Spandrel Type on the Lateral Behavior

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Implications of the spandrel type on the lateral behavior ofunreinforced masonry walls

Fulvio Parisi*,, Nicola Augenti and Andrea Prota

Department of Structures for Engineering and Architecture, University of Naples Federico II, Naples, Italy

SUMMARY

Seismic response of unreinforced masonry (URM) buildings is largely inuenced by nonlinear behavior ofspandrels, which provide coupling between piers under in-plane lateral actions. Seismic codes do not appropri-ately address modeling and strength verication of spandrels, adapting procedures originally proposed for piers.Therefore, research on spandrels has received signicant attention in some earthquake-prone countries, such as

Italy and New Zealand. In the last years, the authors of this paper have performed both monotonic and cyclicin-plane lateral loading tests on full-scale masonry walls with single opening and different spandrel types. Thosetests were carried out in a static fashion and with displacement control. In this paper, experimental outcomes fortwo as-built specimens are presented and compared with those obtained in the past for another as-built specimenwith a wooden lintel above the opening. In both newly tested specimens, the masonry above the opening wassupported by a shallow masonry arch. In one of those specimens, a reinforced concrete (RC) bond beam wasrealized on top of the spandrel, resulting in a composite URM-RC spandrel. Then, the inuence of spandrel typeis analyzed in terms of observed damage, forcedrift curves, and their bilinear idealizations, which allowed tocompare displacement ductility and overstrength of wall specimens. Furthermore, effects of rocking behavior ofpiers are identied, highlighting their relationship with hysteretic damping and residual drifts. Copyright 2014 John Wiley & Sons, Ltd.

Received 28 October 2013; Revised 25 March 2014; Accepted 1 May 2014

KEY WORDS: full-scale testing; in-plane lateral behavior; masonry walls; spandrels

1. INTRODUCTION

Destructive earthquakes have demonstrated that unreinforced masonry (URM) buildings can suffer heavy

damage to spandrels, that is, horizontal strips connecting different piers (Figure 1(a)). This occurs

especially if local out-of-plane mechanisms do not take place, and hence, walls are mainly subjected to

in-plane lateral loads [1]. In this case, spandrels effectively participate to the global seismic response of

the URM building, resulting in the need to incorporate them as primary structural components in the

capacity model of the structure. How to account for this is not appropriately an issue of seismic codes

(e.g., [24]) where simplied procedures for modeling and strength verication of spandrels are

provided by just adapting those formulated for piers (e.g., [57]).Different spandrel types can be identied in URM buildings, depending on the type of lintel above the

opening, namely at the lower edge of the spandrel, and the presence/absence of tensile resistant elements

on top of the spandrel. Dealing with existing URM buildings not designed for earthquake resistance,

spandrels have only stone, masonry, wooden, or steel lintels above openings, which transfer gravity

loads to piers. Especially in the case of old URM buildings with historical value, the masonry above the

*Correspondence to: Fulvio Parisi, Department of Structures for Engineering and Architecture, University of NaplesFederico II, via Claudio 21, 80125 Naples, Italy.E-mail: [email protected]

Copyright 2014 John Wiley & Sons, Ltd.

EARTHQUAKE ENGINEERING & STRUCTURAL DYNAMICSEarthquake Engng Struct. Dyn.2014; 43:18671887Published online 28 May 2014 in Wiley Online Library (wileyonlinelibrary.com). DOI: 10.1002/eqe.2441


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opening is supported by aregular arch(Figure 1(b)) or aat arch(Figure 1(c)), which transfers gravity

loads as both downward vertical forces and outward horizontal forces, the latter referred to as thrusts.

Otherwise, wooden or steel lintels can be found in ancient URM buildings (Figure 1(d) and (e)). A steel

lintel is usually realized with two I-beams that are transversely connected by steel ties to avoid potential

torsional-exural buckling. In this case, lateral external space along the beams is lled with mortar just

to provide at surfaces. In most cases, earthquake damage mainly affects spandrel panels, namely the

parts of spandrels between consecutive openings, whereas spandrelpier intersection zones, also called

joint panels [7], are usually found to be undamaged. Figure 2(a)(c) shows damage to spandrel panels

after the 2009 LAquila and 2012 Emilia-Romagna earthquakes in Italy.

Since the last century, reinforced concrete (RC) lintels have been constructed in most URM

buildings. When the importance of wall-to-wall and oor-to-wall connections to provide a global

box-type seismic response was recognized after major seismic events, tensile resistant elements,

for example, wooden/steel ties and RC bond beams, were inserted on top of spandrels. Particularly,

RC bond beams provide the following types of seismic resistance to URM walls with openings: (i)

out-of-plane bending resistance, because beams transfer most part of horizontal diaphragm actions to

longitudinal walls; and (ii) in-plane bending resistance, because their longitudinal reinforcing steel

bars provide both tensile and compressive strengths to the spandrel, while concrete provides

additional compressive strength. Therefore, URM spandrels with masonry arches or wooden lintelsabove openings are typically observed in ancient URM buildings not designed for earthquake

resistance, whereas composite URM-RC spandrels are usually identied in modern URM buildings

either in as-built conditions or after seismic retrotting.

Experimental programs on URM building models with different spandrel types have been carried

out in the last 15 years. Benedettiet al. [8] performed a pioneering research work based on shaking

table tests of 24 half-scale building models. In this case, some models were made of brick masonry

and wooden lintels, whereas other models were made of stone masonry and regular arches above

openings. Those tests and subsequent energy evaluations [9] revealed a key concept of the seismic

response of URM buildings, that is, their energy dissipation capacity can be maximized if damage

develops within spandrels rather than piers. Bothara et al. [10] performed a shaking table test on a

(a) (c)(b)

Figure 2. Observed damage to spandrel panels: (a) partial collapse of regular arch; (b) exural cracks closeto spandrelpier intersection; and (c) diagonal shear cracks.

(c)(b)

(e)(d)(a)

Figure 1. Spandrel types in ancient unreinforced masonry walls with openings: (a) identication of spandrelsand piers; (b) spandrel with regular arch; (c) spandrel with at arch; (d) spandrel with wooden lintel; and (e)

spandrel with steel lintel.

1868 F. PARISI, N. AUGENTI AND A. PROTA

Copyright 2014 John Wiley & Sons, Ltd. Earthquake Engng Struct. Dyn.2014; 43:18671887

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half-scale masonry building model with both wooden lintels and at masonry arches. That test

conrmed that spandrels can experience not only diagonal shear cracking (Figure 2(c)) but also

vertical exural cracks that originate at the location of opening corners and develop according to

masonry interlocking effects at spandrelpier intersections (Figure 2(b)).

On the basis of observed postearthquake damage evidence and such preliminary experimental results,

research programs have been underway for several years to investigate the mechanics of spandrels and

their inuence on nonlinear seismic behavior of URM buildings. The Italian Network of EarthquakeEngineering University Laboratories, whose Italian acronym is ReLUIS, have funded two research

programs including specic tasks to investigate the role of spandrels in the in-plane behavior of URM

walls. Full-scale testing was believed to be the most effective solution to that end, so different

spandrelpier systems in terms of geometry, masonry assemblage, and lintel type have been tested by

several research groups [1113]. In particular, the authors of this paper have investigated the inuence

of masonry spandrels on the in-plane lateral behavior of tuff masonry walls with single opening,

reproducing some of the typical spandrel types in Mediterranean countries. The rst wall specimen had

a wooden lintel above the opening and was tested in both as-built and predamaged conditions, as well

as after repair and external strengthening with an inorganic matrix grid composite material [13]. This

investigation was motivated by the large presence of tuff masonry constructions (including cultural

heritage goods) in many earthquake-prone regions and consisted of quasistatic lateral loading tests in

displacement control. Furthermore, experimental testing was carried out on walls with openings, in

order to assess the inuence of spandrels including spandrelpier interaction effects, for example, the

modication of boundary conditions to the spandrel as a result of damage to piers. This feature of real

behavior cannot be captured by testing single spandrel elements or spandrel pier assemblages where

the drift on the spandrel is directly imposed as differential vertical displacements or rotations to the

piers, which remain macroscopically undamaged over the entire test duration (e.g., [11, 12]).

In this work, the main experimental ndings of two additional cyclic tests on full-scale tuff masonry

walls with an opening and a regular masonry arch above the opening, rather than a straight lintel, are

presented. One of those specimens also included a RC bond beam on top of the whole spandrel. Such

tests allow to increase knowledge about pros and cons of masonry arches and RC bond beams in

spandrels, giving the chance to improve seismic codes for design and assessment of URM building

structures. Experimental results are compared with those previously obtained for the rst specimen,

in terms of crack patterns, forcedrift curves, displacement ductility, overstrength, and rocking

behavior effects. Finally, hysteretic damping and residual drifts are discussed as they are related toenergy dissipation capacity and seismic loss assessment.

2. EXPERIMENTAL PROGRAM

Quasistatic in-plane lateral loading tests were carried out on two wall specimens with different spandrel

types. In view of a comparative assessment, the rst wall specimen discussed by the authors in [13] is

herein recalled specimen W. The additional specimens forming the core of this paper are labeled as A

and AB in the following. Specimen A had a regular masonry arch above the opening, whereas

specimen AB had both the regular arch and a RC bond beam running on top of the spandrel.

2.1. Description of specimens

Each specimen was a symmetrical tuff masonry wall with a central opening (Figure 3(a)(c)). Tuff

masonry was composed of yellow tuff stones that were 100 150 300 mm3 in size and were bonded

by mortar joints with nominal thickness of 10 mm. Tuff stone masonry layers were alternated along the

height of the specimen in a way to obtain discontinuous head joints and a running bond masonry. The

overall geometry of specimens A and AB was equal to that of specimen W, which was dened on the

basis of the following: (i) mechanical characterization tests on tuff stones, mortar, and masonry as a

whole; and (ii) numerical predictions provided by static pushover analysis of macroelement models

according to [7]. Details on material testing and design of specimens can be found in [1315]. It is only

underlined here that the design of specimens was aimed at developing most of damage within the

IMPLICATIONS OF THE SPANDREL TYPE ON THE LATERAL BEHAVIOR OF URM WALLS 1869


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spandrel, in order to capture its impact on the in-plane behavior of specimens. Because of the nonlinearity

in the lateral behavior of walls with openings, the authors calibrated the magnitude of the initial axial loads

to apply on the piers before lateral loading (Section 2.3). That calibration was carried out by performing

pushover analysis on macroelement models subjected to different axial load magnitudes and hence

deriving a loading condition where the spandrel was expected to fail before the piers.

Each specimen can be regarded as an assemblage of two equal piers connected by a spandrel panel.

Both the piers and the spandrel panel were 1.70 m long, resulting in a total length of the specimen equal

to 5.10 m. Three tuff stone masonry layers were built over the spandrel, to obtain pier continuity in

elevation with an ideal upper story. Indeed, specimens were supposed to be taken out from a typicalmultistory wall with openings. The total height of the specimen was then 3.62 m, while its thickness

was 0.31 m, that is, the sum of two 0.15-m-thick masonry wythes with a collar mortar joint. In the

case of specimen W the height of the opening and spandrel was 2.30 and 1.00 m, respectively.

Therefore, the rise of masonry arches in specimens A and AB was set to 330 mm in order to obtain

the intrados of the keystone at the same height of the wooden lintel in specimen W and hence a

spandrel height at the specimen centerline of 1.00 m for both specimens. The rise-to-length ratio of

the arch was about 1/5, and the depth of the arch cross-section was equal to 300 mm.

In the case of specimen AB, the RC bond beam was cast in place, and its cross-section was

310230mm2. According to specications of the past Italian Masonry Code (IMC) [16], practice

rules, and also some current seismic codes [4, 17], the bond beam was reinforced with four longitudinal

steel bars (14 mm in diameter) and 2-leg steel stirrups (8 mm in diameter) with 200 mm spacing. It is

emphasized that the current Italian Building Code (IBC) [4] establishes that stirrups should be at least6 mm in diameter, and their spacing should not be larger than 250mm. IMC provided the same

detailing rules, while allowing a stirrup spacing up to 300 mm and increasing the minimum stirrup

diameter to 8 mm in the case of URM buildings with more than six oors. No specications about RC

bond beams are provided by Eurocode 8 (EC8) part 1 [17] for new URM buildings.

The tuff stones and mortar used for the construction of specimens were experimentally characterized

through compression, tensile and direct shear tests [14, 15]. Tuff masonry as a whole was also

investigated through both simple and diagonal compression tests to simulate lateral loading tests by

means of macroelement pushover analysis. Masonry joints were made of a hydraulic mortar, which

was a mixture of natural sand and pozzolana-like reactive aggregates with 1 : 6.25 water/sand ratio

by weight. Mean unit weight of tuff stones and mortar was, respectively, equal to 11.72 kN/m3

(a)

(b) (c)

Figure 3. Geometry of (a) specimen W; (b) specimen A; and (c) specimen AB (dimensions in mm).



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(with coefcient of variation CoV= 1.73%) and 16.92 kN/m3 (CoV= 0.65%). Mean values of

uniaxial compressive strength fc, uniaxial tensile strength ft, compressive Youngs modulus Ec,

and shear modulus G are outlined in Table I along with theirCoV in round parentheses.

Direct shear tests provided shear sliding strength at zero conning stress0,bj= 0.15 MPa and initial

friction coefcient = 0.29 for the tuff stone-mortar interface (further properties of that interface can be

found in [15]).

Finally, concrete and reinforcing steel bars employed to build the RC bond beam on top of the spandrelof specimen AB were experimentally characterized by displacement-controlled tests. In particular,

compressive and three-point bending tests were carried out on concrete cubes and prismatic specimens,

respectively. Tensile tests were carried out on steel bar specimens. Mean values and CoV (in round

parentheses) of mechanical properties of concrete and reinforcing steel bars are listed in Table II, where

fccand fctare the cube compressive strength and uniaxial tensile strength of concrete; fy and ftuare the

yielding and ultimate strengths of steel bars, respectively; Ecc is the compressive Youngs modulus of

concrete; andAgtis the maximum elongation of steel bars.

Experimental results in Table II indicate that the concrete and reinforcing steel used for the RC bond

beam are very close to type C20/25 concrete and type B450C reinforcing steel, according to EC2 [18]

and IBC [4], respectively. Actually, such concrete and steel types are also equivalent to the concrete

type Rck250 and steel type Feb44k in the past Italian code on RC and metallic structures [19].

2.2. Test setup and instrumentation

The experimental setup employed for in-plane lateral loading tests is shown in Figure 4. First of all, -

shaped steel plates were bolted to squared holes of the laboratory strong oor at the location of pier

corners, in the transverse direction of the specimen. Then, RC beams (200 310 1900 mm3 in size)

were cast in place over each couple of plates, in the longitudinal direction of the specimen. Shear

keys welded to the -shaped plates ensured a rigid connection of the beams to the strong oor. The

RC beams were cured during 28 days, and the masonry piers were built up over them without

supplemental shear keys that is by realizing just a simple mortar joint as pierbeam connection. This

is in agreement with past practice rules and codes (e.g., [16]). It is also noted that the RC beams at

the base of piers were made of the same concrete and reinforcing steel of the bond beam in

specimen AB. After each specimen was built up, two rigid steel beams were placed over the piers to

uniformly distribute vertical forces simulating gravity loads. Such forces were applied by

bidirectional hydraulic jacks with 500 kN nominal capacity, which were placed over the steel beams

at the centerlines of the piers. Couples of polytetrauoroethylene (PTFE) layers were installed

between hydraulic jacks and rigid beams to minimize friction at their interface.

Table I. Mechanical properties of constituent materials of tuff masonry.

Material fc[MPa] ft[MPa] Ec [GPa] G[GPa]

Tuff stones 4.13 (18.54%) 0.23 (22.06%) 1.54 (6.43%) 0.44 (25.78%)Pozzolana-like mortar 2.50 (7.34%) 1.43 (6.23%) 1.52 (22.12%) 0.66 (10.66%)

Table II. Mechanical properties of constituent materials of reinforced concrete.

Material fcc[MPa] fct[MPa] Ecc[GPa] fy[MPa] ftu [MPa] ftu/fy Agt[%]

Concrete 35.29(14.31%)

2.82(9.48%)

27.86(10.18%)

D8 reinforcing steel bars 545.46(5.36%)

632.63(2.43%)

1.16(3.18%)

11.73(9.63%)

D14 reinforcing steel bars 510.22(5.78%)

605.57(3.74%)

1.19(2.11%)

15.59(6.35%)



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Transverse steel frames were installed over the piers at the location of their centerlines, in order to

provide contrast against vertical loading. Cap beams of transverse frames were properly stiffened to

avoid local buckling, and bolted connections were realized for both beamcolumn and columnbase

connections. The columns were connected by a couple of steel beams at each specimen side to

prevent potential out-of-plane failure modes.

Finally, a second reaction system consisting of a nonprismatic steel wall was used to apply lateral

loading to the specimen. The reaction wall was anchored to the laboratory strong oor by means of

four steel bars, each pretensioned at 400 kN. Cyclic lateral loading was applied by a servocontrolled

hydraulic actuator with 500 kN nominal compressive capacity, 290 kN nominal tensile capacity, and

250 mm stroke. That actuator was bolted on top of the reaction wall and supported by an

additional cable-stayed system. The application of load reversals to the specimen was allowed by six

steel bars (18 mm in diameter), which were bolted to the rigid end plate of the actuator and toanother rigid plate at the opposite edge of the specimen. A spherical hinge was installed between the

end of the horizontal actuator and its rigid plate, in order to apply in-plane lateral loading without

any parasite out-of-plane action for the specimen. A load cell with nominal and maximum capacities

equal to 200 and 250 kN, respectively, was used to measure the actual lateral load, which was

applied to the specimen.

Two different types of displacement transducers were mounted over specimens: linear variable

differential transformers (LVDTs) and wire potentiometer transducers (PTs). Joint panels were not

instrumented as they were expected to experience small deformations and no macroscopic cracks

owing to the geometry of the specimens (assumption conrmed by tests). The typical arrangement

of displacement transducers is shown in Figure 5. LVDTs were installed at end sections of piers and

spandrel panel above the opening, to obtain data on exural deformations, whereas PTs were

employed to measure both shear and rocking-induced deformations.In particular, four vertical LVDTs were installed at the base of the piers and put in contact against the

RC beams (see the front side in Figure 5), in order to measure the width of potential tensile horizontal

cracks induced by the rocking behavior of the piers. The four vertical PTs along the height of the piers

(back side) were anchored to the RC beams as well, to include wire elongations related to rocking-

induced cracks. In the case of specimen AB, the RC bond beam was instrumented as follows: eight

horizontal LVDTs over cross-sections located in the proximity of spandrel pier intersections, an axial

horizontal LVDT at the mid section of the RC bond beam. The rst group of LVDTs was installed to

measure deformations at the location of potential plastic hinges, which were expected to be caused by

large drift demands on the spandrel panel as a result of piers rocking. The horizontal LVDT was

mounted to measure axial deformations of the RC bond beam.

Figure 4. Experimental setup.



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An additional potentiometer tagged as PT #1 was installed at the height of the horizontal actuator

but on the opposite side with respect to it. That potentiometer allowed to measure actual lateral

displacements experienced by the specimens. Real-time readings at the load cell and PT #1 were

assumed to be more reliable than load and stroke measurements provided by the actuator itself, to

plot forcedrift curves for each specimen.

2.3. Loading protocols

Lateral loading tests on specimens W, A, and AB were based on vertical loading of piers and then lateral

loading applied at the height of the spandrel. Vertical loading consisted of two concentrated forces applied

at the top of the piers by the hydraulic jacks. After two initial load cycles were completed to obtain

effective contrast between jacks and piers, the magnitude of vertical forces was monotonically

increased up to 200 kN (corresponding to approximately 10% of the mean ultimate load of pier cross-

sections). Then, vertical forces were kept constant, and the horizontal actuator began to apply lateral

loading on the specimen in a quasistatic fashion and with displacement control, to capture the softened

postpeak forcedrift response under increasing deformation demand. Displacement-controlled loading

was managed by the computer program of the data acquisition system, where a target displacement

time history was implemented. Therefore, the lateral force was changed at each time step in a way tomeasure the target displacement on the horizontal actuator. All measurements were recorded at a

sampling rate equal to 5 Hz. The lateral loading stage of each test started with two displacement cycles

between 0.15 and 0.25 mm to reach good contrast between specimen and actuator. After those cycles

were completed, lateral loading was imposed to the specimen in accordance to the target displacement

time history inputted in the computer program. The time history dened for specimen W was different

from those of specimens A and AB. Indeed, specimen W was tested under monotonically increasing

displacements until a maximum displacement dmax= 28mm was read at the actuator, which

corresponded to the onset of damage to the spandrel. That stop was motivated by the need to assess the

specimen response also in predamaged and repaired-upgraded conditions, as discussed in [13]. On the

other hand, both specimens A and AB were subjected to cyclic loading up to a near-collapse

performance level. Lateral loading was imposed in accordance to the target displacement time history

shown in Figure 6, which was composed of 17 cyclic displacement blocks, each of them consisting ofthree cycles at each amplitude peak, that is, a total number of 51 cycles leading to dmax= 95.1 mm. The

displacement rate was set to 0.70 mm/s, whereas the displacement increment between consecutive

groups of three displacement cycles was equal to 5.6 mm.

If the interstory drift ratio w is introduced as lateral displacementddivided by the height of lateral

loading line from the base of the piers (y0 = 3050 mm), that is,w = d/y0, it can be equivalently stated

that monotonic loading on specimen W was stopped at a maximum driftmax= 0.9%, whereas cyclic

loading on specimens A and AB was dened in a way to reach max= 3.1%. Nevertheless, cyclic tests

were expected to be stopped at the attainment of near-collapse performance level for the specimen. The

target displacement time history was then followed up to actual maximum drifts that were different for

specimens A and AB, depending on the lateral strength drop measured by the horizontal actuator and

Figure 5. Typical instrumentation of specimens (dimensions in mm).



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the damage level observed during the test. The near-collapse performance level was associated with a

lateral strength drop of at least 20% on the postpeak branch of the forcedrift envelope curve related

either to positive or negative loading orientation.

3. OBSERVED DAMAGE

Crack patterns observed during any test were generated by a signicant rocking response of the

specimen. No cracking was detected on spandrelpier intersection zones, conrming the

assumption of small deformations and no macroscopic cracks used for design and instrumentation

of specimens (Section 2.2). With reference to Figure 7, the observed damage mainly consisted

of the following: (i) both diagonal shear and exural cracking in the spandrel panel above the

opening; (ii) horizontal tensile cracks at the base of the piers due to their rocking response;

and (iii) masonry crushing with transverse splitting at pier corners as near-collapse conditions

were attained. Those failure modes depended on the specimen geometry and magnitude of

axial loads imposed to the piers. In fact, the moderate slenderness of both piers and spandrel,

as well as low axial load levels on the piers, were expected to be associated with exuralcracking even at low lateral force levels, according to analytical predictions by limit strength

domains of URM cross-sections [6, 7]. Nevertheless, most part of damage concentrated in the

spandrel panel.

Damage detected on specimen W is herein briey discussed for comparative purposes. Horizontal

exural cracks were observed at the base of the piers at small drift levels (w0.06%). As w was

increased, the spandrel panel suffered vertical exural cracks at its end sections and a vertical crack

close to its midspan. Finally, diagonal shear cracking of the spandrel panel occurred atw = 0.65%.

For specimen A, diagonal shear cracking started at drift levels smaller than those recorded for

specimen AB. Diagonal shear cracking originated at both lower ends of the spandrel panel, and

then, it propagated along diagonals involving both tuff stones and mortar joints. In this case, the rst

portion of arch fell down atw = 0.37%, which is a drift level close to half the drift measured in the

case of specimen AB at the cracking onset in the masonry arch. Diagonal cracks propagated until asecond part of arch collapsed atw = 0.56%. That drift level also induced horizontal exural cracks

at the base of piers. Then, the spandrel panel totally failed in shear atmax= 1.12%, which produced

the sudden loss of connection between the piers and hence a signicant lateral resistance drop for

the entire specimen (it is noted thatmax is intended to be the maximum drift corresponding to the

lateral displacement measured by potentiometer PT #1). After that failure, the upper triangular

fraction of masonry above diagonal shear cracks did not provide any contribution to the lateral

resistance, and the test was stopped.

As the spandrel of specimen AB included a masonry arch and a RC bond beam, the damage

observed at near-collapse (corresponding tomax= 3.1%) was the result of a failure sequence, which

can be summarized as follows:

-100

-50

0

50

100

0 5000 10000 15000

d[mm]

t [s]

Figure 6. Target displacement time history for cyclic tests.



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formation of axial cracks in the arch atw = 0.69% (Figure 8(a));

collapse of little central piece of arch atw = 0.86% (Figure 8(b));

collapse of arch wedge atw = 1.03%, after rocking beginning of piers (Figure 8(c));

collapse of third arch portion and early masonry crushing at the base of the piers atw = 1.21%

(Figure 8(d));

collapse of fourth arch portion and early concrete cover spalling in the RC bond beam (top of end

section) atw = 1.72% (Figure 8(e));

complete concrete cover spalling atw = 2.58%; and collapse ofrst and second masonry layers above the arch (at the rst and second displacement

cycles, respectively) and plastic hinging of the RC bond beam atw = 2.67% (Figure 8(f)).

Therefore, the spandrels of specimens A and AB were notably prone to damage, as a result of a

signicant fragility of the masonry arch under differential displacements at its supports. Such a

difference in displacements was caused by different rocking rotations of the piers and the

associated drift demand on the spandrel panel. In fact, the piers of a wall with openings can

experience a different response even if they are subjected to equal boundary conditions and axial

load levels. This is generated by axial load variations resulting from a sort of frame behavior

under a global overturning moment, as well as nonlinear behavior due to smeared cracking of

(a)

(b)

(c)

BACKFRONT

BACKFRONT

BACKFRONT

Figure 7. Crack patterns to (a) specimen W; (b) specimen A; and (c) specimen AB.



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masonry. The combination of such effects gives a dependence of pier exibilities on both axial

and lateral loads.

The rocking behavior of the piers was the source of large drift demands on the RC bond beam,

which suffered plastic hinging even though shear cracks propagated from the masonry below. That

propagation highlights that the 200 mm stirrup spacing was too large to favor full plastic hinging

without shear cracking in the RC beam. This also means that, when designing a wall with openings

where piers are expected to experience signicant rocking, a complete exploitation of plastic

resources of the RC bond beam is not guaranteed if specications by the past IMC [16] and current

IBC [4] are met and also when seismic design is carried out according to EC8 part 1 [17]. This is a

typical situation for modern URM buildings located in Italy and other European countries.

Figure 9(a)(f) shows both overall damage to specimens and crack patterns suffered by their

spandrel panels at the maximum drift imposed during the tests. As the maximum drift level wasreached during the test, specimen W experienced small horizontal tensile cracks at the base of the

piers and moderate cracking of the spandrel. In fact, the latter suffered both exural and diagonal

cracking. Two types of vertical cracks were observed: cracks in the proximity of the spandrel pier

interfaces, which were induced by the rocking behavior of the spandrel panel under lateral loading,

and a vertical crack close to the mid section, which was chiey caused by exural deformation of

the wooden lintel under masonry self-weight. Diagonal cracking formed only along a single

diagonal of the spandrel panel, as a result of monotonic loading from the left to the right (Figure 9(b)).

That loading orientation is denoted as positive in the case of cyclic tests on specimens A and AB.

The lintel did not lose its supports even though the anchorage length was just 150 mm, so the

masonry above was effectively supported during the test.

The test on specimen A revealed a more brittle behavior resulting from diagonal shear cracking of

the spandrel panel at small drift levels. Figure 9(d) reveals that diagonal shear cracks involved thewhole height of the spandrel panel and the complete loss of connection between the piers.

The cyclic in-plane loading test on specimen AB conrmed a signicant rocking response of all

the walls, which caused heavy damage to both piers and spandrel. The onset and propagation of

diagonal cracking from the masonry arch to the RC bond beam forced the masonry to fall down

progressively. Only two masonry layers did not collapse and did not lose their bond to the upper

RC bond beam (Figure 9(f)). The lower edge of fractured masonry arch was found to be almost

aligned with the diagonals of the spandrel panel. The latter did not behave as a simple rod able to

transfer just axial loads, because it dissipated drift demands through plastic hinging at its ends. It

is emphasized that the RC beam lost its upper concrete cover and suffered shear cracking as a

result of partially effective transverse reinforcement. Although diagonal cracks veried for the

(a) (c)

(e)

(b)

(d) (f)

Figure 8. Damage to the spandrel of specimen AB under increasing drift levels: (a) w = 0.69%; (b)w =0.86%; (c)w = 1.03%; (d)w = 1.21%; (e)w = 1.72%; and (f)w = 2.67%.



DOI: 10.1002/eqe


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spandrel panel of specimen AB almost completely reproduced those observed on specimen A, the

RC bond beam had an important benecial effect as it delayed the propagation of shear cracks to

larger drift levels.

Specimens A and AB reached the near-collapse performance level as a result of different phenomena,namely, complete loss of connection between the piers when diagonal shear cracking developed along the

whole height of the spandrel panel, in the case of specimen A, and heavy crushing of masonry at the base

of the piers when the drift level attained max= 3.1%, in the case of specimen AB. For this last specimen,

Figure 10(a)(d) demonstrates that the rocking response of the piers starts with a single tensile crack along

a bed joint and then results in a combination of transverse splitting and moderate-to-heavy crushing of

masonry at the pier toes. In the case of specimen AB, the width of the horizontal crack exceeded

30 mm in the left pier (front side), while it reached approximately 15 mm in the right pier. Transverse

splitting of masonry was particularly evident in the left pier, where the vertical crack width was larger

than 50 mm (Figure 10(c)). Conversely, in the case of specimens W and A, horizontal cracks at the

base of the piers reached a width smaller than 10 mm.

Figure 9. Damage at maximum drift levels: (a) specimen W and (b) its spandrel (max= 0.9%); (c) specimenA and (d) its spandrel (max= 1.1%); and (e) specimen AB and (f) its spandrel (max= 3.1%).



DOI: 10.1002/eqe


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4. EXPERIMENTAL FORCEDRIFT CURVES

Force and displacement readings recorded by the data acquisition system were processed in order to

obtain forcedrift curves, where the lateral resisting force and specimen drift were denoted by Hand

w, respectively. A comparative analysis of such curves is performed herein, considering drift

measurements corresponding to displacement readings of PT #1, that is, the wire potentiometer

placed at the height of the horizontal actuator.

A roughly bilinear (exure-dominated) forcedrift behavior up to the peak resistance was found in

the case of specimen W under monotonic lateral loading (Figure 11(a)), allowing to identify an ideal

cracking point, and hence both initial and postcracking lateral stiffness. When the peak lateral

resisting force was attained, diagonal shear cracking occurred in the spandrel panel, and a 15% drop

in the lateral resisting force was measured. As the lateral drift was further increased, the lateral

resisting force raised again. The test was stopped at the attainment of max= 0.9% to allow further

cyclic testing in predamaged conditions, which is out of the scope of this paper ([13]).

Typical narrow forcedrift loops forexural response with both stiffness and strength degradation,

as well as low energy dissipation capacity, were found for specimens A and AB. Figure 11(b) shows

that specimen A experienced very different drift levels depending on the orientation of lateral loading,

even though the peak resisting levels were almost the same. A higher displacement capacity was found

in the case of specimen AB, which also experienced small residual drifts compared with maximumdrifts, namely, a signicant recentering capacity (Figure 11(c)). It is noted that recentering capacity

is the ability of the structure to return the center of mass to its initial position after unloading. After

the peak resisting force, specimen AB suffered a signicant, but gradual, strength degradation as the

lateral drift increased up to the maximum drift imposed during the test.

On the basis of the forcedrift responses (their envelopes for cyclic tests), the following performance

levels were identied for each specimen: (i) cracking onset; (ii) peak resistance; and (iii) maximum

displacement. Forces and drifts associated with such performance levels are listed in Table III. The

cracking point of the experimental forcedrift curves was dened at a lateral stiffness reduction

equal to 10%. That stiffness drop was captured by monitoring the ratio between secant lateral

stiffness k and simple moving average stiffness ksma at each resisting force level. The secant

stiffness was dened as the ratio of the measured resisting force H to the corresponding

displacement d. The simple moving average stiffness at a given force level was de

ned as thearithmetic mean of the secant stiffness values ranging between the rst nonzero value and that

corresponding to the force level under consideration. In that way, signicant variations in stiffness

were identied, and secant lateral stiffness at cracking was dened as the ratio between the

estimated cracking forceHcrand the displacementdcr(corresponding to the cracking driftcr). The

maximum resisting force and its corresponding lateral drift are, respectively, denoted by Hmax and

Hmax, whereas the maximum drift imposed during the test and the corresponding resisting force are

indicated as maxand Hmax, respectively.

In the case of specimen W, the cracking resisting force was approximately 0.5 times the peak

resisting force, whereas Hcr=0.7Hmax is assumed by IBC [4]. The cracking drift was instead about

one order of magnitude lower than the drift associated with peak resistance.

(a) (b) (d)(c)

Figure 10. Rocking response of the piers of specimen AB: (a) tensile crack close to the base of the left pier(front side) and (b) crack width measured and (c) masonry splitting and crushing of left pier and (d) masonry

splitting and crushing of right pier.



DOI: 10.1002/eqe


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Specimen A suffered a signicant stiffness reduction during lateral loading, even in the prepeak

range of forcedrift curves. Indeed, on the basis of stiffness monitoring, cracking onset was

conventionally identied at resisting force and drift levels, that is, Hcr and cr, which were almost

equal to those measured in the case of specimen W but signicantly lower than those recorded on

specimen AB. The ratio Hcr/Hmax was approximately equal to 0.6 and 0.5 in the positive and

negative loading orientations, respectively. Specimen A was also characterized by a signicantly

lower peak resistance compared with other specimens, as a result of the regular masonry arch,instead of the wooden lintel, above the opening. This caused that Hmax was 17% and 22% lower

than that of specimen W in the positive and negative loading orientations, respectively.

In the case of specimen AB, the RC bond beam allowed to reach a peak resistance 46% and 57%

higher than those computed for specimen A in the positive and negative orientations, respectively.

Dealing with the rising branch of the force drift envelope curve, the ratio Hcr/Hmaxwas about 0.6

and 0.8 in the positive and negative loading orientations, respectively. The RC bond beam delayed

the 10% stiffness drop at force and drift levels notably higher than those related to specimen W. The

secant lateral stiffness at cracking for specimen AB was approximately 10% lower than that

estimated for specimen W. This can be motivated by the presence of the masonry arch above the

opening, which was less effective than the wooden lintel in connecting the piers, thus resulting in a

-250-200

-150

-100

-50

0

50

100

150

200

250

H[kN]

Specimen W

Specimen A

Specimen AB-250-200

-150

-100

-50

0

50

100

150

200

250

H[kN]

0

50

100

150

200

250

-4 -3 -2 -1 0 1 2 3 4

0 1 2 3 4

H[kN]

(a)

(d)

-250

-200

-150

-100

-50

0

50

100

150

200

250

-4 -3 -2 -1 0 1 2 3 4

-4 -3 -2 -1 0 1 2 3 4

H[kN]

Cyclic response

Envelope

(b)

Cyclic response

Envelope

(c)

Figure 11. Forcedrift curves: (a) specimen W; (b) specimen A; (c) specimen AB; and (d) envelopes.

Table III. Parameters of the experimental response of specimens.

SpecimenLoading

orientation Hcr[kN] cr[%] Hmax[kN] Hmax[%] Hmax[kN] max[%]

W Positive 99 0.06 184 0.65 172 0.89

A Positive 96 0.06 153 0.38 141 0.59Negative 73 0.07 143 0.62 84 1.10

AB Positive 132 0.09 224 0.52 126 2.67Negative 185 0.16 224 0.88 109 3.13



DOI: 10.1002/eqe


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lower lateral stiffness of the whole specimen. Specimen AB reached the same peak resistance in both

loading orientations, even though their corresponding drifts were different. In fact, in the case of

specimens A and AB, the difference was that theirrst cracked loading was conditioned on cracking

previously suffered in the opposite orientation. This interpretation of asymmetrical response at

different performance levels is also supported by nonlinearity effects because of masonry cracking

and axial load variations in piers generated by global overturning moment corresponding to lateral

loading. Therefore, asymmetrical cyclic behavior can occur even in the case of symmetricalspecimens in terms of geometry, loads, and external boundary conditions. The peak resistance of

specimen AB was approximately 22% higher than that measured in the case of specimen W. The

gradual deterioration of the spandrel panel in the postpeak force drift range caused a progressive

reduction in the connection between the piers, resulting in a signicant strength degradation

(Figure 11(c)). As the maximum drift level was imposed to specimen AB, the lateral resisting force

was found to be less than half the peak resistance.

Finally, the envelopes of hysteretic forcedrift curves are compared in Figure 11(d), where the curve

of specimen W is plotted in the positive quadrant only, because of monotonic loading. It can be

observed that specimen A had the lowest levels of displacement capacity and peak resistance. The

realization of the RC bond beam on top of the spandrel (specimen AB) considerably increased both

displacement capacity and peak resistance. The total vertical load Vapplied to specimens W, A, and

AB was estimated to be 429, 468, and 470 kN, respectively, considering self-weight of constituent

materials and initial axial loads of 200 kN on the piers. Such a computation allowed to compute the

ratio H/V(typically referred to as seismic coefcient), which was found to be approximately 43% in

the case of specimen W, 3133% in the case of specimen A, and 48% in the case of specimen AB,

at the peak resistance level.

5. BILINEAR IDEALIZATION OF FORCEDRIFT CURVES

The envelopes of the experimental forcedrift curves were approximated as bilinear diagrams to

characterize the response of an idealized SDOF system, which is widely used to estimate seismic

demand on inelastic structures (e.g., [20]). Furthermore, idealized bilinear force deformation diagrams

are usually adopted to dene the in-plane seismic capacity of masonry walls under lateral loading in a

simplied way [6, 21]. In the case of walls with openings, the SDOF system approximation does notinduce major errors for single-story specimens (case under study). The methodology used in the past by

the authors for specimen W was also employed for specimens A and AB, in order to consistently

compare their idealized SDOF models and to discuss current code values. Therefore, two bilinear

idealization procedures according to Tomaevi[22], and both EC8 part 3 [3] and IBC [4], were used to

estimate the following capacity features: (i) ultimate force Hu; (ii) yielding drifte(corresponding to de);

(iii) elastic stiffness ke; (iv) displacement pseudoductility ; and (v) overstrength factor . I t i s

underlined that, in this paper, the authors use drifts rather than displacements in order to provide

dimensionless deformation capacity estimates that do not depend on the height of specimens. Regardless

of the bilinear idealization procedure being used, displacement ductility was computed as = u/e,

where u is the ultimate drift. The latter was assumed to be the experimental drift corresponding to a

resisting force level equal to Hu = CSd Hmax, where CSd is the strength degradation factor. This factor

was assumed to be not lower than 0.8, that is, corresponding to a strength degradation not greater than20% on the postpeak falling branch of the forcedrift envelope. This assumption derived from the fact

that the tests were stopped at strength degradation levels usually lower than that typically considered in

seismic codes, that is, 20%. Typically, the ultimate capacity point (u,Hu) of the forcedrift curve was

different from the experimental point at maximum deformation (Hmax,Hmax). By the way, it is also

emphasized that the bilinear idealization for specimen W is presented only for comparative purposes

without providing actual capacity estimates, because the lateral loading test on that specimen was

stopped when just a moderate damage was detected on the spandrel.

As far as lateral stiffness is concerned, the bilinear idealization procedure according to [22] sets keequal to the experimental secant stiffness at cracking (kcr). The experimental force and drift associated

with kcr are denoted as Hcr and cr, respectively. Conversely, in the procedures reported in [3, 4],



DOI: 10.1002/eqe


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cracking is supposed to be associated with a force equal to 0.7Hmaxso that elastic lateral stiffness is

dened as ke = 0.7Hmax/dcr, where dcr is the lateral displacement corresponding to 0.7Hmax on the

forcedisplacement envelope curve. In both bilinear idealization procedures, e is derived by

imposing equal areas below the envelope forcedrift curve and idealized diagram, which means that

the same energy dissipation capacity is assumed for the actual and idealized structural systems. The

ultimate force of the idealized SDOF system was then derived as Hu = ke de. A graphical denition

of idealized force

drift diagrams according to both procedures is provided in Figure 12(a), wheredifferent estimates of cracking, elastic, and ultimate limits are pointed out.

According to IBC [4], the overstrength factor was dened as =Hu/Hcr, which is a part of the

strength reduction factor to be used in linear equivalent seismic analysis (e.g., [23]). It is noted that

IBC [4] allows one to set= 1.4 in the case of single-story URM buildings and = 1.3 in the case

of single-story RM buildings.

As described in Section 2.2, two types of displacement readings were obtained during the tests,

namely, stroke measurements provided by the horizontal actuator and displacements measured by

potentiometer PT #1. When displacement readings of potentiometer PT #1 were considered,

the features of idealized SDOF systems corresponding to specimens were found to be those outlined

in Tables IV and V, which are, respectively, associated with the bilinear idealization procedures

according to [22] and [3, 4]. If the estimates related to positive and negative orientations of lateral

loading are averaged (bracketed values), differences between results of the two bilinear idealization

procedures tend to vanish. This gives the chance to identify the main effects of the spandrel type on

the in-plane seismic capacity. Nevertheless, the idealization procedure provided by EC8 part 3 [3]

and IBC [4] was more sensitive to the postcracking branch of force drift envelopes, overestimating

the yielding drift and underestimating displacement ductility in the case of specimens W and A. The

last was not the case of specimen AB, whose capacity estimates provided by both idealization

procedures were almost the same. This occurred because postcracking stiffness of specimen AB was

(a) (b)

00

0.4

0.2

0.6

0.5 1 1.5 2 2.5 3

H/V

Figure 12. (a) Bilinear idealizations and (b) idealized in-plane capacity diagrams of specimens.

Table IV. Capacity features adopting procedure by Tomaevi.

SpecimenLoading

orientation Hu [kN] e[%] u[%]

W Positive 162 0.10 0.89 8.51 1.64A Positive 147 0.09 0.59 6.44 1.53

Negative 136 0.13 1.04 7.74 1.85(142) (0.11) (0.82) (7.09) (1.69)

AB Positive 211 0.14 1.63 11.85 1.60Negative 212 0.18 1.47 8.00 1.14

(212) (0.16) (1.55) (9.93) (1.37)



DOI: 10.1002/eqe


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not signicantly lower than the initial stiffness, resulting in the same estimate ofkeafter both bilinear

idealizations. On the basis of such considerations, the estimates of specimen W and bracketed estimates

of specimens A and AB listed in Table IV can be preferentially used to compare seismic capacity

features of specimens, because their experimental response seems to be more accurately described

by the idealized SDOF system derived in accordance to Tomaevi [22]. Idealized forcedrift

diagrams averaged over both loading orientations are plotted in Figure 12(b), where resisting force

His replaced by the corresponding seismic coefcientH/V.

First of all, the yielding drift was found to be between 0.10% and 0.11% in the case of specimens W

and A, respectively, and 0.16% in the case of specimen AB, that is, 1.6 and 1.4 times those computed

for specimens W and A, respectively.

This indicates that the RC bond beam on top of the spandrel with masonry arch delayed the

attainment of the elastic limit to larger drift levels. It is also worth noting that the spandrel with

wooden lintel (specimen W) allowed to reach an average ultimate drift larger than that computed for

the spandrel with masonry arch above the opening (specimen A), even if strength degradation of

specimen W at u = max was lower than half that of specimen A. In other words, if specimen W

would have been tested up to a maximum drift corresponding to a strength degradation factor equal

to that of specimen A, the spandrel with wooden lintel would have favored an average value ofusignicantly larger than that actually measured in the case of spandrel with masonry arch. This

remark is consistent with the fact that the wooden lintel was more effective than the arch in

allowing the spandrel to behave as coupling beam for the piers. The realization of the RC bondbeam in the spandrel with masonry arch (specimen AB) almost doubled the average displacement

capacity, increasing u from 0.82% to 1.55% (bracketed values in Table IV). In any case, ultimate

drift limits of piers provided by current codes were exceeded by those estimated for each

specimen. In fact, EC8 part 3 [3] allows one to set u to 0.4% and 0.8% H0/D for shear and

exural failure modes of piers, respectively, where H0 is the distance between the section where

exural capacity of the pier is attained and the contraexure point and D is the length of the pier.

IBC [4] sets u to 0.4% and 0.8% for shear and exural failure modes, respectively, reducing the

exural limit value to 0.6% in the case of existing URM buildings. Such drift limits are related to

the same performance level, which is referred to as signicant damage in EC8 and life safety

prevention in IBC.

The lowest displacement ductility was found for specimen A, whereas the highest estimate was

found for specimen AB, showing a 40% increase in ductility induced by the addition of the RCbond beam to the spandrel with masonry arch. It is also noteworthy that the average displacement

ductility in the case of spandrel with wooden lintel (specimen W) was 20% higher than that related

to the case of spandrel with masonry arch (specimen A).

Approximately, the same estimates of overstrength factor were found in the case of specimens W

and A, which were, respectively, 20% and 23% higher than that related to specimen AB.

Seismic coefcients corresponding to Huwere estimated to be 38%, 30%, and 45% in the case of

specimens W, A, and AB, respectively. This means the following: (i) that the spandrel with wooden

lintel contributed to resist a horizontal acceleration about 27% greater than that related to the

spandrel with masonry arch; and (ii) that the RC bond beam doubled the seismic coefcient of

specimen A. An increase in the seismic coefcient of walls can have great impact on the overall

Table V. Capacity features adopting procedures in Eurocode 8 and Italian Building Code.

SpecimenLoading

orientation Hu[kN] e [%] u [%]

W Positive 171 0.19 0.89 4.68 1.73A Positive 150 0.11 0.59 5.23 1.56

Negative 135 0.12 1.04 8.51 1.84

(143) (0.12) (0.82) (6.87) (1.70)AB Positive 212 0.15 1.63 10.76 1.61

Negative 210 0.16 1.47 8.91 1.13(211) (0.16) (1.55) (9.84) (1.37)



DOI: 10.1002/eqe


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capacity of a masonry building, because higher horizontal seismic accelerations can be resisted without

additional structural mass. This can be an advantage not only in seismic design of URM buildings but

also in retrot of existing buildings. In the case of substandard buildings without historical value,

replacing masonry arches with wooden lintels above openings can be a low-cost seismic upgrading

strategy able to increase the in-plane resistance of walls. If masonry arches are to be preserved for

their architectural value, lintels can be installed above them. Conversely, although the installation of

RC bond beams on top of existing spandrels is a more complex and invasive retrot measure, it alsoconnects orthogonal walls favoring a box-type seismic response of the building.

6. DIFFERENTIAL PIER DRIFTS, HYSTERETIC DAMPING, AND RESIDUAL DRIFTS

Displacement readings provided by the horizontal actuator and PT #1 at opposite edges of each

specimen allowed to estimate lateral drifts of both left and right piers (on the front side), that is, leftand right(note that the latter was previously assumed to be representative of the specimen driftw).

The difference between such drifts, which is denoted here by = left right, can be used as ameasure of the effectiveness of the spandrel in connecting the piers. In Figure 13, the lateral

resisting force is plotted against , on the basis of the envelopes of experimental forcedrift

curves in the positive loading orientation. When was not zero, the spandrel shortened as a

result of horizontal compression (up to the corner points on the rising branches of forcedrift

envelopes) or its progressive failure. When spandrel shortening was caused by horizontal

compression, the spandrel was more able to transfer a fraction of horizontal force from the left

pier to the right pier. Otherwise, the progressive failure of the spandrel induced a gradual loss of

connection between the piers. To associate the relative distance between the piers with given drift

levels of specimens, the ratio /w is used herein. If the forcedrift diagram of the wall with

openings is known, the ratio /w allows to dene the spandrel deformation and damage as the

lateral drift changes.

In the case of specimen W, which was subjected to monotonic testing, the relationship between H

and was rather linear up to H= 120 kN (point CW in Figure 13), namely, the horizontal force

level associated with the corner point on the rising branch of the experimental forcedrift curve

(Figure 11(a)). Such a corner point was recorded when exural cracks at the end sections of thespandrel panel were observed. In that condition, was approximately equal to 0.12%, because it

was associated withleftandrightof about 0.22% and 0.10%, respectively. Therefore, the ratio /wwas

about 1.2 at the rst corner point of the experimental forcedrift curve of specimen W. As H

increased up to its maximum level and then decreased until the maximum drift was reached,

gradually reduced. This indicates a progressive reduction in shortening of the spandrel panel and

hence an elongation up to its initial length. By contrast, if the vertical exural crack on the right-

CW

CA

CAB

0

50

100

150

200

250

0 0.1 0.2 0.3 0.4 0.5 0.6

H[kN]

[%]

Specimen W

Specimen A

Specimen AB

Figure 13. Lateral resisting force versus variation between pier drifts (positive orientation).



DOI: 10.1002/eqe


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19/21

energy dissipation capacity of spandrels with wooden lintels rather than masonry arches above

openings. In all cases, the low damping levels were basically a consequence of small areas per cycle

compared with elastic areas under varying drift. This is a feature of structural systems experiencing

rocking behavior under lateral loads. It is noteworthy that past shear compression tests on individual

brick masonry piers led to estimate an average value of eq,h equal to 5% in the case of exural

failure [6].

Finally, residual drifts of specimens were analyzed to quantify repairability conditions for each spandrel

type. The residual drift is a nonzero drift corresponding to a complete unloading of the structure, that is, a

zero force level (zero crossing of hysteretic forcedrift curve). Residual drifts were then computed

for the specimens under study in each half cycle. This is an important task because the spandrel

type may affect the ability of an URM wall with openings to return to its initial position after

severe lateral loading. It is emphasized that, although collapse does not occur, large residual drifts

make the structure no longer usable or may require too expensive repair measures. This motivates

the following: (i) the use of residual deformations as additional engineering demand parameters

together with maximum deformations, to quantify earthquake damage at different seismic

intensity levels; and (ii) seismic loss assessment procedures considering the impact of residual

interstory drifts.

Figure 14(b) shows residual driftsrof specimens under increasing driftwat each loading cycle.Specimen W, which again is considered in its predamaged conditions, experienced zero residual

drifts until w = 0.15% was reached. Asw increased, a logarithmic increase in roccurred until the

residual-to-maximum drift ratio RMDR = r/wreached 11% atw = 1.26%.

The presence of the masonry arch in specimen A did not signicantly modify the onset of residual

drifts, which became different from zero atw = 0.18%. As the lateral drift increased, the residual drift

increased according to a logarithmic trend but more rapidly than in the case of specimen W. The

residual drift reached 0.26% atw = 1.12%, resulting in RMDR = 23%.

Almost the same trend in residual drifts is found in the case of specimen AB, but it suffered residual

drifts signicantly smaller than those of specimen A. Again, the residual drift was zero until the

specimen reached a lateral drift equal to 0.18%. The increase in residual drift under increasing

w was less important compared with other specimens. Indeed, r was found to be 0.16% at

w = 3.13%, resulting in RMDR = 5%. This indicates that the RC bond beam was able to limitresidual deformations, even if no signicant improvements in terms of hysteretic damping were

found. This nding satisfactorily agrees with previous remarks, because large drifts associated with the

rocking behavior of piers were observed during the test on specimen AB, and most part of permanent

damage (and hence residual drifts) was caused by masonry crushing at pier toes and plastic hinging of

the RC bond beam. Finally, if the curves in Figure 14(b) are compared, it can be noted that, in general,

the lateral driftw produced the following: (i) comparable residual drifts in the case of spandrel with

masonry arch and spandrel with wooden lintel; and (ii) almost a halved residual drift in the case of

spandrel with both masonry arch and RC bond beam. For instance, r= 0.1% was associated with

specimen drifts approximately equal to 0.4%, 0.6%, and 1.1% in the case of spandrels with masonry

arch, wooden lintel, and both masonry arch and RC bond beam, respectively.

3.9%

2.3%2.6%

0

1

2

3

4

5

eq,h

[%]

w

[%] w

[%]

Specimen W

Specimen ASpecimen AB0

0.1

0.2

0.3

0 1 2 3 4 0 1 2 3 4

r

[%]

Specimen W

Specimen ASpecimen AB

(a) (b)

Figure 14. Graphs of (a) hysteretic damping ratio and (b) residual drift versus specimen drift.



DOI: 10.1002/eqe


20/21

7. CONCLUSIONS

The main ndings of two full-scale lateral loading tests on tuff masonry walls with an opening and

different spandrel types have been presented and compared with those obtained in the past for

another specimen. Comparisons have been carried out in terms of crack patterns, force drift curves,

capacity features of idealized SDOF systems, pier drifts, hysteretic damping, and residual drifts.

Three specimens with the following spandrel types have been compared: spandrel with woodenlintel (specimen W), spandrel with masonry arch (specimen A), and spandrel with masonry arch and

RC bond beam (specimen AB). The geometry of specimens and the magnitude of axial loads on the

piers inuenced the forcedrift curves, which showed a stable hysteretic rocking behavior. No

damage was detected on spandrelpier intersection zones, conrming typical modeling assumptions

in macroelement methods. Large horizontal cracks formed at the base of piers because of rocking.

The spandrel type inuenced the observed damage as follows:

The wooden lintel did not slip off from the piers, even if its anchorage length was small. After the

spandrel panel suffered diagonal cracking, the lintel ensured the connection of the piers avoiding

the collapse of the masonry above.

The spandrels with masonry arch progressively fell down at small drift levels as a result of the

arch fragility against differential displacements at its supports, induced by rocking of piers.

The RC bond beam on top withstood large drift demands by plastic hinging, delaying the com-plete loss of connection between piers at large drift levels. Nevertheless, to inhibit the propagation

of shear cracks from the masonry to the RC bond beam, a stirrup spacing lower than those allowed

by codes is suggested for beam sections close to spandrel pier intersections.

The lowest resistance was found in the case of spandrel with only masonry arch. The RC bond beam

on top favored a peak resistance increase of about 50%. In the case of spandrel with wooden lintel, the

peak resistance was up to 1.3 times that related to the spandrel with only masonry arch.

Bilinear idealizations of forcedrift envelopes allow to suggest reference values for walls with

rocking piers and different types of spandrels, as follows:

The peak lateral resistance can be assumed to be 35%, 30%, and 45% of the total gravity load in

the case of spandrel with wooden lintel, only masonry arch, and both masonry arch and RC bond

beam, respectively. The yielding drift can be set to 0.1% in the case of spandrel with wooden lintel or masonry arch

and 0.16% in the presence of RC bond beam on top (corresponding to a 60% increase).

The ultimate drift can be assumed to be 0.8% in the case of spandrel with wooden lintel or ma-

sonry arch and 1.5% in the presence of RC bond beam.

The displacement ductility can be set to 7, 8.5, and 9.5 in the case of spandrel with only masonry

arch, wooden lintel, and both masonry arch and RC bond beam, respectively.

The overstrength factor can be set to 1.6 in the case of spandrel with wooden lintel or masonry

arch and 1.3 in the presence of RC bond beam.

The analysis of pier drifts has conrmed the effectiveness of the wooden lintel in connecting the

piers after spandrel failure and, by contrast, the gradual loss of connection induced by progressive

collapse of spandrels with masonry arches. Dealing with the energy dissipation capacity of walls

with openings and rocking piers, the authors suggest a hysteretic damping ratio equal to 2% in thecase of spandrels with masonry arch regardless of the presence of RC bond beam and 4% in the

case of spandrels with wooden lintel.

Finally, the analysis of residual drifts under varying lateral drift of specimens has allowed to

investigate the inuence of spandrels on recentering capacity of walls. The ratio between residual

and maximum drift at near-collapse can be set to 10%, 20%, and 5% in the case of spandrel with

wooden lintel, only masonry arch, and both masonry arch and RC bond beam, respectively.

Different residual-to-maximum drift ratios show that the spandrel type also affects the ability of

URM walls to be repaired after severe in-plane lateral loading.

Further research is needed to assess the inuence of the spandrel type in the case of piers subjected

to moderate-to-high compression levels, where rocking behavior may be less evident or even



DOI: 10.1002/eqe


21/21

negligible compared with that observed in this study. Moreover, lateral loading tests should be carried

out to investigate the in-plane lateral behavior of walls with steel ties in the spandrel and irregular walls

with openings.

ACKNOWLEDGEMENTS

This research was carried out in the framework of the ReLUIS-DPC 20102013 project (line AT1-1.1 Evaluation of the Vulnerability of Masonry Buildings, Historical Centers and Cultural Heritage) and PON0102324 PROVACI project (Technologies for Earthquake Protection and Valorization of Cultural HeritageSites), which were, respectively, funded by the Italian Civil Protection Department and the Italian Ministry ofEducation, University and Research.

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Spandrel Type on the Lateral Behavior