
and permits for repairs and to carry out repair work.
Two structural design alternatives, damped outriggers and base isolation, are evaluated. Both the
damped outrigger option and the isolated building are effective at limiting the drifts experienced by
the structure. The isolated building in particular was found to significantly reduce the expected direct
financial loss and downtime following a 475-year return period earthquake event to approximately
10% of the total building cost.
Substantially improved performance of the building is achieved following the resilience-based design
approach in the REDiTM Rating System. The adoption of the REDi guidelines is shown to reduce
losses following the DBE event to between 25% and 60% of the performance-based losses and
result in significantly less downtime (on the order of half) of the performance-based schemes.
The results of a cost-benefit analysis indicate that all design alternatives are preferable to the
baseline case. The payback period for enhanced performance is expected to be between 4.6 years
and 6.6 years for the structural design alternatives and between 5.3 years and 9 years for the REDi
83
designed buildings. A sensitivity analysis reveals that, in the worst case, the payback period for
enhanced building design is on the order of 35 years.
Some of the key limitations of this research include:
- The analysis was confined to a single site location, therefore the sensitivity of the results to
the building location cannot be evaluated
- A number of modeling assumptions were made, including neglecting the effect of soil-
structural interaction
- The structural design alternatives were, for the most part, not optimized.
- Only one kind of structural system, a concrete core wall, was evaluated.
- Only new tall building construction was considered.
Future studies should address the limitations of the research presented herein, particularly in the
consideration of additional structural systems and additional building heights. Further understanding
the performance of non-structural components and pursuit of enhancements is shown to be a key
contributor to reducing losses. Future research into the performance of non-structural components
should focus on the cost of improved performance and the optimal balance between non-structural
performance and increased costs to achieve this.
84
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