Vung Tau Go Cong Dam Vietnam - TU Delft

Vung Tau – Go Cong Dam

Vietnam

Preliminary Design Study

Appendices

Project group Flood Defence HCMC Water Resources University, Second Base

TU Delft Institute for Water and Environment Research

October 2011

Project Flood Defence HCMC Page 2 / 140

Project group Flood Defence HCMC

TU Delft

Water Resources University, Second Base

Institute for Water Environment Research

Vung Tau – Go Cong Dam

Preliminary Design Study

Appendices

Authors

Bart Dekens

Lisanne Meerdink

Gertjan Meijer

Emma Sirks

Renate van Vliet

This report is the result of the multidisciplinary project, part of our MSc-study Civil Engineering at Delft

University of Technology, and is therefore written for educational purposes.

This project is made possible by:

Ballast Nedam Iv-groep

Damen Dredging Equipment CVB-fonds TU Delft


TABLE OF CONTENTS

TABLE OF CONTENTS ................................................................................................................................ 3

REFERENCE PROJECTS ....................................................................................................................... 5 A

ESTIMATION OF SOIL LAYERING AND PARAMETERS ....................................................................... 32 B

CALCULATION OF THE CREST LEVEL ................................................................................................ 45 C

SWANONE CALCULATION OF WAVE HEIGHT .................................................................................. 47 D

CALCULATION OF THE 2% WAVE RUN-UP ....................................................................................... 57 E

CALCULATION OF OVERTOPPING ................................................................................................... 63 F

SEA LEVEL RISE ............................................................................................................................... 66 G

BEARING CAPACITY ........................................................................................................................ 71 H

SETTLEMENTS ................................................................................................................................. 85 I

STABILITY ....................................................................................................................................... 97 J

MCA SCORES ALTERNATIVES ........................................................................................................ 103 K

SHEARING .................................................................................................................................... 104 L

MICRO INSTABILITY AND SEEPAGE ............................................................................................... 106 M

SLOPE REVETMENT ....................................................................................................................... 111 N

TOE PROTECTION FOR QUARRY STONE REVETMENT .................................................................... 117 O

MODEL FOR TIDAL PREDICTION .................................................................................................... 119 P

DETERMINATION OF GRAIN SIZES FOR THE CLOSURE ................................................................... 122 Q

BOTTOM PROTECTION ................................................................................................................. 127 R

UNSTEADY STATE SALT INTRUSION MODELS ................................................................................ 135 S

MODEL AND MEASUREMENTS ..................................................................................................... 137 T

DRAWINGS ................................................................................................................................... 140 U


Project Flood Defence HCMC | APPENDIX Reference projects Page 5 / 140

REFERENCE PROJECTS A

Reference projects are studied to obtain knowledge and conceptual solutions for a design of the VT-GC

dam. Data will be collected about construction type, foundations, water and wave heights, side functions,

et cetera. The following projects will be investigated:

1. Afsluitdijk, The Netherlands

2. Eastern Scheldt Barrier, The Netherlands

3. Haringvliet Dam, The Netherlands

4. IHNC barrier New Orleans, USA

5. Saemangeum Sea Wall, South Korea

6. Gulf of Khambhat Dam, India

7. Feni River Closure Dam, Bangladesh

8. MOSE project, Italy

9. Ems Barrier, Germany

These projects will be discussed in the following sections.

A.1 Afsluitdijk (The Netherlands)

A.1.1 Introduction and basic data

The Afsluitdijk in the northern part of the Netherlands (see Figure A-1) was built in the 20th century, but

the plans to make a dam to protect the area against flooding did already exist in the 17th century

(Stichting Deltawerken Online). As with most projects with new techniques a disaster, in this case a severe

storm, had to happen before the structure was built. In 1932 the building of the 30 km long dam with 2

ship locks and 25 discharge sluices was finished (Afsluitdijk.org). Figure A-3 shows one of the ship locks.

Figure A-1: Location of the Afsluitdijk Figure A-2: Closing of the Afsluitdijk in 1932

A.1.2 Construction

Located in a deltaic area, the Afsluitdijk is built on soft soil. Soil improvement was needed for the stability

of the dam. Therefore the soil is reinforced with osier mats with dump stones. After that, the dam was

built up out of boulder clay, sand and stones. Unfortunately, a worm ruined the osier mats and the

galvanized steel was corroded. To guarantee the stability of the dam barite was used as replacing

material. The slopes are protected under water with boulders and above the water with basalt or


concrete. Due to the boulders the boulder clay remains in place with velocities up to 4 m/s. At the closure

point, this velocity was even higher: 5 to 6 m/s due to the smaller stream channel.

Figure A-4 shows the building pit of the sluices, Figure A-2 the closing of the final gap.

Figure A-3: Sluice at Kornwerderzand Figure A-4: (Poldered) building pit of the sluices

Some characterics of the Afsluitdijk are presented in Table A.1.

Table A.1: Data of the Afsluitdijk

Year of construction 1920 till 1932

Construction costs 120 million euro ≈ 80 million US$

Head difference [m] 0.3 m

Head based on 3,5 m above highest known water level

Length 30 km

Dimensions On the water 90 m width, crest height +7,50/7,80 m above MSL

Foundation Soil improvement: osier mats with dump stones

Soil conditions Soft soil

Amount of discharge 5000 m3/s in 25 locks of 12 m width and 4 m deep

Barrier type Sector gates (sluice)

Ship locks size 6.000 ton (Stevin sluice) and 6.000 + 2.000 ton (Lorentz sluice)

Ship locks capacity Stevin sluice: 3531 commercial vessels, 19921 pleasure boats Lorentz sluices: 452 commercial vessels, 37116 (in 2001) pleasure boats

Construction method Vertical closure and building pits for locks

Side functions Road (highway)

Figure A-5: Cross section Afsluitdijk, on the left the sea and the right the IJsselmeer (IJsselJLake)

A.1.3 Use in Vietnam

The length of the Afsluitdijk is comparable to the length of the VT-GC dam. Soil improvement is used,

which will probably be needed for the VT-GC dam. Also the Afsluitdijk has a road on top, which is a plan


for the VT-GC dam as well. Discharge sluices and ship locks have to be incorporated in the design of the

VT-GC dam, as is the case for the Afsluitdijk.

A.1.4 References

Afsluitdijk.org (sd). Retrieved on August 23, 2011, from De Afsluitdijk 25-05-1932: http://www.afsluitdijk.org/

Stichting Deltawerken Online. (sd). Afsluitdijk. Retrieved on August 23, 2011, from Deltawerken Online:

http://www.deltawerken.com/

A.2 Eastern Scheldt Barrier (The Netherlands)


The Eastern Scheldt barrier is a barrier with movable parts, in The Netherlands. It is part of the Delta

Works and is located in the province of Zeeland, in the Eastern Scheldt, see Figure A-6. The barrier is open

under normal conditions, to maintain salinity levels in the Eastern Scheldt. These levels are needed to

prevent damage to the existing ecosystem.

Figure A-6: Location and plan of the Eastern Scheldt barrier

A summary of information about this project is presented in Table A.2.

Table A.2: Eastern Scheldt barrier data. Information based on (Stichting Deltawerken Online, 2004) and

(Rijkswaterstaat, 2009)

Year of construction 1969 - 1986

Construction costs 2.5 billion Dutch Guilders ≈ 760 million US$

Head difference [m] Design water level seaside: NAP +5.8 m (Dutch reference level)

Head based on Flooding chance: 1/4000 year Sea level rise of 30 cm

Design rules used Designed lifetime: 200 years


Length Total width Eastern Scheldt: approx. 7.5 km Total length barrier parts: 3.6 km

Dimensions Construction height: 30 – 40 m

Foundation Soil improvement, densification of sand layer and placing of several mattresses

Soil conditions Clay, sand

Amount of discharge Almost none, Eastern Scheldt is dammed of from fresh water area’s

Barrier type Barrier with vertically movable doors 62 doors and 65 pillars

Ship locks size One small lock, length 100 m, width 15 m

Construction method

Creating of artificial islands Soil improvement and foundation works Placing of pillars Placing sill of stones around pillars

Side functions National road on top of the dam

A.2.2 Change of design

First the plan was to construct a completely closed dam. Several islands were constructed (Roggeplaat,

Neeltje Jans, Noordland). Hereafter, under political pressure to preserve the Eastern Scheldt ecosystem

and fishery, in 1976 it was decided (after investigation) to build a barrier with movable parts to maintain

salinity levels in the Eastern Scheldt.

A.2.3 Construction and foundation on soft soil

The subsoil on which the barrier was to be placed was far too weak to build for building such a heavy

barrier. Therefore, this weak layer was first completely removed by dredging and replaced by sand. This

sand was subsequently compacted with a ship which was constructed especially for this purpose, which

compacted the sand until a depth of 15 meters using vibration piles. On this stiff sand, mattresses filled

with sand and gravel were placed to prevent the scour of the sand under the pillars. On these mattresses

the pillars were placed. To ensure a good connection a mixture of sand, cement and water was used to fill

the gaps. (Rijkswaterstaat, 2009).

For a schematic view of the foundation construction, see Figure A-7.


Figure A-7: Schematisation of foundation works of the Eastern Scheldt barrier

A schematisation of the pillars is given in Figure A-8. These hollow concrete pillars are 30 to 40 meters

high and have a weight of max. 18000 tons. Around these pillars a sill of large stones, with weight up to

10000 kg, were placed which can withstand the flow velocities when one of the doors cannot be closed.

Five million tons of stone were used for this purpose.

Figure A-8: Schematic view of 2 pillars and a door of the Eastern Scheldt barrier



The Eastern Scheldt barrier is a heavy construction founded on delta soils. Soil replacement and

improvement techniques were needed in the construction process, which could prove useful for the VT-

GC dam.

The barrier is designed as an open barrier, to maintain salinity levels in the Eastern Scheldt. This concept

can be useful in the design of the VT-GC dam, because the mangrove forest may not be influenced.

A.2.5 References

Rijkswaterstaat. (2009, Februay 8). De bouw van de Oosterscheldekering. Retrieved August 25, 2011, from

Rijkswaterstaat:

http://www.rijkswaterstaat.nl/water/feiten_en_cijfers/dijken_en_keringen/oosterscheldekering/bouw_van_de_oost

erscheldekering/

Stichting Deltawerken Online. (2004). De Oosterscheldekering. Retrieved August 25, 2011, from Deltawerken Online:

http://www.deltawerken.com/Oosterscheldekering/44.html

A.3 Haringvliet dam (The Netherlands)


Figure A-9: Location of the Haringvlietdam in the

Netherlands (Ferguson, Blokland, & Kuiper, 1970)

Figure A-10: Location of the Haringvlietdam (Stichting

Deltawerken online, 2004)

The Haringvlietdam (Figure A-11) is part of the Delta Works. It is located between Goeree-Overflakkee

and Voorne Putten in the south-west of the Netherlands, see Figure A-9 and Figure A-10. The construction

of the Haringvlietdam took 14 years. The two main functions of this barrier are protection against

potential flood and drainage of water from the Rhine and Meuse River into the North Sea.


The dam is made of concrete blocks that were dumped on the sea bottom from cable cars. The holes

between the blocks were filled with sand. The specifications of this dam are listed in Table A.3.

Figure A-11: Haringvlietdam (ThinkQuest)

Table A.3: Haringvlietdam data. Information based on (Deltares) (Ferguson, Blokland, & Kuiper, 1970) (Stichting

Deltawerken online, 2004) (ThinkQuest)

Year of construction 1957 - 1971

Construction costs 720 million Dutch Guilders ≈ 220 US$

Head difference [m]

Sea level may be swept up to 5 m above A.O.D. (Amsterdam Ordnance Datum, = N.A.P. in Dutch). The sill of the sluices is located 5.5 m below A.O.D. The water level on the river side can come up to A.O.D. 2.65 m

Head based on Flood protection

Length 4.5 km

Dimensions Width 65 m (crest) to 330 m (at bed level)

Foundation The sluices were constructed on 22.000 concrete piles, some more than 20 m long (40 x 40 and 45 x 45 cm2)

Soil conditions Clay and soft sandy layers from 20 to 30 m below A.O.D.

Max wave height 4 m

Amount of discharge 17 openings (length 1 km) regulate the amount of discharge; circa. 30∙1015 m3 per year

Barrier type Concrete blocks, sand

Ship locks size Professional: CEMT VIa/Va Recreational purposes: AZM Length = 144 m, width = 16 m, sill height at N.A.P.

Construction method Cable car method

Side functions

Discharge water from Rhine and Meuse into the North Sea Special fish tunnels in pillars Traffic can run over the dam (national road)

A.3.2 Construction

First the discharge sluices were constructed in a polder in the middle of the Haringvliet. On the pile

foundation a 3 m thick concrete floor was cast. Eighteen pillars were constructed with girders in between.

These girders carry steel arms that handle the (56 x 8.5 to 10.5 m) slides. The openings have a width of 56

m. This opening was chosen to allow for ice to reach the sea in the winter time. Each opening can be


closed by or two gates. The seaward gate breaks the waves and the inner gate takes the excess volume

during storm tides. Figure A-12 shows the dross section of one of the discharge sluices.

The bottom of the sluice on the seaward side is protected with a 65 m concrete slab that reaches to a

filter bed. On the river side a concrete slab on a shallow foundation reaches 33 m onto another filter bed.

Figure A-12: Cross section of Haringvliet sluice at base of pier (Ferguson, Blokland, & Kuiper, 1970)

To construct the dam large concrete blocks were dropped into the water from a cable car, see Figure

A-13. To close the southern gap, sand that was reclaimed from other parts of the sea bottom was raised

until a dam body appeared. Wind and water made this southern part look more like a dune than a dam.

The 1 km wide northern gap was closed by dumping over 100.000 concrete blocks (2500 kg each) from a

cable car. Figure A-13 shows how the blocks were dumped. Holes between the blocks were filled with

sand. To reduce the strong currents the locks were open during gap closure. After construction of the

body a road was made on top of the dam.

Figure A-13: Cable car method (Stichting Deltawerken online, 2004)


A.3.3 Discharge sluices

The gates of the seventeen discharge sluices are closed at high tide to prevent salty water from entering

the Haringvliet lake and form a protection against flooding. At low tide discharge sluices are only partially

opened to discharge water. The sluice gates are opened in conjunction with three weirs on the Rhine

River between Arnhem and Rotterdam. This makes the distribution of outflows of river water possible in

the desired manner.

After construction tides had no influence anymore in the area behind the dam. Plants and fish that

depended on salty water didn’t survive. Other species however took their place.

On the seaward side of the sluices there is a 7 m deep step to break the waves partially and reduce the

strain on the gates by approximately 50%.


This barrier has seventeen discharge sluices to discharge water from the Rhine and Meuse River. The VT-

GC dam will have to discharge a large amount of water from the rivers surrounding Ho Chi Minh City.

The soft delta soils require a proper foundation and bed protection. The Haringvlietdam is constructed on

clay and soft sandy layers and has a proper foundation and bed protection

The barrier was constructed using concrete blocks and sand. These materials are available almost

everywhere.

A.3.5 References

Deltares. (sd). Haringvlietsluizen. Retrieved on August 25, 2011, from Deltares Public Wiki:

http://public.deltares.nl/display/Dashboard/Public+Wiki

Ferguson, I. H., Blokland, I. P., & Kuiper, I. d. (1970). The Haringvliet Sluices. Den Haag: Rijkswaterstaat.

Stichting Deltawerken online. (2004). Haringvlietdam. Retrieved on August 24, 2011, from Deltawerken online:

http://www.deltawerken.com/Haringvlietdam/46.html

ThinkQuest. (sd). Haringvlietdam. Retrieved on August 24, 2011, from Projects for Students by STudents:

http://library.thinkquest.org/19846/data/ne/delta/haring/haring.htm

A.4 IHNC barrier, New Orleans (United States of

America)


The IHNC barrier was built after hurricane Katrina caused a major flood in New Orleans in 2005. The

barrier (Figure A-14) will seal off the inner barrier navigation channel to the harbour of New Orleans from

storm surges from the Gulf of Mexico. This barrier was built on soft delta soils and is partially open for

navigation and recreational shipping. In case of a storm surge movable gates can close off the openings

(Figure A-15). The specifications of the barrier are listed in Table A.4.

The very weak subsoil resulted in a concrete construction (Royal Haskoning, 2008) instead of a

conventional dam. The barrier was made of almost 1400 concrete piles with a length of 42 to 49.5 m and

diameter of 1.7 m. The wall is supported by 88 m long steel batter piles. To withstand wave and surge

forces it is necessary for these piles to be very long.

The northern and southern end will tie into existing barriers.


The design and construct method was used for the construction. One of the reasons this design was

chosen was the aggressive building schedule of the contractor: the barrier had to be finished before the

hurricane season.

Figure A-14: The IHNC Surge Barrier, being built by the US

Army Corps of Engineers (Wikipedia)

Figure A-15: Sector gates of the IHNC Surge Barrier

Table A.4: IHNC barrier data. Information based on data from (US Army Corps of Engineers) (Royal Haskoning, 2010)

(Royal Haskoning, 2008) (Brown, july 2009)

Year of construction 2008 to 2011

Construction costs 700 million US$

Head difference [m] 6m above mean sea level; surge wave heights 2 to 3 m The piles rise 4.3 m above water level, the concrete cap on top of them adds 2 m.

Head based on Hurricane and storm damage risk reduction to 1% per year

Length 3.2 km

Foundation

The deep portion of the channel was partially filled using rock and sand to serve as a foundation. For the lift gate and navigation opening with sector gates: concrete piles. Estimated settlements: 80 to 150 mm over 50 yr. If settlements are higher panels can be installed to restore the wall to its proper height.

Soil conditions Soft alluvial soils

Barrier type Lift gate for recreational boat access, sector gates in the navigation opening (46 m wide)

Construction method

Circular piles diam. (1680 mm) were supplied (by boats) in sections of approximately 4.8 m and driven in. Gaps between the piles were filled with smaller square shaped piles (side 460 mm) and grout. Batter piles (diam. 910 mm) on the inland side serve for stability. The piles are filled with steel reinforcing bars and concrete.

Side functions Roadway for operation and maintenance crews


This barrier is constructed on very soft soils. Therefore the choice was made for a concrete solution. The

VT-GC dam will also be constructed on soft river delta soils. This light weight design can be an example for

a construction on soft alluvial soils as the settlements are low and it can be easily restored to the required


height if settlements are larger than expected. The VT-GC dam doesn’t have to be built as quickly as the

IHNC barrier.

Parts of the structure are open for navigation purposes and will be closed in case of a storm surge. An

opened structure for navigation is favourable as is decreases port delays. Depending on how frequently

they have to close this could work for the VT-GC dam as well.

For a length of 30 km this will be an expensive solution.

Figure A-16: the IHNC barrier (Brown, july 2009)

A.4.3 References

Brown, J. L. (july 2009). Corp's Largest Design/Build Civil Works Project To Shield New Orleans. Civil Engineering, 22-

23.

Engineers, U. A. (Regisseur). (sd). Lake Borgne [Film].

Royal Haskoning. (2008, 12 4). New Orleans IHNC Barrier. Retrieved on August 23, 2011, from WaterLink

international: http://www.waterlink-international.com/news/id351-New_Orleans_IHNC_Barrier.html

Royal Haskoning. (2010). IHNC Barrier under constructon. Retrieved on August 24, 2011, from Royal Haskoning:

http://www.haskoninginc.com/en-gb/Projects/Pages/StartofIHNCbarrier.aspx

Wikipedia. (sd). IHNC Lake Borgne Surge Barrier. Retrieved on August 24, 2011, from Wikipedia:

http://en.wikipedia.org/wiki/IHNC_Lake_Borgne_Surge_Barrier

A.5 Saemangeum Sea Wall (South Korea)


To create new farmland and a fresh water reservoir, the South Korean government started in 1991 with

building a connection between two land heads (Wikipedia). This connection was finished in 2010. Now

the government is developing the area enclosed by the dam, which will consist of a new city and space for

agrarian and industrial development. (Nederlandse Overheid) The big seawall will connect 3 islands and

the two land heads and will consist of 4 parts.


Figure A-17: Location of the Saemangeum seawall. (Sande, 2006)

The four dam parts are different in height due to differences in bottom depth. The dam with the lowest

mean average height has a height of 16 m and a base width of 198 m. The highest one is 35 m high and

has a base width of 290 m. Also the height of the part above the water level differs: between EL +8.5 and

EL +11 m.

A road is constructed on top of the dam.

Table A.5: Saemangeum Sea Wall data

Year of construction 1999 till 2006 (basic construction) / 2010 (opening)

Construction costs 2 trillion KRW on Construction Dam, 220 billion for strengthening the dam

Length 33 km

Dimensions Dike dimensions vary at each dike part: height of 16-35 m and bottom width 198-290 m. The height above water is between EL(+)8,5 and EL(+)11,0 m.

Foundation Bottom protection: mat paving

Soil conditions Soft sand layer on top of a rock bed layer (±50 m below MSL)

Average wave height Tidal wave = 8 m

Construction method Fill-up construction method with stone and sea sand

Side functions Road on top of the dike

A.5.2 Foundation and gap closure

The foundation of the dam consists of a mat paving on sandy soil/mud. Stones are used to protect the

foundation from the tide. The dam is made of stones (riprap) and the gap is filled up with dredged sea

sand. The final construction is made by placing rocks to protect the dam construction for the waves.

At the location of the two closing gaps, a high velocity of 6.9 m/s and with that a lot of erosion is

expected. (Eo & Kim, 2003) Therefore bottom protection is used over a length of 200 m. The gaps are

closed in steps (Figure A-18), in which the orange part is the last. The filling material for the closing gap


will consist of 50% rocks weighing 3-5 ton and 50% gabions weighing 3 ton. The gap is 100 m wide and 10

m deep.

Figure A-18: Top view of the final gap closure. Green parts are closed first, orange last..

Table A.6: Main construction works of Garyeok Sluice, dewatering sluice (Saemangeum Project Office)

Sluice gate in sea side RADIAL GATE 30 m (W) × 15 m (H) × 8 pairs

Sluice gate in freshwater side RADIAL GATE 30 m (W) × 12.5 m (H) × 8 pairs

Vessel passage gate MITER GATE 4 m (W) × 13.85 m (H) × 4 pairs

Water level controlling and vessel passage gate

SLIDE GATE 2.5 m (W) × 2.5 m (H) × 2 pairs

Water gate for lower section drainage

ROLLER GATE 6 m (W) × 2.5 m (H) × 1 pair Fluid controlling valve Φ1,000 × 1

Figure A-19: Sluices at Sinsi island


The Saemangeum sea wall has almost the same length as the VT-GC dam. Also this wall is located at the

end of an estuary, more or less offshore. It has discharge facilities in it and a road on top of it. Therefore it

has great similarities with the VT-GC project.

A.5.4 References

Eo, D. S., & Kim, J. S. (2003). Hydraulic structures for Semangeum project. Gyeonggi, Korea: Agricultural and Rural

Infrastructure Corporation.


Nederlandse Overheid. (sd). Zuid-Korea: inpoldering- en ontwikkelingsproject Saemangeum. Retrieved on August 24,

201 1,from Agentschap NL, Ministerie van Economische zaken, Landbouw en Innovatie:

http://www.agentschapnl.nl/onderwerp/zuid-korea-inpoldering-en-ontwikkelingsproject-saemangeum

Saemangeum Project Office. (sd). Business, status of construction. Retrieved on August 25, 2011, from Saemangeum

Project Office: http://www.isaemangeum.co.kr/eng/sub/02_business_06.html

Sande, M. v. (2006). Computational modeling final gap. TU Delft: TU Delft.

Wikipedia. (sd). Saemangeum Seawall. Retrieved on August 24, 2011, from Wikipedia:

http://en.wikipedia.org/wiki/Saemangeum_Seawall

A.6 Closure dam Gulf of Khambhat (India)


The Gulf of Khambhat is located in the western part of India, in the state of Gujarat. At the time of the

first feasibility study (1989) it would have been the largest closure of a tidal estuary in the world, although

it is not constructed. However, the project is reported to be economically and technically feasible

(Haskoning, 1998). Construction will start in the near future.

It would have a length of 60 km, 30 km through deep water and 30 km through shallow water. It functions

would be twofold: 1) creating water reservoir, and 2) electric power generation. The whole construction

consists of a dam, discharge sluices, locks and a power generator. The reservoir will be split in 2 parts, one

for power generation (salt part) and one for water supply (fresh part), see Figure A-20.

Figure A-20: Location of the Khambhat dam Figure A-21: Layout of the Khambhat dam

A summary of information about this proposed project is presented in Table A.7.


Table A.7: Khambhat closure dam data, based on (Broos & Wiersema, 1998) and (Haskoning, 1998)

Year of construction Construction starts in 2011 (Projectmonitor.com, 2002)

Construction costs Rs 19.253 Crores = 192.530 million rupees ≈ 4.2 billion US$

Head difference [m] Max. 9 m Tidal head difference: 8 m, during springtide 10 m

Length approx. 60000 m (total length of dams)

Dimensions dam

Base width: variable Height: 13 meters above MSL. Height in shallow parts: 18 m, Height in deep parts: 53 m. Slopes: 1:5 (seaside), 1:3 (river side) Crest width: 5 m

Foundation No additional foundation works besides bottom protection

Soil conditions River sediment: sand and clay, susceptible to erosion

Wave height Significant wave height: 3.5 m

Amount of discharge 9.6∙10

6 m

3/s (1/100-year value)

Required width of discharge structure: 700-800 m 65 discharge openings with a width of 17 m. Radial gates, sill level -11 m

Barrier type Rock/earth dam

Ship locks size Length: 200 m Width: 35 m Sill level: GTS -10 m

Ship locks capacity Max. allowance ships: 50.000 deadweight tonnes

Construction method Placing bottom protection Building of dams, layer for layer Filling the closure gap with sluice caissons

Side functions Creating water reservoir Power generation Road & railway

A.6.2 Construction method and cross sections

Bottom protection is used over the whole alignment to prevent local scour. This protection is made of

geotextile bag filled with sand. The dam has different cross sections at different locations. A cross section

of the part of the dam at shallow parts in the area where gullies occur (section of 6 km long) is given in

Figure A-22. The core of the dam is made out of rock.

Figure A-22: Proposed cross section of the Khambhat dam in shallow parts (Broos & Wiersema, 1998)


At the closure, where the tidal power generator will be, a different profile is used. This profile is given in

Figure A-22. The core of the dam is made out of rock.

Figure A-23: Cross section of the dam near the tidal power generator (Broos & Wiersema, 1998)

The large part of the dam (30 km) which is located in the shallow area where no gullies occur (30 km long)

will have the same profile as in Figure A-22, but the core will consist of clay instead of rock. This is because

the local bottom consists of clay, velocities are lower than at other parts of the dam, rock is expensive and

because most of the alignment is dry during the tidal cycle.

On top of the dam a protective layer is placed. Dam protection against waves requires stone with a

diameter of 1.2 m to withstand the design storm waves.

A.6.3 Closing methods

The maximum acceptable flow velocity during closure is adopted as 6 m/s, to keep the handling of the

rock manageable. This flow occurs when the width of the opening is 10 m. This gap has to be closed

suddenly. A suggested plan is doing this with the use of sluice caissons which can be closed suddenly by

lowering the gates (Haskoning, 1998). The caissons will be placed on a sill. A cross section of such a

caisson is given in Figure A-24.

Figure A-24: Cross section of a Khambhat sluice caisson


Another solution for the closure is the use of ‘superbags’ According to (Broos & Wiersema, 1998) this is a

cheaper and faster closure solution. Such a geotextile bag (dimensions approximately 25 x 50 x 22 meters,

see Figure A-25) is filled with geotextile tubes filled with local sand. The bags are placed in the closure gap

with a floating crane.

Figure A-25: Schematic figure of a superbag, filled with geotextile tubes filled with sand


The Khambhat dam will be a large dam, consisting of two sections of approximately 30 km. It is founded

on soft soils, susceptible to erosion. Special attention had to be paid to prevent scour, something that will

also be important when construction the VT-GC dam. Also several interesting closure techniques are

investigated, like the use of gated caissons and the use of large geosynthetics bags filled with local

material. Finally on the dam there will be a road/railway, something that is planned for the VT-GC dam as

well.

Because the dam is not yet constructed the efficiency of the Khambhat construction techniques cannot

yet be proven. This makes the project somewhat less interesting for the VT-GC dam.

A.6.5 References

Broos, E., & Wiersema, K. (1998). Closure of the gulf of Khambhat. Delft: TU Delft.

Haskoning, D. F. (1998). Khambhat gulf development project (Kalpasar), pre-feasibility report, executive summary.

Projectmonitor.com. (2002, January 1). Kalpsar project to be commenced in 2011. Retrieved on August 11, 2011, from

Projectmonitor.com: http://www.projectsmonitor.com/detailnews.asp?newsid=3407

A.7 Feni River Closure Dam (Bangladesh)


The Feni River closure dam is a dam located at the mouth of the Feni River in Bangladesh. It is built in a

tidal area. Main goal of the dam is to form a reservoir filled by the Feni and Muhuri rivers for irrigation

purposes. Besides this main function it functions as protection against tidal waves of extreme heights.

A regulator in combination with a new diversion canal next to the dam was constructed to discharge

surplus river water in the rainy season.


Figure A-26: Location and plan of the Feni River dam (scales on both figures should be 10 times bigger, distances are

too small in the figure (Stroeve, 1993)


Table A.8: Feni closure dam data. Information based on data from (Stroeve, 1993)

Year of construction 1985

Construction costs 478.1 million Bangladesh Taka (base construction) ≈ 6.4 million US$

Head difference [m] Mean sea level in the wet season 1.0 m above MSL dry season Max height of water 9.1m, height dam 10.4 m

Design rules used

Probabilistic design rules:

For a monsoon situation having a return period of 50 years only 2% of the waves map overtop.

For a cyclone with a return period of 20 years wave overtopping should not result in an average water depth in the polder exceeding 1.0 m.

For a cyclone with a return period of 40 years the crest level should not be lower than the still water level.

Length Approx. 1200 m (main dam)

Dimensions

Base width: 150 m Height: 14.5 meters Slopes 1:5-6 (seaside), 1:4 (river side) Crest width: 6.0 m

Foundation No additional foundation works besides bottom protection

Soil conditions Predominantly loosely packed silts, sandy silts


Average wave height Significant wave height Hsig = 1.70 m

Amount of discharge Separate channel for discharge next to the dam

Barrier type Earth dam, constructed from local (weak soil) materials

Ship locks size No locks

Ship locks delay No locks

Ship locks capacity No locks

Construction method

Levelling of river bottom to uniform profile

(Almost) immediate closure with small dam during neap tide

Building of winter springtide dam on top of ‘neap tide’ dam

Building of full size dam

Side functions Creating water reservoir

A.7.2 Earthquake considerations

Although the risk of earthquakes was rather small some precautions are taken. Because soil improvement

was not economically feasible measures were taken to minimize the consequences. These measures

consisted of using a geotextile at the base of the dam and the use of ‘earthquake resistant’ earth, which

would not collapse during an earthquake.

A.7.3 Hydrological problems

Modelling showed that initial narrowing was not possible, because the delicate balance of plates and

gullies would be disturbed too much. Predominantly present soils at the closure location are loosely

packed silty sands and sandy silts. During closure therefore the danger for soil slides and mud waves

existed. Therefore was decided to close the whole river (almost) instantly.

A.7.4 Construction method

First bottom protection (bamboo frames with composite filter and filled with river boulders) were placed

on the river bed. Afterwards, the bottom was raised to equal height by placing a sill made of dumped

boulders. This was done because a uniform bottom profile was needed to be able to build the neap tide

dam in construction phase 1. The sill was then covered with another layer of bed protection matrasses.

The dam itself was built in three stages:

1 Neap tide dam: the whole gap was closed almost instantly during neap tide with clay bags,

using 1200 labourers.

2 Winter spring tide dam: the neap tide dam was elevated with clay to be able to withstand

the spring tide which would occur 8 days later. This dam was built out of local clay.

3 Main dam: before the monsoon season started the dam was heightened to the final height.

The fill material used was local dredging material from the river bottom (sandy silts /silty

sand) and clay excavated from one of the shores.

Figure A-27 gives a cross section of the final dam. In this picture also the neap tide dam (A) and the winter

tide dam (B) are visualised.


Figure A-27: Cross section of the Feni dam, right side is seaside. (Stroeve, 1993)

After this third stage slope protection was placed at both sides of the dam using concrete blocks, bricks

and boulders, to protect the dam from waves from both directions.

Coastal embankments were constructed at both ends of the dam, see Figure A-32, to connect the dam

with existing dams. In Figure A-28 the whole construction process is schematically visualized.


The dam is founded on a river estuary with a risk of unfavourable change of the gully pattern. The dam is

placed in Asia, where conditions could be similar to the ones in Vietnam (monsoons/storms)

Besides that, the dam is build out of local soil, which is advantageous for the building time and the costs.

A.7.6 References

Haskoning. (1985). Report on the construction of the Feni River closure dam.

Stroeve, F. (1993). The Feni River closure dam reviewed, Delft: TU Delft.


Figure A-28: Construction method of the dam. Building stages are chronologically visualized from front to back.

Seaside is at the left side (Haskoning, 1985)

A.8 The MOSE project (Italy)


Figure A-29: Location of the MOSE Project (BBC News, 2003)


In 1966 a record high tide caused major trouble to the city of Venice. In the 30 years after that Venice has

been plagued by several floods. The city has subsided 23 centimetres over the past 100 years and high

waters occur more frequently. In 2007 the construction of the MOSE project started.

The 78 mobile steel barriers of the MOSE project will lie on the seabed most of the time. When Venice is

threatened by high tides or storm they are filled with air to create a dam. A 1.5 km long dam can be

created by these floating barriers.

Table A.9: Information about the MOSE project

Year of construction 2007-2012

Construction costs 4,272 million euro ≈ 6,155 million US$

Head difference [m] The gates operate from 1.10 up to 3 m tide

Head based on Venice has sunk 23 cm during the past 100 years High waters occur more frequently Expected sea level rise has been taken into account

Length 1.5 km

Dimensions Large metal box type structures, with a length of 20 to 30 m, a width of 20 m and thickness of about 5 m

Foundation The gates rest on concrete piles

Barrier type Oscillating buoyancy flap gates

Side functions Navigation, water exchange between sea and lagoon, no influence on landscape

A.8.2 Construction

First a series of measures to protect the coast was implemented. This included reconstruction of 45 km of

beach, restoration of 8 km of dune and reinforcement of 11 km jetty and 20 km seawall. Outside of

Malamocco and Chioggia two breakwaters have been constructed to protect the locks. The construction

work on the barriers proceeded parallel at all three inlets.

A.8.3 Operation of the gates

The oscillating buoyancy flap gates do not modify the water exchange between the sea and the lagoon.

Because they are lying in the sea bottom most of the time they do not alter the landscape and do not

interfere with port activities.

The gates are large metal box type structures, with a length of 20 to 30 m, a width of 20 and thickness of

about 5 m. In normal conditions they are full of water and rest in prefabricated concrete caissons on the

seabed. When a tide higher than 1.10 m is forecast compressed air is introduced into the gates and

therefore they rise from the seabed. Rising takes about 30 minutes, retracting 15.

Because the construction consists of multiple gates not all inlets have to be closed at the same time and

also parts of inlets can be closed off. The use of the gates is estimated to be limited to 3-5 times per year.

To guarantee navigation while the mobile barriers are in operation, three ship locks are constructed. At

Malamocco there is a lock for large ships, at Lido and Chioggia these are locks for recreational shipping,

fishing and emergencies.


Figure A-30: Operation of the gates (BBC News, 2003)

Figure A-31: Ship locks guarantee navigation when gates are closed (Nuova, 2008)

A.8.4 Criticisms

The main criticism of the project is that it is very expensive compared to alternative solutions.

Management and maintenance costs are very high.

Levelling and reinforcing will be carried out at the seabed where the barrier is constructed and this will

influence the ecosystem of the lagoon.



It cannot yet be said if the result of this solution is as good as expected for the environment as the project

has not yet been finished. Softer and less expensive measures could have been taken, which would also

provide a good exchange of water between sea and lagoon.

This solution doesn’t delay port activities and allows for the tide to reach the lagoon. These are some

preferences for the VT-GC dam. In Vietnam an earth structure could be combined with movable parts, so

that influence on navigation will be less than in case of ship locks. Ship locks can be applied for when the

barrier is closed.

The solution is quite expensive, both in construction and in management and maintenance, which makes

it less desirable in Vietnam.

A.8.6 References

BBC News. (2003). In Pictures: Protecting Venice. Retrieved on August 25, 2011, from BBC News:

http://news.bbc.co.uk/2/shared/spl/hi/pop_ups/03/europe_protecting_venice/html/2.stm

Nuova, C. V. (Director). (2008). Venice Lagoon: The MOSE System for the Defence Against High Water [Motion

Picture].

Wikipedia. (2011). MOSE project. Retrieved on August 26, 2011, from Wikipedia:

http://en.wikipedia.org/wiki/MOSE_Project

A.9 Ems barrier (Germany)


The Ems Barrier is located on the Ems River, near Gandersum in Germany, see Figure A-33. This dam was

built for storm flood protection on the Ems River. The Ems barrier consists of seven barrier spans. Two of

the seven openings are being used for navigation.

The barrier is made of underwater concrete, reinforced concrete (ca. 46 000m3) and steel for the service

bridges (ca. 1650t). Connection levees connect the barrier (with a length of 476m) on both sides to the

main levees along the river. The total distance between the North- and Southbank is 1040 m.


Figure A-32: Ems barrier (Meinhold)


Figure A-33: Location Ems barrier in Germany (Meinhold) Figure A-34: Location Ems barrier (Meinhold)

Table A.10: Ems barrier data. Information based on data from (Meinhold) and (NLWKN)

Year of construction 1998 to 2002

Construction costs €220 Million euro, costs of mobile gates circa 10% ≈ 150 million US$

Head difference [m] HThw (Storm flood and wave effects): NN+6.60m, closes at: NN+3.70m MThw: NN +1,60 m NN is the German reference level

Head based on Storm flood protection

Dimensions L: 476 m

Foundation Sheet piling and steel piles (circa 10 000 t), dredging works (ca. 400 000 m

3),

riprap (ca. 150 000 m3), underwater concrete for sill and pier foundation

(ca. 12 000 m3). Sill depth from under NN -5.0 up to NN-9.0 m

Barrier type Rotary segment gate for main shipping, segment gate for inland shipping, vertical lift gates

Ship locks size

Ships with a draught of max. 8.50 m, width of 38 m and length of max. 300 m. Max. Storage level: NN +2.70 m Main shipping opening: width: 60.0 m, sill depth: NN-9.0 m, width inland navigation opening: 50.0 m, sill depth: NN -7.0 m

Construction method The sill is immersed and connected to the piling with underwater concrete by divers.


Figure A-35: Top view Ems barrier (Meinhold)

Figure A.36: Front view Ems barrier (NLWKN)

Figure A.37: Cross section rotary segment gate (Meinhold)



The idea of a rotary segment gate and a segment gate with a certain height above the water may be used

for the VT-GC dam as it allows ships to pass with a short delay compared to a solution with ship locks.

With high sea levels the gates will be closed. As high tides at the location of the VT-GC dam occur a lot

more frequently than storm surges on the Ems River the gates will have to close more frequently. The

closed gates will cause delay for shipping.

The structure allows for water to pass, which would be of positive influence for the mangroves behind the

VT-GC dam. The concrete columns that hold the vertical lift gates need a strong foundation as the weight

is spread out over a small area.

The connection between the structure and the levees can be used for a partially open structure. This

connection is one of the weak spots in the structure.

A.9.3 References

Meinhold, W. (sd). Project Review. Ems Barrier. Federal State Lower Saxsony, Germany: PIANC.

NLWKN. (sd). Das Emssperrwerk bei Gandersum - Stormflutschutz und Staufunktion. Retrieved on August 23, 2011,

van Nedersachsischer Landesbetrieb fur Wasserwirtschaft Kusten- und Naturschutz:

http://www.nlwkn.niedersachsen.de/live/live.php?navigation_id=8434&_psmand=26

Project Flood Defence HCMC | APPENDIX Estimation of soil layering and parameters Page 32 / 140

ESTIMATION OF SOIL LAYERING AND PARAMETERS B

B.1 Method

There is no information about the soil conditions at the exact location of the barrier. Therefore an

assumption will be made, based on the following sources of information:

1. Thu Bo barrier data (for the deeper layering and parameters)

2. Information from the Southern Institute of Water Resources Planning

a. Boring at Go Cong (for the layering)

b. Seismic data near Vung Tau (for the layering)

c. Material in the top 30 cm of the seabed

3. A report about a Vung Tau port study (Lyon Associates inc. Consulting Engineers, 1974) (for the

layering op top layers and parameters)

4. For unknown soil parameters: the Dutch geotechnical code NEN 6740, where typical Dutch

parameters are given for several soil types (Dutch Normalisation Institure NEN, 2006)

By putting these three sources of information together a soil layering profile and characteristic values for

soil parameters will be estimated.

B.2 Gathered information

B.2.1 Thu Bo barrier data

From the Thu Bo barrier project the soil layering and parameters as in Table B.1 are derived.

Table B.1: Soil parameters for assumed dam subsoil. Data of layers 1, 2, 3, 5 and 6 based on (Soil investigation Thu Bo

barrier, Book 1: (in Vietnamese), 2010). Data of layer 4 based on (Soil investigation Thu Bo barrier, Book 2: Appendix

experimental results (in Vietnamese), 2010). SPT blow count-values based on (Water Resources University).

Category Parameter Unit Layer 1 Layer 2 Layer 3 Layer 4 Layer 5 Layer 6

Layer

Geological era

[-] Holocene Holocene Pleistocene Pleistocene Pleistocene Pliocene

Description

[-] Clay mud Clay

Sand, fine to medium, loose to medium compacted

Clay

Sand, fine to medium, medium to dense compacted

Clay

Depth compared to surface

[m] 0 -8.5 -16 -58.5 -63.5 -79

Depth compared to reference

[m] +1.5 -7 -14.5 -57 -62 -77.5

Thickness

[m] 8.5 7.5 42.5 5.0 15.5 -

Contents

Sand [%] 16.4 43.8 80.5 83.1 29.1

Silt [%] 30.4 24.5 11.2 9.4 25.3

Clay [%] 53.2 31.7 8.3 7.5 45.6

Unit weights Saturated γw [kN/m3] 14.66 18.63 19.19 19.59 19.18 19.49

Dry γd [kN/m3] 8.60 15.01 15.83 16.23 16.10 16.40

Atterberg limits

Liquid limit LL [%] 61.3 38.2 - 39 - 40.2

Plastic limit PL [%] 30.9 19.9 - 16.8 - 19.8


Category Parameter Unit Layer 1 Layer 2 Layer 3 Layer 4 Layer 5 Layer 6

Plasticity index

PI [%] 30.4 18.3 - 22.2 - 20.4

Direct shear test?

Cohesion c [kPa] 7.45 30.48 5.10 29.50 5.19 35.38

Friction angle ϕ0 [deg] 4.43 16.77 28.57 16.91 29.62 17.10

UU-triaxial test data

Cohesion cuu [kPa] 11.76 33.22 - 33.22 - 48.51

Friction angle ϕuu0 [deg] 1.48 2.08 - 2.06 - 2.22

CU-triaxial test data

Cohesion ccu [kPa] 11.76 30.77 - - - 51.25

Friction angle ϕcu0 [deg] 11.92 19.37 - - - 19.53

CU-triaxial effective parameters?

Cohesion c' [kPa] 8.82 26.85 - - - 47.43

Friction angle ϕ'0 [deg] 23.15 22.77 - - - 22.72

Permeability k [cm/s] 7.6E-06 4.6E-07 5.8E-03 - 6.3E-03 3.2E-07

Porosity n [%] 66.3 41.9 38.7 37.8 38.9 37.8

Void ratio e0 [-] 1.981 0.725 0.633 0.608 0.641 0.611

SPT blow count

N [#] < 2 < 10-19 11-29 - 28-45 -

To be able to make predictions about settlements data is required about the consolidation behaviour of

the layers. From the appendix of the soil investigation of the barrier (Soil investigation Thu Bo barrier,

Book 2: Appendix experimental results (in Vietnamese), 2010) results of consolidation tests are used to

make an estimate.

The model to describe the settlements will be the Bjerrum model, an easy and internationally known

consolidation model. Because not much is known about the consolidation behaviour of the soil, a simple

model which can easily be verified is better to use then an extensive and difficult model for which the

input parameters are not known. The Bjerrum-formula is given by:

(B.1)

where is the initial void ratio, the primary consolidation parameter, the effective stress on the

sample after loading, the stress at time (before loading), the creep parameter, the time

and the initial time.

The relation between void ration and strain is given by:

(B.2)

Because the consolidation tests are performed with a maximum pressure of

only

reliable values for can be obtained for layers above approximately 10 – 15 m below surface level,

because otherwise the original preconsolidation pressure is not reached (at 15 m depth approx. 250 kPa.

For the following Thu Bo samples the primary consolidation coefficients are calculated. It is assumed that

there is no creep, so . Therefore:


can now be determined by fitting a straight line though a graph with on the vertical axis the void ratio

and on the horizontal axis the 10-log of the total stress .

The estimation for the two layers (mud clay ( layer 1) and clay ( layer 2)) is given in Table B.2.

Table B.2: Estimation of Cc and data about used samples Thu Bo barrier

Thu Bo borehole

Depth Soil type Estimated preconsolidation pressure

Estimated Cc (based on Figure B-1)

Average Cc per soil type

[#] [-] [m] [-] [kPa] [-] [-]

1 TBIII-4 2.4 Clay mud 114 1.10

1.01

2 TBIII-5 2.4 Clay mud 114 1.10 3 TBIII-2 2.4 Clay mud 113 1.14 4 TBIII-1 8.7 Clay mud 139 0.96 5 TBIII-3 7.4 Clay mud 118 0.90 6 TBIII-6 7.4 Clay mud 118 0.86

7 TBIII-5 4.9 Clay 157 0.13

0.10 8 TBIII-1 12.0 Clay 204 0.09 9 TBIII-3 12.4 Clay 211 0.08 10 TBIII-3 9.9 Clay 148 0.08 11 TBIII-6 9.9 Clay 162 0.14

The graphs on which the estimations are based are given in the Figure B-1.

Sample 1

Sample 2

Sample 3

Sample 4

Sample 5

Sample 6


Sample 7

Sample 8

Sample 9

Sample 10

Sample 11

Figure B-1: Graphs of consolidation measurements. Blue line is measured data, dotted red line the estimated

preconsolidation pressure, and green line the linear fit to estimate Cc

Finally values for the vertical consolidation coefficient are determined. The values from CU-triaxial tests

at the Thu Bo barrier (Soil investigation Thu Bo barrier, Book 2: Appendix experimental results (in

Vietnamese), 2010), and averaging of the -value, are given in Table B.3.

Table B.3: Determination of the coefficient of consolidation cv based on Thu Bo barrier data

Layer Borehole Depth Measurement Average Average per layer Thu Bo 1 2 3 cv [#] [-] [m] m

2/day m

2/day m

2/day m

2/day m

2/day m

2/s

1 TBIII-1 7 0.14 0.19 0.05 0.13

0.14 4.52∙10-9

1 TBIII-2 2 0.14 0.19 0.06 0.13 1 TBIII-3 4.5 0.14 0.19 0.06 0.13 1 TBIII-4 2 0.14 0.21 0.08 0.14 1 TBIII-6 4.5 0.2 0.26 0.09 0.18

2 TBIII-1 14.5 0.17 0.09 0.09 0.12 0.12 3.70∙10-9


B.2.2 Information from the Southern Institute of Water

Resources Planning

B.2.2.1 Boring Go Cong

In a presentation provided by the Southern Institute of Water Resources Planning part of a boring at Go

Cong is given, see Figure B-2. In Figure B-3 a schematisation is given, and in Figure B-7 the location of the

boring is given.

The pleistocene sand layer is not visible in the boring, so this layer has be lower than -18 m deep.

Figure B-2: Boring Go Cong

0

-2

-4

-6

-8

-10

-12

-14

-16

Go Cong

Boring

+0.0

-2.7

-11.7

-14.7

Key

Clay, silty

(67.5% clay, 25.6% silt, 6.5%

sand)

Sand, muddy

(43.6-57.9% sand, 24.3-38.9%

silt, 17.5-17.7% clay)

Mud, with sands, silts and clays

(17.4-41% sand, 22.7-48.0% silt,

20.2 – 63.5% clay)

Sand

(93,7% sand, 6.3% silt)

Ho

loce

ne

Ple

isto

ce

ne

-18 -18.0

Reference level

unknown

Figure B-3: Schematisation boring Go Cong


B.2.2.2 Seismic data near Vung Tau

In a presentation provided by the Southern Institute of Water Resources Planning two seismic profiles

near Vung Tau are given, see Figure B-4 and Figure B-5. For the location, see Figure B-6.

Allthough the reference level is not given from these figures is estimated that in deep gully west of Vung

Tau the Holocene sediments extend to a maximum depth of approximately 28 m below the sea level, and

even more to the west to a depth of 23 m below sea level. The thickness varies between 6 and 16 m.

Figure B-4: Seismic profile 26

Figure B-5: Seismic profile 27


Figure B-6: Location of the seismic profiles

Several investigations are performed in the vicinity of Vung Tau. Based on these data a map with the

thickness of the Holocene layer is drawn, see Figure B-7.

In the gully near Vung Tau the Holocene layer is approximately 10 – 12 m thick, and some kilometres

more to the east 14 m. Depths are estimated on respectively -25 and -20 m below sea level.

Figure B-7: Map with equivalent thickness lines of Holocene layers


B.2.2.3 Top layer material

In Figure B-8 the material in the top 30 centimetres is given. This information is based on a geological

survey and provided by the Southern Institute of Water Resources Planning. They estimate 2 - 5 m of

young Holcene layers (from rivers) on top of 5 - 12 m of old Holocene layers consisting of muddy

sand/clay.

Figure B-8: Material on sea bottom between 0 and 30 cm below sea bottom. gS = Sandy gravel, (g)S = Sand with

pebbles, mS = Sandy mud, sM = Muddy sand, M = Mud

B.2.3 Vung Tau port project

In 1974 a port study was performed for the United States Agency for International Development Mission

to Vietnam (Lyon Associates inc. Consulting Engineers, 1974). In this report also the results of borings in

the estuary north of Vung Tau are given.

The results of the 12 borings are summarised in Figure B-9. Because the results were badly readable the

results have been reproduced. The locations of these borings are given in Figure B-10. From these borings

appears that the surface layer mainly consists of weak marine clay layers, sometimes mixed with sand or

clayey sand layers.

The marine clay located close to the gully is in general weaker (N-values in the order of 0-5) than the clay

more to the east (N-value 0-24).


Key

SP – Poorly graded sand, little or

no fines

SC – Light gray clayey sands,

wet & loose, low plasticity

CH – Grey marine clays, wet &

soft to stiff, high plasticity

LLW = 0m

-4

-8

-12

-16

-20

-24

-28

-32

12

0

17

16

16

15

11

0

0

16

18

24

18

1

0

0

1

12

2

0

0

0

3

3

1

?

?

?

4

0

0

7

13

5

0

0

0

0

0

0

1

2

0

0

3

6 7

2

3

4

?

2

8

0

12

?

?

2

9

0

7

3

4

2

3

0

10

0

0

7

8

15

13

Figure B-9: Vung Tau boring profiles with SPT blow-values

Figure B-10: Locations of the Vung Tau Port Study borings

Known soil parameters are given in Table B.4.


Table B.4: Soil parameters Vung Tau borings

Category Parameter Symbol Unit CH(*) CH SP SC

Description

[-]

Soft to very soft grey marine clays

Stiff clay Poorly graded fine sands

Sandy clays

SPT blow count

N [#] 0 - 1 9 - 41 7 - 30 18 - 105

Unit weights Saturated γw [kN/m

3]

16.4 - 19.3 17.7 - 20.9 17.7 - 21.7

Dry γd [kN/m3]

Atterberg limits

Liquid limit LL [%] 75 - 107 Plastic limit PL [%]

Plasticity index

PI [%] 43 - 70

Unknown tests (presumably undrained parameters)

Cohesion c [kPa] 7.35 – 19.6 (design 9.8)

35.80 - 89.40

0 13.40 - 26.80

Friction angle ϕ [deg]

0 32 - 38 29 - 36

Consolidation coefficient

cv [m2 /day] 2.8∙10

-3

Compression index

Cc 0.83 - 1.05

Moisture content

[%] 89 - 129

B.3 Comparison of available information

B.3.1 Choice of profiles

By putting all these sources of information together soil profiles for the dam can be determined.

However, several sources contradict. For example, according to the Vung Tau Port Study the top layer

north of Vung Tau consists of marine clays and clayey sands, whereas Figure B-8 expects sand with

pebbles as well. Also all sources give different profiles with different soil descriptions and names.

Therefore several profiles will be worked with:

1. One based on the weakest profile observed in all the information (this is a boring in Vung Tau).

The Holocene boundary is assumed on -25 m and below this boundary Pleistocene Thu Bo barrier

layering and parameters are assumed. (Vung Tau Weak)

2. One based on the strongest profile observed at Vung Tau. (Vung Tau Strong)

3. One based on the Go Cong boring and Thu Bo barrier soil parameters. (Go Cong)

The following assumptions are made:

For the Vung Tau profiles: the Holocene boundary is assumed on -25 m below MSL, based on an

average derived from the seismic profiles

For all profiles: unknown parameters are estimated using the Dutch national geotechnical code,

NEN 6740, based on the soil description from data. This is done by comparing the descriptions of

the layers, comparing parameters and comparing known SPT-values with CPT-values in the Dutch

codes.

To correlate the SPT data the following graphs and tables are used.


Table B.5: SPT correlations according to Peck & Hanson (1953)

Sources of these parameters are indicated with colours:

Red: Vung Tau port project (Lyon Associates inc. Consulting Engineers, 1974)

Blue: Thu Bo barrier project (Soil investigation Thu Bo barrier, Book 2: Appendix experimental

results (in Vietnamese), 2010), (Soil investigation Thu Bo barrier, Book 1: (in Vietnamese), 2010)

Green: NEN 6740, table 1. (Dutch Normalisation Institure NEN, 2006)

Known parameters, known soil descriptions and an estimation of the cone resistance , based

on SPT N-values, are used to find the corresponding soil type in the NEN. The chosen soil type are

given in the tables.

Orange: Cannot be found in the code or other sources of information, so these are estimations

based on own judgement and experience.

B.3.2 Profile 1: Weak Vung Tau

Table B.6: Assumed layers and parameters for the Weak Vung Tau profile

Category Parameter Symbol

Unit Layer 1 Layer 2

Description in the source Soft marine clay Sand, fine to medium, loose

to medium compacted Geological era Holocene Pleistocene

Geometry Depth top of layer below mean sea level

d [m] 0 to -12 -25

Thickness h [m] 13 - 25 27

SPT blow count N [#] 0 20

(Range 11 – 29)

Consistency based on SPT according to Table B.5 Very weak Medium

Soil type assumed to use Dutch code for unknown parameters Clay, clean, soft Sand, clean, medium

Unit weights Saturated γw [kN/m3] 17.85 20 Dry γd [kN/m3] 18

Undrained Cohesion cu [kPa] 9,8 - Friction angle φ [deg] 0 -

Effective Cohesion c' [kPa] 0 0 Friction angle φ’ [deg] 17.5 32.5

Consolidation coefficient cv [m2/s] 3.2∙10-8 1.0∙10-3 Bjerrum settlement parameters

Compression index Cc [-] 0.94 0.023

Creep index Cα [-] 0.0131 0

Permeability k [cm/s] 5.0∙10-7 5.8∙10-3 Void ratio e0 [-] 2.0 0.633

Name given in report Soft marine clay Fine to medium sand, medium compacted


B.3.3 Profile 2: Strong Vung Tau

Table B.7: Assumed layers and parameters for the Strong Vung Tau profile

Category Parameter Symbol Unit Layer 1 Layer 2 Layer 3 Layer 4

Description from source Soft marine

clay Clayey sand

Soft marine clay

Sand, fine to medium, loose to medium

compacted Geological era Holocene Holocene Holocene Pleistocene


d [m] 0 to -8 -8 -10 -25

Thickness h [m] 0 - 8 2 15 27

SPT blow count N [#] 0 15 15 20

(Range 11 – 29)

Consistency based on SPT according to Table B.5 Very weak Medium Stiff Medium

Soil type assumed to use Dutch code for unknown parameters Clay, clean,

soft

Sand, very

silty, clayey

Clay, clean,

stiff

Sand, clean,

medium

Unit weights Saturated γw [kN/m3] 17.85 20 19 20 Dry γd [kN/m3] 18 19 18

Undrained Cohesion cu [kPa] 9.8 -

50 (Range 35.80

- 89.40) -

Friction angle φ [deg] 0 - 0 -

Effective Cohesion c' [kPa] 0 0 13 0 Friction angle φ’ [deg] 17.5 25 20 32.5

Consolidation coefficient cv [m2/s] 3.2∙10-8 1.0∙10-3 1.0∙10-8 1.0∙10-3

Bjerrum settlement parameters

Compression index Cc [-] 0.94 0.015 0.16 0.023

Cc / (1+e0)

0.008 0.080 0.038

Creep index Cα [-] 0.0131 0 0.0035 0

Permeability k [cm/s] 5.0∙10-7 5.0∙10-4 5.0∙10-7 5.8∙10-3 Void ratio e0 [-] 2.0 0.7 1.0 0.633

Name given in report Soft marine

clay Sand, very silty, clayey

Moderate marine clay

Fine to medium

sand, medium

compacted

B.3.4 Profile 3: Go Cong (based on Thu Bo)

Table B.8: Assumed layers and parameters for the Go Cong profile


Description from source Mud Sand, muddy Clay, silty

Sand, fine to medium, loose to medium

compacted


d [m] -3 -12 -15 -18

Thickness h [m] 9 3 3 12

SPT blow count N [#] < 2 - < 10-19 11-29

Consistency based on SPT according to Table B.5 Very soft - Stiff Medium

Soil type assumed to use Dutch code for unknown parameters Clay, clean,

sift

Sand, very

silty, clayey

Clay, clean,

stiff

Sand, clean,

medium

Unit weights Saturated γw [kN/m3] 14.6 20 19 19.2 Dry γd [kN/m3] 14.6 18 19 15.8

Undrained Cohesion cu [kPa] 11.8 - 33.2 -



Friction angle φu [deg] 1.5 - 2.0 -

Effective Cohesion c' [kPa] 8.8 0 25.8 5 Friction angle φ’ [deg] 23 25 22.8 28.6

Consolidation coefficient cv [m2/s] 4.5∙10-9 1.0∙10-4 3.7∙10-9 1.0∙10-3

Bjerrum settlement parameters

Compression index Cc [-] 1.01 0.014 0.10 0.023

Cc / (1+e0)

0.009 0.038

Creep index Cα [-] 0.013 0 0.004 0

Permeability k [cm/s] 7.6∙10-6 5.8∙10-4 4.6∙10-7 5.8∙10-3 Void ratio e0 [-] 1.981 0.6 0.725 0.633

Name given in report Mud Sand, very silty

Clay, silty

Fine to medium

sand, medium

compacted

Project Flood Defence HCMC | APPENDIX Calculation of the crest level Page 45 / 140

CALCULATION OF THE CREST LEVEL C

C.1 Storm surge: Calculation for blocked profile

The formula for the total storm surge in an area with a constant depth:

(C.1)

The wind set-up at the near shore end of the deeper block profile (section 1) is half of the total storm

surge:

(C.2)

The total storm surge at the shore, or in this case the dam, is the wind set-up of section 1 plus the total

storm surge of the shallow zone.

*

+

(C.3)

Profile 1

3.5∙10-6

- constant 42 m/s wind velocity 10 m/s

2 gravitational acceleration

20 m water depth section 1 (constant) 5 m water depth section 2 (constant) 87.5∙10

3 m fetch length section 1

12.5∙103 m fetch length section 2

2.1 m total wind set-up

Profile 2

3.5∙10-6



20 m water depth section 1 (constant) 5 m water depth section 2 (constant) 70∙10


30∙103 m fetch length section 2


C.2 Storm surge: Calculation for sloping profile

The formula for calculating the storm surge for a depth that varies with distance to the shore:

∫

( )

(C.4)

The wind set-up of the first section is calculated the same way as for the blocked section (constant depth

section 1):

(C.5)

Project Flood Defence HCMC | APPENDIX Calculation of the crest level Page 46 / 140

The wind setup values that follow from these storm surges are half of the total setup height:

(C.6)

The depth functions of the profiles will be in this form (linear shore profile):

( ) (C.7)

The integral of this depth function will be in this form:

∫

( )

[

( )]

(C.8)

The total storm surge in the second section will then become:

[

( )]

(C.9)

The total setup at the dam will then be the sum of the wind set-up in section 1 and the storm surge in

section 2:

(C.10)

Profile 1

3.5∙10-6



20 m water depth section 1 (constant) 81∙10


101 - constant of depth function 0.001 - constant of depth function 19∙10



Profile 2

3.5∙10-6



20 m water depth section 1 (constant) 52.5∙10


41 - constant of depth function 0.0004 - constant of depth function 47.5∙10



Project Flood Defence HCMC | APPENDIX SwanOne calculation of wave height Page 47 / 140

SWANONE CALCULATION OF WAVE HEIGHT D

SwanOne is a computer program that determines wave properties for near shore locations. This

programme was used to determine the wave height ( ) and period ( ) at the toe of the structure.

This appendix describes the method and the conclusions that were drawn from the SwanOne calculation.

D.1 Input data

D.1.1 Wind and wave data

Input in SwanOne consists of a bottom profile, wave conditions at an offshore boundary and wind speed

data. Wave conditions are described in the Vietnamese design code and for a return period of 200 years

are = 10.03 m and = 12.0 s between Vung Tau and the Dinh An estuary. The location of the offshore

boundary is about 110 km from the location of the dam and the mean wave direction is 130° with respect

to the north (Figure D-1 and Figure D-2). For the VT-GC project currents are not included as their effect is

usually small and this is a preliminary design stage.

Figure D-1: Diagram of 4 zones of near shore wave computation; from Ba Ria – Vung Tau to Kien Giang (Vietnamese

Design Code)


Figure D-2: Mean wave direction with respect to north (SwanOne)

The water level that was used for these calculations is MSL +3.2 m. Information on what this design Still

Water Level is based on can be found in the main report, section 4.1.2.

Wave setup is a rise in the mean water level above the still-water elevation of the sea due to onshore

mass transport of the water by wave action alone. This factor is also taken into account in SwanOne.

A conservative value for the wind speed of 42 m/s is assumed in the same direction as the waves. During a

typhoon wind can come from all possible directions, therefore this unfavourable situation of the wind and

waves having the same direction is considered. The return period of 1/200 per year is relatively high for

this situation. For comparison a situation without wind has been plotted as well.

D.1.2 Bottom profiles

Three different bottom profiles have been considered: one for an average depth that holds for the largest

stretch of the dam; one for the shallow part near Go Cong and one that holds for the area near the

navigation channel, where the depths are largest. In Figure D-3 to Figure D-8 the locations of the depth

profiles and Excel plots of these profiles are shown. For the thick part of the red lines in Figure D-3, Figure

D-5 and Figure D-7 results for and as a function of the distance to the offshore boundary are

given in the section D.2 of this appendix. The depth profiles were made using the nautical map (Myres,

1993). A depth of 26 m below chart datum was assumed near the offshore boundary.


Figure D-3: Location average cross section

Figure D-4: Bottom profile average cross section


Figure D-5: Location shallow cross section

Figure D-6: Bottom profile shallow cross section


Figure D-7: Location deep cross section

Figure D-8: Bottom profile deep cross section


D.2 Results

For the last 7 km of each profile , and the water depth above the sea bottom are plotted. In

Figure D-9 the results for the average profile with a wind speed of 42 m/s are given.

Figure D-9: Results SwanOne calculation location with average depth

For this average profile a design significant wave height ( ) of around 3.5 m was found, with a design

wave period ( ) of 7 to 7.5 seconds. When the water depth increases the wave height and period

increase rapidly.


Figure D-10: Results SwanOne calculation location with average depth, no wind

In Figure D-10 can be seen that winds increase the significant wave height with approximately 0.5 m. The

wave period increases when no wind is taken into account.

For calculations = 3.2 m and = 8.0 s are used. In case of a storm, when the water level is as

high as the design still water level it is assumed that there are winds blowing over the area. A wave height

in between the two cases (wind of 42 m/s in the direction of the waves and no wind) will be used for

further calculations.

A wave surge level of 0.25 m was found.


Figure D-11: Results SwanOne calculation more shallow location

For the shallow part and are lower than for the part of the dam with an average depth. A

significant wave height of 3 m and a period of about 8 s may occur for a 200 year return period (Figure

D-11). A wind speed of 42 m/s was used for these calculations. These values are considered conservative.


Figure D-12: Results SwanOne calculation deeper location

For the deep part is higher than for the part with an average depth. A significant wave height of 4.5

m may occur for a 200 year return period (Figure D-12). The wave period is nearly the same, around 7.5 s.

A wind speed of 42 m/s was used for these calculations. These values are considered conservative.


D.3 Conclusions and recommendations

D.3.1 Conclusion

For the largest part of the dam a design significant wave height of 3.2 m and a design period

of 8 s were determined for calculations. For the deeper cross sections the significant wave height

increases, for more shallow parts it reduces. The wave height for the entire dam has a value between 3

and 4.5 m. The design wave period for the dam has values between 7 and 8.5 s.

A wave surge level, this is a water level elevation due to wave action, of 0.25 m was found for the cross

section located on average depth.

D.3.2 Recommendations

As can be seen in section D.2 wind effects play a role in determining the significant wave height. The wind

and its direction that occur in a 200 year return period storm can be determined more accurately from

wind and wave data measurements.

Currents are not included in this determination of the significant wave height. A current speed of 1.5 m/s

was recommended by Nguyen Huu Nhan PhD, the Vice Director of the Institute of Coastal and Offshore

Engineering (ICOE).

Depth profiles were constructed using a nautical map of the area. This map only covers the area until 30

km offshore of the VT-GC dam. A more detailed profile could be constructed with data on the water depth

further offshore.

The design still water level doesn’t hold for the entire cross section. During a storm, the water level

increases towards the coast with a maximum value of the design still water level that was used as input.

Calculated values are considered conservative.

SwanOne was not developed for construction purposes. Deficiencies or other limitations in the program

may result in different values than the ones made in reality. The values that were calculated with

SwanOne should be verified, for instance with measurement data.

Project Flood Defence HCMC | APPENDIX Calculation of the 2% wave run-up Page 57 / 140

CALCULATION OF THE 2% WAVE RUN-UP E

With use of Excel a deterministic calculation was made for the 2% wave run-up. In this appendix the

calculation method and results are described.

E.1 Wave run-up formula

For a deterministic calculation the following formula is recommended:

(E.1)

With a maximum of:

(

√

) (E.2)

And:

( )

√ ⁄ (E.3)

Where:

(E.4)

(E.5)

In Table E.1, the values that were used for the calculation of the 2% wave run-up can be found.

Calculations were made for different slope angles of the outer slope and different berm lengths.

Table E.1: Calculation values

Parameter Unit Explanation Calculation value

m Wave run-up height exceeded by 2% of the incoming waves

See table

m Incident significant wave height at the toe of the structure

3.2

- Influence factor for a berm Depends on berm length

- Influence factor for roughness elements on a slope

1 or 0.65 dependent on revetment of the outer slope

- Influence factor for oblique wave attack 1 - Breaker parameter (Iribarren parameter) Dependent on slope m Deep water wave length 102

rad Seaward slope steepness 1:3 1:4 1:5 or 1:6 m/s

2 Gravitational acceleration 10

s Calculated wave period that follows from the wave spectrum

8

s Peak wave period 8.9

E.2 Influence factors

For a berm, slope roughness, oblique wave attack and certain influence factors are used in the formula for

the wave run-up. In this section it is described how these influence factors are obtained.


E.2.1 Roughness

The influence factor for roughness ( ) takes into account the roughness and permeability of the surface.

In Table E.2 roughness factors for different revetment materials are given.

Table E.2: Surface roughness factors for typical elements for coastal dikes and embankment seawalls

Revetment type

Concrete 1.0 Asphalt 1.0 Closed concrete blocks 1.0 Grass 1.0 Block mats 0.95 Basalt 0.90 ¼ of stone setting 10 cm higher 0.90 Small blocks over 1/25 of surface 0.85 Small blocks over 1/9 of surface 0.80 Ribs (optimum dimensions) 0.75 Quarry Stone 0.65

Roughness elements have no or little effect below 0.25·Ru2%,smooth below the still water line and above

0.50·Ru2%,smooth above the still water line.

For the VT-GC dam calculations were made with an influence factor of 0.65 (quarry stone revetment) and

1.0 (asphalt revetment). These materials were chosen as rocks are available in the area and asphalt is a

flexible revetment that can deal with unequal settlements.

E.2.2 Berm

The influence factor for a berm is calculated as follows:

( ) (E.6)

For:

(E.7)

Where:

(E.8)

And for a berm above still water line:

(

) (E.9)

For a berm below still water line:

(

) (E.10)

- Influence factor for a berm - Relative berm length Depends on and

- Relative distance between berm and design still water level

0.034

m Length of the berm Different lengths

- Length of interruption of the slope (see Figure E-1)

Depends on slope angle

m Distance between berm and design still water 0.75 (Average between sea level rise and


level settlements)

m Wave run-up height exceeded by 2% of the incoming waves

-

m Incident significant wave height at the toe of the structure

3.2

Figure E-1: Definition of Lberm (EurOtop, 2007)

It can be seen from these formulas that a berm located on the still water line has the most favourable

influence on the wave run-up. Because of settlements and sea level rise the berm cannot be located on

the still water line during the entire lifetime of the dam. For the design of the barrier it is assumed that

the berm will be constructed above design SWL and that due to settlements and sea level rise the berm

will be located below design SWL at the end of the lifetime. For calculations a berm location of 0.75 m

below design SWL is assumed. Settlements below the berm and sea level rise are 1 and 0.75 m

respectively. It is assumed that the berm will be constructed above design SWL and will be located below

design SWL at the end of the lifetime. A berm location of 0.75 m below design SWL is used for

calculations.

Other factors then the location below design SWL that determine the influence factor for a berm are: the

length of the berm the Ru2% wave run-up and the interrupted slope length, which depends on the slope

angle and the significant wave height Hs (Figure E-1).

E.2.3 Oblique wave attack

The influence factor for oblique wave attack is calculated as follows:

| | (E.11)

For:

For wave overtopping:

| | (E.12)

For:


Figure E-2: Definition of angle of wave attack β (EurOtop, 2007)

The angle of attack between the significant waves and the VT-GC dam will be between 0 and 30°.

Therefore the factor for oblique wave attack has a value between 1 and 0.94 for calculating the run-up.

For the design of the VT-GC dam a value of 1 is used.

E.3 Calculation results

Using the formulas that were presented in the previous section, the 2% wave run-up was calculated for

different slope angles and berm lengths. The results are shown in

Table E.3 and Table E.4. The first table is the run-up height for an asphalt revetment on the outer slope.

The second one for a quarry stone revetment. It can be seen that a gentler slope and a longer berm

reduce the wave run-up.

Table E.3: Ru2% (in m above SWL) for different berm lengths and slope angles for asphalt revetment

B [m] \ Slope 1:3 1:4 1:5 1:6

0 10.0 7.9 6.3 5.3 2 9.5 7.3 5.9 5.0 4 8.7 6.8 5.6 4.8 6 8.0 6.4 5.3 4.5 8 7.4 6.0 5.0 4.3 10 6.9 5.7 4.8 4.2 12 6.5 5.4 4.6 4.0 14 6.3 5.1 4.4 3.8 16 6.3 4.8 4.2 3.7 18 6.3 4.7 4.0 3.6 20 6.3 4.7 3.9 3.4

Table E.4: Ru2% (in m above SWL) for different berm lengths and slope angles for quarry stone revetment

B [m] \ Slope 1:3 1:4 1:5 1:6

0 6.5 5.1 4.1 3.4 2 6.2 4.8 3.9 3.2 4 5.6 4.4 3.6 3.1 6 5.2 4.1 3.4 3.0 8 4.8 3.9 3.3 2.8 10 4.5 3.7 3.1 2.7 12 4.2 3.5 3.0 2.6 14 4.1 3.3 2.8 2.5 16 4.1 3.1 2.7 2.4 18 4.1 3.1 2.6 2.3 20 4.1 3.1 2.5 2.2


E.4 Material use

Because using more material will result in higher costs, the content of the dam per m1 length was

calculated for different slopes and berm widths. From these figures an optimal berm length can be

estimated for different slope angles.

For this calculation a crest width of 7 m is assumed and an inner slope with an angle . The

layout of the inner slope has not been determined yet. Future design probably includes an inner berm.

The difference in material use for the inner berm for different dam designs is not taken into account, but

is assumed to be relatively small. The seabed is assumed to be located on MSL -5 m (this bed level holds

for the largest stretch of the dam between the two navigation channels) and the design SWL on MSL +3.2

m. The berm is assumed to be located SWL -0.75 m. In Figure E-3 the cross section that is used for volume

calculation is drawn.

SWL

α

α

B 7

1:3

Ru 2% level

MSL +3.2m

MSL

MSL -5m

SWL -0.75m

Figure E-3: Cross section, used to calculate the volume per m dam

Results of the calculations are shown in Table E.5 and Table E.6. For each slope angle the lowest volume is

highlighted in yellow. It can be seen that for a gentler slope more material is required. Except for the

situation of a 1:3 slope with asphalt revetment. The optimum berm width when using quarry stone

revetment, is shorter for a lower slope angle. For the asphalt slope the results are spread.

Table E.5: Content in m3 per m

1 dike for different berm length and slope angle for asphalt revetment

B [m] \ Slope 1:3 1:4 1:5 1:6

0 1124 1020 945 910 2 1082 967 915 893 4 1006 926 891 879 6 949 894 873 869 8 905 870 859 861 10 871 851 848 856 12 846 836 841 853 14 849 826 835 851 16 865 818 832 852 18 882 824 831 853 20 898 841 832 856

Table E.6: Content in m3 per m

1 dike for different berm length and slope angle for quarry stone revetment

B [m] \ Slope 1:3 1:4 1:5 1:6

0 753 716 692 689 2 738 694 683 686 4 705 679 676 686 6 681 669 673 687


B [m] \ Slope 1:3 1:4 1:5 1:6

8 664 662 672 689 10 653 659 673 693 12 646 658 676 698 14 655 659 680 705 16 672 661 685 712 18 688 672 692 719 20 705 688 699 728

For an asphalt slope a slope angle of 1:4 gives the lowest volume per m1 of the dam. The optimum berm

length is 16 m. For further design these values for the slope and berm length are used.

For a slope with quarry stone revetment a slope of 1:3 gives the lowest volume per m1. Because this steep

slope angle has a big chance of being instable and because steeper slopes result in larger stone sizes for

the revetment, a slope of 1:4 is chosen for further design. As can be seen in Table E.6, the content of the

dam per m1 is barely differs for a berm of 12, 14 or 16 m length. In the previous section is concluded that

a longer berm results in a lower dike. A lower dike results in lower settlements in the subsoil, as the load

will be lower. Therefore a berm length of 16 m is chosen for the quarry stone slope.

Project Flood Defence HCMC | APPENDIX Calculation of overtopping Page 63 / 140

CALCULATION OF OVERTOPPING F

In this appendix the method to calculate the overtopping discharge for a certain freeboard height is

described. For an overtopping discharge of 0.1, 0.5, 10 and 50 l/m/s the required freeboard height is

calculated.

F.1 Deterministic calculation method

The principal equation used for overtopping is:

(

)√

(F.1)

With:

√ ( ) (F.2)

(F.3)

( )

√ ⁄

(F.4)

(F.5)

(F.6)

The maximum overtopping discharge is:

(

)√

(F.7)

This formula is valid for . Other formulas are recommended in case of a shallow foreshore. The

wave steepness ( ) can vary between 0.04 (steep storm waves) and 0.01 (long waves due

to swell or wave breaking).

Parameters are:

Parameter Unit Explanation Calculation value

m3/s/m Overtopping discharge 0.1; 0.5; 10 and 50 l/m/s

m Wave height at the toe of the structure 3.2 m Crest height above design still water level Depends on discharge - Influence factor for a berm 0.61 (see appendix E)

- Influence factor for roughness elements on a slope

1 or 0.65 (see appendix E)

- Influence factor for oblique wave attack 1 (see appendix E)

- Influence factor for vertical wall on top of the crest

1

- Breaker parameter (Iribarren parameter) 1.41

rad Seaward slope steepness 0.24 m/s

2 Gravitational acceleration 10

s Peak wave period 8.9

The influence factor for a wall is 1.0. It is considered that there is no wall constructed on the dike crest.


F.2 Calculation results

Using the calculation method that is described in the previous section, the necessary freeboard height is

calculated for a discharge volume of 0.1, 0.5, 10 and 50 l/m/s. The freeboard height is the elevation of the

crest above design Still Water Level (Figure F-1). This calculation is done for a 1:4 slope and a berm length

of 16 m (as calculated in appendix E), for a revetment with asphalt and a revetment with quarry stone.

Figure F-1: Definition of some parameters for the calculation of overtopping (EurOtop, 2007)

Results are displayed in Figure F-2.

Figure F-2: Required freeboard for different overtopping discharge and different revetment type

From Figure F-2 can be seen that allowing a larger overtopping discharge can reduce the necessary

freeboard height significantly. The required freeboard height for quarry stone reduces with almost 50%


compared to the 2% run-up level when an overtopping discharge of 50 l/m/s is allowed. For an asphalt

revetment a reduction of even more than 50% is found.

For small values of the overtopping discharge the calculation is not very accurate. For an overtopping

discharge of 0.1 and 0.5 l/m/s the same freeboard heights were calculated.

For the design of the dam a discharge of 50 l/m/s will be acceptable if the revetment on the inner slope is

designed to withstand this discharge and if it is possible to close the road over the dam for traffic when

much overtopping is expected, e.g. when a storm is predicted. Closing the road for part of the time during

storm conditions is considered worthwhile as the design crest height reduces significantly.

For a dam with an outer slope revetment with asphalt this results in a freeboard height of 2.4 m. For

quarry stone the required freeboard height is 1.6 m.

Project Flood Defence HCMC | APPENDIX Sea level rise Page 66 / 140

SEA LEVEL RISE G

G.1 Introduction

Over the last decades, climate change has been investigated intensively. Climate change has an impact on

many things on earth, and sea level rise is one of them. In the design of the dam, which have to protect

the hinterland from flooding, the future water height and the therefore the change in mean sea level has

to be taken in account.

In the fourth assessment report of the IPCC (Intergovernmental Panel on Climate Change) the definition

of sea level change is defined as follows (IPCC, 2007):

“Sea level can change, both globally and locally, due to (i) changes in the shape of the ocean basins, (ii)

changes in the total mass of water and (iii) changes in water density.”

With (iii) it meant that water will expand when it warms up.

G.2 History

For a long time research is done on the sea level height. Since 1950 tidal gauge measurements of the sea

level are done and in 1993 (more accurate) satellite measurements started. Together with the

reconstruction of sea level height in the period 1870 - 1950, Figure G-1 with the annual averages of the

global mean sea level could be constructed. It can be seen that there is an accelerated sea level rise since

the 30’s of the 20th

century. The reason for this increase will not be discussed here. The IPCC assesses the

rate for global sea level rise as 1.8 (±0.5) mm/year for the period 1961 - 2003 and 3.1 mm/year for the

period 1993 - 2003.

Figure G-1: Annual averages of the global mean sea level (mm). The red curve shows reconstructed sea level since

1870; the blue curve shows coastal tide gauge measurements since 1950; the black curve shows the satellite

measurements since 1993. (IPCC, 2007)

G.3 Causes of the sea level rise

There are different sources that contribute to the sea level rise. First there is the thermal expansion: the

sea level will rise if the ocean warms, because the density of the water changes. Second there is the

contribution to melting glaciers and ice caps. The Greenland and Antarctic ice sheets are expected to

contribute significantly to the rise of the sea level, therefore they are treated separately. The expected


contributions are listed in Table G.1. From the table it can be seen that still a part of the contribution to

sea level rise is unknown.

Table G.1: Overview sea level rise (mm/year) due to different causes (IPCC, 2007)

Source Sea level rise [mm/year] 1961-2003

1993-2003

Thermal expansion 0.42 ± 0.12 1.6 ± 0.5 Glaciers and ice caps 0.50 ± 0.18 0.77 ± 0.22 Greenland ice sheet 0.05 ± 0.12 0.21 ± 0.07 Antarctic Ice Sheet 0.14 ± 0.41 0.21 ± 0.35 Sum 1.1 ± 0.5 2.8 ± 0.7 Observed 1.8 ± 0.5 3.1 ± 0.7 Difference 0.7 ± 0.7 0.3 ± 1.0

G.4 Global sea level rise

With (a part of) the contributors to sea level rise known, investigation is done on the different

contributions. Also different scenarios are composed. Together, predictions of the sea level (due to

climate change) can be made. Table G.2 shows the expected sea level rise at different world-development

scenarios. In scenarios A1FI and A2 high emission are expected, in A1B and B2 medium emissions and in

A1T and B1 low emissions. A further explanation of the scenarios can be found in Figure G-3.

Table G.2: Possible sea level rise for different scenario's (IPCC, 2007)

The IPCCC expects a global sea level rise of 0.18 – 0.59 m in 2090 – 2099, compared to the level in 2980 –

1999, see Figure G-2. The different scenarios give a wide bandwidth.


Figure G-2: Projection of the sea level rise, measured and projections (IPCC, 2007)

Scenario A1 = a future world of very rapid economic growth, global population that peaks in mid-

century and declines thereafter, and the rapid introduction of new and more efficient

technologies. Major underlying themes are convergence among regions, capacity building and

increased cultural and social interactions, with a substantial reduction in regional differences in

per capita income. The A1 scenario family develops into three groups that describe alternative

directions of technological change in the energy system. The three A1 groups are distinguished

by their technological emphasis: high emission scenario fossil-intensive (A1FI), medium

emission scenario non-fossil energy sources (A1T) or a low emission scenario: a balance across

all sources (A1B) (where balanced is defined as not relying too heavily on one particular energy

source, on the assumption that similar improvement rates apply to all energy supply and end

use technologies).

Scenario A2 = a very heterogeneous world. The underlying theme is self-reliance and preservation

of local identities. Fertility patterns across regions converge very slowly, which results in

continuously increasing population. Economic development is primarily regionally oriented and

per capita economic growth and technological change more fragmented and slower than other

storylines. This is a high emission scenario.

Scenario B1 = a convergent world with the same global population, that peaks in mid-century and

declines thereafter, as in the A1 storyline, but with rapid change in economic structures toward

a service and information economy, with reductions in material intensity and the introduction

of clean and resource-efficient technologies. The emphasis is on global solutions to economic,

social and environmental sustainability, including improved equity, but without additional

climate initiatives. This is a low emission scenario.

Scenario B2 = a world in which the emphasis is on local solutions to economic, social and

environmental sustainability. It is a world with continuously increasing global population, at a

rate lower than A2, intermediate levels of economic development, and less rapid and more

diverse technological change than in the B1 and A1 storylines. While the scenario is also

oriented towards environmental protection and social equity, it focuses on local and regional

levels. This is a medium emission scenario.

The SRES scenarios do not include additional climate initiatives, which means that no scenarios are included that explicitly assume implementation of the United Nations Framework Convention on Climate Change or the emissions targets of the Kyoto Protocol.

Figure G-3: Emission Scenarios of the IPCC Special report on emission scenarios (SRES) (Ministry of Natural

Resources and Environment, 2009)

Other studies predict different (higher) global sea level rises. The difference could be contributed to the

fact the IPCC does not take into account one of the major processes contributing to sea level rise: the fact

that the melting process of ice is not linear (National Research Council US, 2001).


By the National Research Council of the National Academies in the US a literature study is done on sea

level rise. In Table G.3 the overview can be observed.

Table G.3: Predicted sea level rise (National Research Council US, 2001)

Author Special Sea level rise (basic year is 1990)

Horton et al. (2008) Incorporating the empirical effect into models

0.62 to 0.88 meters in 2100

Rohling et al. (2008) Using paleoclimatic evidence from past interglacial periods

1.6 m per century

Kopp et al. (2009) Looking to interglacial period 0.56 to 0.92 meters per century

Pfeffer et al. (2008) Using geophysical constraints of ice loss

2 meter in 2100

G.5 Local sea level rise and Ho Chi Minh City

For the design of the VT-GC dam, a local expected sea level rise has to be determined, because the sea

level will not rise with the same amount everywhere. Regionally there can be differences due to the

different thermal expansion and the exchange of water with other sources (e.g. rivers, reservoirs and ice

caps).

From the IPCC follows that for the location of the VT-GC dam the local sea level rise will be around 0.025

m higher than the global rise (see Figure G-4). Following from that: the by the IPCC expected sea level in

2090-2999 will be 0.20 – 0.61 m higher than in 1980 - 1999.

Figure G-4: Local sea level change (m) due to ocean density and circulation change, relative to the global average

(IPCC, 2007)

Finally, the Vietnamese Ministry of Natural Resources and Environment (MoNRE) (Ministry of Natural

Resources and Environment, 2009) did research on how the climate change will vary in Vietnam. They did


the study based on the same scenarios of the IPCC report, but with local data. The predictions are given in

Table G.4.

Table G.4: Sea level rise in Vietnam in cm, relative to 1980-1999. (Ministry of Natural Resources and Environment,

2009)

The data for a medium emission scenario is used for the design of most hydraulic structures in Vietnam.

G.6 Conclusions

Global sea level rise predictions vary between 0.18 and 1.00 m sea level rise in 2100. Further research is

needed to make more accurate predictions.

For the design of the VT-GC dam the data in Table G.4 for a medium emission scenario will be used.

Therefore a sea level rise of 75 cm in 2100 is assumed.

Project Flood Defence HCMC | APPENDIX Bearing capacity Page 71 / 140

BEARING CAPACITY H

Bearing capacity is one of the important aspects in the foundation engineering: will the soil fail or will it

keep its properties? First is looked whether the bearing capacity is large enough, when the dam is built

directly on the soil. This will be schematised as a shallow strip foundation, and to calculate the bearing

capacity the formula from Netherlands Eurocode 7 (EC7-NL), (Geotechniques Normcommission NEN,

2008) and (Geotechniques Normcommission, NEN, 2007) based on the formula of Brinch-Hansen, is used.

The load induced by the dam is therefore simplified to a uniform load.

The bearing capacity check has to be done for the undrained as well the drained situation. Undrained is

needed directly after placing the embankment, when the increase in stress is mainly carried by the water.

After a while, the excess water pressure will be dissipated, so a drained calculation has to be carried out.

All calculations will be done for a cross section of the dam with 1 m width.

H.1 Assumed parameters

The assumed soil layering and parameters can be found in Appendix B. In short (see Figure H-1):

0 m MSL

-10

-20

-30

-40

Fine to medium sand, medium

compacted

Mud

Soft clay

Stiff clay

Sand, very silty, clayey

Vung Tau

Weak

Vung Tau

Strong

Go Cong with

Thu Bo

parameters

Key

Figure H-1: Soil profiles

The mean depth of the dam is assumed to be 5 meter below Mean Sea Level. Also situations where the

bottom is located on MSL -10 m and MSL -2 m are taken into account. Safety factors are taken from NEN

6740, table 3:

(safety factor for soil) (safety facor for internal friction)

(safety factor for cohesion) (safety factor for undrained shear

strength)

The cross section as given in Figure H-2 is assumed, consisting of only sand

( kN/m3 and kN/m

3). The cross section area and foundation width are given in Table

H.1.


+6.5 m MSL

+3.2 m MSL

+0.0 m MSL

-5.0 m MSL

713.2812.820 9.9 25 9.6 15

1:4

1:41:3

1:3

Figure H-2: Cross section of the dike

Table H.1: Parameters of the cross section

Depth dam [m –MSL] Area cross section [m2] Width of dam B’ [m]

0 299 85.5 -1 388 92.5 -2 484 99.5 -5 814 120.5 -10 1504 155.5

H.2 Calculation method ‘shallow foundation’ with

Eurocode 7

In the beginning of the 20th

century, Prandtl developed a method to calculate the failure load of a strip

foundation. This formula satisfied the horizontal and vertical equilibrium and the Mohr-Coulomb criterion

and divides the soil in 3 areas (Figure H-3). In all areas the stress state is assumed to be critical. In area 1

the horizontal stress is higher than the vertical, in area 3 it is vice versa and in area 2 it is somewhere

between. The basic formula to calculate the bearing capacity [kN/m] is (per meter length):

(H.1)

In which:

c'd design value of effective cohesion [kN/m]

design value of effective stress at lowest level of foundation [kN/m]

Nc, Nq dimensionless parameters [-], the formulas are:

(

) (H.2)

( ) ( ) (H.3)

Figure H-3: Strip foundation


By Keverling Buisman, Terzaghi, Brinch Hansen and others the formula is supplemented with other

possible characteristics of a strip foundation. The formula of Brinch-Hansen is as following:

(H.4)

In which:

c'gem,d design value of the mean effective cohesion [kN/m]

ic, iq, iγ inclination factors of the load [-]

sc, sq, sγ shape factors [-], with an rectangular shape, the formulas are:

(H.5)

(

) (H.6)

(

) (H.7)

Nγ dimensionless parameter [-], the formula is:

( ) ( ) (H.8)

B’ width of the effective strip foundation [m] (when the load is eccentric)

L’ length of the effective strip foundation [m] (when the load is eccentric)

γ’gem,d Design value of the effective weight of the soil [kN/m2 ], in which

The mean angle of internal friction, mean effective cohesion and mean effective unit weight have to be

used. To calculate these means, all the layers above the influence depth need to be taken into account.

Figure H-4: Shape factor, rectangular shape Figure H-5: Inclination of a load

When condidering the undrained situation, Nγ = 0 and Nq = 1 (because φ’ = 0) so the formula can be

simplyfied to:

( ) (H.9)

In which:

cu,d Design value of the undrained shear stress [kN/m2]

And: (

) (H.10)

In the undrained calculation, the weakest layer which lies in the depth of influence is the critical one.


H.2.1 Depth of influence

The depth of influence Ze depends on the width of the dam and the angle of internal friction and can be

determined from Figure H-7.

Figure H-6: Area of influence of a foundation

Figure H-7: Graphics to determine the area of influence

For a representative angle of internal friction of 17.5 , = 2,26 and = 1.08. In the undrained

calculations a depth of influence of 0.7*B’ is assumed ( ).

H.2.2 Squeezing

Squeezing is the effect that when a load is put on a non-cohesive layer which lies on a cohesive layer, the

cohesive layer is pushed away in lateral direction. So, in case that there is a cohesive layer within the

depth of influence squeezing has to be taken in account.

(

) (H.11)

In which:

σ'sq,d design value of the effective stresses on the upper side of the squeezing cohesive layer

[kN/m]

design value of effective stress at the upper side of the cohesive layer [kN/m]

hsq height of the cohesive layer [m]


H.2.3 Punch

Punch is the effect that when a load is applied on a cohesive layer which lies on top of a non-cohesive

layer, the non-cohesive layer is put aside. It can occur when the difference in the angle of internal friction

between two layers is more than 6 degrees. The effective width of the foundation has to be increased to

make a fictive effective width of the foundation above the layer which is punched. For this an angle of 8

degrees compared to the vertical has to be used, on both sides of the foundation. The calculation will

therefore be iterative.

H.3 Calculations without soil improvement

H.3.1 Hand calculation of the bearing capacity: example for the

Weak Vung Tau profile

The dam with a cross section of Figure H-2 is checked for the bearing capacity. It is simplified, without

iterating, squeezing and punch. The parameters are chosen as follows:

ic, iq, and iγ are 1 There is no inclination of the load

is 0 The foundation is on surface level

for undrained

sc and sq are 1 for this drained calculation

sγ = 0.998

3

B = 120.5 m

L = 2000 m

Undrained

( ) = 37.4 kN/m (this is per m width)

Drained

This hand calculation is done with rounded marks: with more decimals it is 3849 kN/m. The load of the

dam will be calculated using the area of the cross section.

The uniform load on the soil is calculated as follows:

( ) (H.12)


For at depth MSL -5, and the water at MSL -1 this load is equal to:

( )

The lowest water level due to the tide is MSL-2.25 m. This water level is very conservative, because of the

short duration of low water and the fact that on the other side of the dam the water will probably higher.

So, the water will stay at a higher level in the dam, therefore the dam will have has less effective weight

and a lower uniform-load. A water level of MSL -1 m is also low, but more representative and therefore

taken.

Table H.2: Hand calculation results bearing capacity and load, profile 1:Weak Vung Tau

Bottom at [MSL m]

Pd undrained Pd drained Fd Undrained check Drained check

-2 37.4 kN/m (3714 kN)

2513 kN/m (250050 kN)

87.8 kN/m (8736 kN)

Fail Pass

-5 37.4 kN/m (4498 kN)

3849 kN/m (463810 kN)

102.6 kN/m (12368 kN)

Fail Pass

-10 37.4 kN/m (5804 kN)

6482 kN/m (1007946 kN)

128.3 kN/m (19951 kN)

Fail Pass

In Table H.2 the calculated design values of bearing capacity and loads with different water depth are

given (as well the uniform load as the total load Pd = pd * B’). In the undrained situation, the bearing

capacity is not enough (Pd < Fd). The drained situation has enough capacity.

H.3.2 Computer calculation of the settlements: example Weak

Vung Tau profile

The computer program D-Foundations is used to check the hand calculation. This computer program is

based on the same formulas as the hand calculation form Netherlands Eurocode 7 (EC7-NL). The

foundation is schematized in the same way: a strip of 120.5 m width.

Figure H-8: Schematisation of the foundation in D-Foundations


This program also schematises the dam as a uniform load. The calculated values are given in Table H.3.

Squeeze is also calculated, but this doesn’t change the outcome of the checks. Punch doesn’t happen in

this situation. The conclusion is that the bearing capacity of the soil is not big enough and the soil will fail

when constructing the dam.

Table H.3: Computer calculation results bearing capacity and load, profile 1:Weak Vung Tau Undrained

Bottom at [MSL m]

Pd undrained Fd Rd squeeze Fpull [kN] Undrained check

Undrained squeeze check

-2 37.4 kN/m (3714 kN)

87.8 kN/m (8738 kN)

68.7 kN/m (6837 kN)

361 Fail Fail

-5 37.4 kN/m (4498 kN)

102.6 kN/m (12368 kN)

81.0 kN/m (9766 kN)

437 Fail Fail

-10 37.4 kN/m (5804 kN)

128.3 kN/m (19958 kN)

112.6 kN/m (17504 kN)

564 Fail Fail

Table H.4: Computer calculation results bearing capacity and load, profile Weak Vung Tau Drained

Bottom at [MSL m]

Pd drained Fd Rd punch Vd Punch Drained check

-2 3216 kN/m (320018 kN)

87.8 kN/m (8738 kN)

3216 kN/m (320018 kN)

87.8 kN/m (8738 kN)

Pass

-5 4872 kN/m (587104 kN)

102.6 kN/m (12368 kN)

4872 kN/m (587104 kN)

102.6 kN/m (12368 kN)

Pass

-10 7952 kN/m (1236484 kN)

128.3 kN/m (19958 kN)

7952 kN/m (1236484 kN)

128.3 kN/m (19958 kN)

Pass

There is a difference in calculated drained values with the hand and computer calculation. The computer

gives much higher values for the drained calculation. The reason for this is that in the hand calculation the

iteration process with punch and squeezing have not been taken into account. For the following

calculation the computer program D-Foundations will be used.

H.3.3 Calculations for the other profiles

For the profile Strong Vung Tau and profile Go Cong (based on Thu Bo) the same calculations are

performed, see Table H.5 to Table H.8. It can be conclude that the drained bearing capacity is sufficient in

all situations, but in the undrained situation the bearing capacity is only sufficient when squeezing is

allowed. One exception is the when the dam is founded at MSL -10 m with profile Strong Vung Tau: the

dam is then founded on stronger clay and the bearing capacity is sufficient.

Table H.5: Computer calculation results bearing capacity and load, profile Strong Vung Tau, undrained

Bottom at [MSL m]

Pd undrained Fd Rd squeeze [kN]

Fpull [kN] Undrained check


-2 37.4 kN/m (3714 kN)

87.8 kN/m (8738 kN)

48733 1843 Fail Pass

-5 37.4 kN/m (4498 kN)

102.6 kN/m (12368 kN)


-10 190.4 kN/m (29612 kN)

128.3 kN/m (19958 kN)

89307 2880 Pass Pass


Table H.6: Computer calculation results bearing capacity and load, profile Strong Vung Tau, drained

Bottom at [MSL m]

Pd drained Fd Rd punch [kN] Vd Punch [kN] Drained check

-2 3768 kN/m (374918)

87.8 kN/m (8738 kN)

526029 14409 Pass

-5 5523 kN/m (665486)

102.6 kN/m (12368 kN)

764832 15785 Pass

-10 8510 kN/m (1323342)

128.3 kN/m (19958 kN)

1025455

19958 Pass

Table H.7: Computer calculation results bearing capacity and load, profile 3:Go Cong (based on Thu Bo), undrained

Bottom at [MSL m]

Pd undrained Fd Rd squeeze [kN]



-2 44.9 kN/m (4472 kN)

87.8 kN/m (8738 kN)


-5 44.9 kN/m (5415 kN)

102.6 kN/m (12368 kN)

139980 1482 Fail Pass

-10 44.9 kN/m (6988 kN)

128.3 kN/m (19958 kN)

222709 1912 Fail Pass

Table H.8: Computer calculation results bearing capacity and load, profile Go Cong (based on Thu Bo) drained

Bottom at [MSL m]

Pd drained Fd Rd punch [kN] Vd Punch [kN] Drained check

-2 2808 kN/m (279415 kN)

87.8 kN/m (8738 kN)

0 0 Pass

-5 3714 kN/m (447561 kN)

102.6 kN/m (12368 kN)

0 0 Pass

-10 5274 kN/m (820182 kN)

128.3 kN/m (19958 kN)

0 0 Pass

H.4 Remarks on the hand and computer calculation of

the bearing capacity

There are some remarks on the assumptions and calculation methods that are used, these are listed

below.

Schematisation of the dam

When calculating the bearing capacity in the way described above, a remark has to be made about the

schematisation of the dam and the used method. Both the hand and computer calculation use a rough

schematisation of the dam. The real load in the middle of the dam is much higher than the uniform q-load

which was used now. Otherwise, the triangular parts of the dam will work positive for the bearing

capacity. Another point of attention is the depth of influence, this is linked to the width of the foundation.

The dam has a large width, and therefore a large depth of influence. All the strength of the underlying

sand therefore also counts for the drained calculation. With ‘small’ foundations this is okay, but for this

big dam it might be different. Finally the width of the dam is big a factor of influence of the drained

bearing capacity. It is not clear if this is correct for very large constructions.

Despite of the critical points above, the computer program D-Foundations (based on the Eurocode 7) will

be used for the preliminary design of the dam. For the final dam, it is recommended to investigate the

real functioning of the triangular parts and the influence of the width of the foundation.


Undrained/drained calculation

In the preliminary design the undrained and drained calculations are done. But when building such a big

dam there is not only undrained or only drained situation, but it might be between.

Squeezing and punch

In this design squeezing and punch is not allowed. When knowing more about the consequences of these

situations, it might be economical to allow some squeezing or punch.

Soil improvement

The Eurocode 7 describes the needed area of soil improvement before the parameters of the replaced soil

(sand) can be used. Due to the big depth and width of influence, this is a large area. More investigation

can be done to investigate whether this area could be smaller for special constructions like dams. This

would work positive on the needed replacing material.

Extra load due to settlements

Due to settlements, the dam has to be heightened to keep the crest level on the right level. This gives an

extra load. Despite of the big expected settlements and therefore big extra load, this effect is not taken

into account in the calculations of the needed bearing capacity. In the final design this has to be done.

H.5 Influence of soil improvement

In almost all situations the undrained bearing capacity is not large enough to withstand the weight of the

dam. So something has to be done regarding the foundation. In the report section 4.2.2 different types of

foundation/soil improvement are discussed. The best option for improving the bearing capacity and

reducing settlements appeared to be replacing soil by sand. Less favourable but also optional are using

sand/stone piles or concrete piles. Finally, vertical drainage with or without preloading can be applied, so

the duration of the undrained situation in the clay is shortened and the needed bearing capacity is

reached sooner. It is explained in the report, section 4.2.2, that not the whole layer has to be consolidated

to create enough bearing capacity, only a little drainage can be enough to create enough bearing capacity.

Because the differences in soil, the three soil profiles will be looked at separately but in soil profile 1 all

the assumptions and remarks are explained. The calculations are only for the situation when the bottom

and therefore foundation of the dam is at MSL-5 m. In the Table H.9 an overview is given and the

conclusion. Below that, all profiles and options are discussed.

Table H.9: Overview and conclusion soil improvement

Profile Replacing sand [m3] Drainage and preloading

[days] Combination Best option

1 9650 70 days with D=3 m and preloading

- Drainage + Preloading

2 1448 5 days with no drain - Sand

3 6273 Only pre loading, mud layer = 5 days

3374 m3 of sand and

preloading Sand and preloading

Replacing sand implies that the soft soil in the whole area of influence has to be replaced. The column

‘drainage and preloading’ says that 3.5 m of preloading is used in combination with drainage, to create

enough bearing capacity. In case of the Go Cong profile the replacing by sand will also work as preloading

for the lower clay layer.


H.5.1 Profile Weak Vung Tau

Replacing by sand

This profile consists of soft marine clay until 25 m under MSL. For the undrained calculation the cohesive

layer in the depth of influence with the lowest undrained cohesion is critical. Because the depth of

influence is very deep (0.7*B’ = 84 m for a bottom level of -5 m MSL), the whole soft marine clay layer is in

it. When removing only a part of the soft marine clay layer, the undrained layer is still leading. The bearing

capacity is higher due to the influence of punch, but also the load is higher (Table H.10 and Table H.11). So

when using sand replacement as foundation, all the soft marine clay has to be removed. According to the

Eurocode 7 this soil improvement has to be done for 4*B’ under the level of foundation, and 1.5*B’ = 181

m next to the dam. With an average replacement depth of 20 m, this gives an estimate of

(181+120.5+181)*20= 9650 m3 of soft marine clay that has to be replaced by sand, for each meter length

of the dam. A remark has to be made to the rules of 1.5*B’ and 4*B’, because of the shape and big width

of the dam this width of influence might be smaller (see remarks, H.4). More investigation has to be done

for that.

Table H.10: Soil improvement by replacing sand, profile 1; Weak Vung Tau, undrained

Sand until [MSL m]

Pd undrained [kN]

Fd [kN] Rd squeeze [kN]



-9 9537 19640 0 0 Fail Fail -20 199961 34436 0 0 Fail Fail -25 Not relevant

Table H.11: Soil improvement by replacing sand, profile 1; Weak Vung Tau, drained

Sand until [MSL m]

Pd drained [kN] Fd [kN] Rd punch [kN] Vd Punch [kN] Drained check

-9 668973 12368 19640 758697 Pass -20 845278 12368 34436 1187978 Pass -25 938971 12368 0 0 Pass

Drainage and preloading

The second option to improve the soil is by using drainage in combination with pre-loading. The drained

situation will be reached sooner and this drained bearing capacity is high enough to build on. The reason

that the bearing capacity is too low in the undrained situation is that the top material has a low undrained

shear strength of , whereas a which is 3 to 4 times higher is required for a sufficient

strength of the soil. A link between the undrained shear strength and water content can be described

with (Davidson & Springman, 2000):

(H.13)

Where is the undrained shear strength for soil with a water content equal to the plastic limit, and

the undrained shear strength for soil with a water content equal to the liquid limit.

To calculate the strain required to increase the undrained shear strength to 50 kPa (which is sufficient for

the bearing capacity) the following assumptions are made:

The water content in the weak layer is assumed to be equal to the liquid limit. This can be

justified by the slow deposition of the layer and the lack of loading op top.

The pores in the material are assumed to be completely filled with water.


, , ( - ), ( - ), Moisture content

( - , taken equal to the LL, because assumed is that the soil is very slowly

deposited and therefore very wet.

With these assumptions the required loss of water content can be calculated to obtain an increase in

undrained shear strength. An increase of undrained shear strength of 500% is chosen to be on the safe

side.

(H.14)

With known values is found that . Now the required decrease in void ratio can be calculated:

(H.15)

With the made assumptions follows that the required void ratio is equal to:

Now the required settlement can be calculated:

So, when the weak top layers have settled approximately 2% the increase in undrained shear strength is

large enough the give sufficient undrained bearing capacity for the whole dam.

In Table H.3 is calculated that the undrained bearing capacity is equal to , so approximately

to 3.5 m of sand under the water table. When this surcharge is applied the time needed to reach 2%

settlements for the weak layers is calculated with D-settlement, see Table H.12.

Table H.12: Time required to reach sufficient undrained shear strength with 3.5 m preloading, profile 1

Profile Drain centre-to-centre distance [m]

Time required to reach 2% settlement of soft top layer [days]

Weak Vung Tau (20 m soft clay) 1 8 2 32 3 70 5 170 10 450 No drains 800

After the time required to create enough bearing capacity, the whole dike core can be constructed.

Although after these required times the bearing capacity is sufficient, one has still to take into account

that large settlements can still occur, in contrast with soil replacement.

When looking to the possibilities for increasing the bearing capacity, drainage in combination with

preloading is chosen. While a lot of sand is needed for replacing, drainage might be a lot of work to install.

But, when having a good planning (with overload and shortening the building time) it might be the best

solution. The drainage distance not only depends on the waiting time to construct the dam, but also on

the total settlement curve: in the life time of the dam only small settlements are allowed. More details

about settlements are described in appendix I. From the settlements the real drainage distance will be


chosen. For only bearing capacity a drainage distance of 3 m might be sufficient to shorten the building

time.

H.5.2 Profile Strong Vung Tau

Replacing by sand

In the strong Vung Tau profile, the soil is much easier to build on. Already on MSL -8 m there is a sand

layer present. The clay layer under it is strong: cu,d = 37 kPa. Therefore no squeezing will occur. The

volume of needed sand is: (181 + 120.5 + 181) * 3 = 1448 m3 per m’. In Table H.13 and Table H.14 the

calculations are given.

Table H.13: Calculation Strong Vung Tau, Undrained

Sand until [MSL m]

Pd undrained [kN]

Fd [kN] Rd squeeze [kN] Fpull [kN] Undrained check

-8 28201 19640 0 0 Pass -10 28201 19640 0 0 Pass

Table H.14: Calculation Strong Vung Tau, Drained

Sand until [MSL m]


-8 718065 12368 774189 16721 Pass -10 730861 12368 827505 19640 Pass


When using preloading to increase the bearing capacity, after a few days with a loading of 3.5 m sand the

soil is sufficiently strong enough to build the dam (Table H.15).

Table H.15: Time required to reach sufficient undrained shear strength with 3.5 m preloading, profile Strong Vung Tau

Profile Drain centre-to-centre distance [m]

Time required to reach 2% settlement of soft top layer [days]

Strong Vung Tau (3 m soft clay) No drains 5

When looking to the possibilities for increasing the bearing capacity, preloading might be seen as the

easiest one: only 3.5 m preloading for a few days and drainage is not needed because of the short period

needed to create enough bearing capacity. But the settlements during lifetime will then be big (see

appendix I). Only a few meters of clay has to be replaced by sand, then the settlements are much lower.

So the replacing by sand-option is chosen.

H.5.3 Profile Go Cong (based on Thu Bo)

Replacing by sand

The Go Cong profile looks like profile 2: clay on the bottom, then layers of sand and clay, and finally sand.

In this profile, the middle clay (clay, silty) is less strong then the clay layer in the strong Vung Tau profile.

When schematizing the soil profile in D-Foundations it seems that the middle clay layer does not have

enough bearing capacity.

In Table H.16 and Table H.17 the calculation results of the bearing capacity when replacing clay by sand

are given. Because squeeze is not allowed in the project, the situation with no sand but with squeeze is

not an option. Between the clay layers is a sand layer of 3 meter. The needed amount of sand for


replacing is (181 + 120.5 + 181) * 13 = 6273 m3 of sand per m’, but 1448 m

3 sand per m’ can be taken from

the project location.

Table H.16: Calculation results bearing capacity and load, profile 3: Go Cong (based on Thu Bo) Undrained

Table H.17: Calculation results bearing capacity and load, profile 3: Go Cong (based on Thu Bo) Drained

Sand until [MSL m]


-5 (so no sand) 476314 12368 0 0 Pass -8 476134 12368 508269 16721 Pass -12 515485 12368 0 0 Pass -15 515485 12368 644022 26996 Pass -18 548697 12368 0 0 Pass Sand until [MSL m]



When again looking at the possibilities to increase the undrained shear strength by preloading with 3.5 m

of sand, as well the first layer (mud) as the third layer (clay) need to be strengthened. Again the method

as explained in H.5.1 is used. Used parameters and results are presented in Table H.18, and in Table H.19

the results are given. It seems that with a preloading of 3.5 m the middle clay layer doesn’t reach the

required strain. A maximum strain of 4% is possible, this gives a of 36.4 kPa. Hand calculations and

calculations in D-Foundations are done, and this undrained shear strength is high enough to create

enough bearing capacity: a bearing capacity (design value) of 137.8 kN/m is created. This gives enough

bearing capacity for the load of the dam (102 kN/m). The soil above the clay layer is not calculated in the

load, but also not in the bearing capacity.

Table H.18: Assumed parameters and results to obtain sufficient undrained shear strength

Layer 1: mud Layer 3: clay

Liquid limit LL [%] 61.3 38.2 Plastic limit PL [%] 30.9 19.9 Water content W [%] 61.3 23.9 Initial undrained shear strength cu [kPa 11.8 33.2 Initial void ratio e0 [-] 1.981 0.725

Required strain ε [%] 1.1 4.0, max cu = 36.4 kPa

Sand until [MSL m]

Pd undrained [kN] Fd [kN] Rd squeeze [kN] Fpull [kN] Undrained check

-5 (so no sand)

5415 12368 139980 1482 Pass

-8 8432 16721 0 0 Fail -12 25681 26996 0 0 Fail -15 25681 26996 0 0 Fail -18 Not relevant


Table H.19: Time required to reach sufficient undrained shear strength with 3.5 m surcharge for profile Go Cong

Layer Drain centre-to-centre distance [m]

Time required to reach required settlement of layer [days]

Go Cong layer 1: mud No drains 55

Go Cong layer 3: clay No drains never

It is possible to use only preloading for increasing the bearing capacity, but then the settlements during

lifetime will probably be large. Therefore the two options are combined: the upper layer will be replaced

by sand. Then a surcharge of 3.5 m sand will be placed to create enough bearing capacity in the middle

clay layer.

Now the preloading is more than 3.5 m (the mud layer with an effective weight of 4.6 kN /m3 is replaced

by sand of σ’=10 kN/m3) also the settlements and drained situation will take place sooner. The bearing

capacity of the original soil is still enough to handle this, the load on the clay layer is 102 kN/m for the

dam and (10-4.6)*7=37.8 kN/m. This is more than the allowable 137.8 kN/m, but with building phases the

drainage will already start after the soil improvement of the upper layer.

The amount of soil that has to replaced is then only the upper layer, so (181+120.5+181)*7 m = 3374 m3.

Project Flood Defence HCMC | APPENDIX Settlements Page 85 / 140

SETTLEMENTS I

I.1 Introduction

The VT-GC will be located in the estuary. Therefore, a weak subsoil built up out of fine sediments can be

expected. This can impose large settlements. The settlement will consist of two parts:

1. Consolidation (or primary settlement)

This is consolidation due the slow decrease of excess pore pressure caused by the loading (and

thus a slow increase of effective pressure densification of the soil settlements).

Impermeable materials, such as clays, take a long time to consolidate.

2. Creep (or secondary settlement)

This is the on-going settlements of visco-elastic materials, such as clays and organic material.

Although the imposed stress does not change the material slowly deforms. Settlements due to

creep behave logarithmic with time.

Whilst consolidation usually causes the majority of the settlements creep is important to take into

account as well, because of the long lifetime of the structure. A typical settlement figure for a clay is given

in Figure I-1.

Figure I-1: Typical figure of settlement in time for a clay

I.2 Assumed dam profile for settlement calculations

The assumed dam profile for settlement calculations is presented in Figure I-2. The bottom is chosen at a

level of MSL -5 m, because this is the average depth for most of the trajectory of the dam.

5 9 25

+6 m above MSL

+3 m

+ 0.0 m

30 9

1:5

1:3

1:3

-5 m

1525

118 m

Figure I-2: Dam profile for settlement calculation

I.3 Hand calculation of the settlements: example for

the Weak Vung Tau profile

A settlement calculation by hand will be made. First the Flamant method will be used to calculate the

stress distribution under the dam caused by the self-weight of the dam, and then the actual settlement

calculation will be made using the Bjerrum method.


The stress distribution formula of Flamant reads: (Verruijt, 2001)

( (

)

) (I.1)

Where:

= the pressure put on the soil [kPa]

= half the width of the load [m]

= the depth below the surface [m]

A dam profile at a depth of 10 m under MSL is assumed. The load caused by the dam is schematised as

Figure I-3. The increase of vertical stress caused by the load is given in Figure I-4

p = Δσ’v = 140 kPa≈ weight 5 m dry sand (above water) + 5 m of

wet sand (below water) – 5 m of water

z

a ≈ 40 m

Figure I-3: Schematisation to calculate Flamant stress

redistribution

Figure I-4: Increase in effective stress as a function of

depth with Flamant method

An increase of vertical stress causes settlements. Because of the large dimensions of the dam this increase

will influence many layers below the dam. It is therefore important to have information about the

presence of compressible layers until a fairly large depth.

The soil is subdivided into 1.5 or 1.0 m thick layers (because the stress differs with depth), and

subsequently in the middle of each layer the initial vertical stress , the stress increase (Flamant), the

change in void ratio, the strain and the settlement is calculated. The following Bjerrum formula is used to

calculate the change is void ratio (Verruijt, 2001):

⏟

⏟

(I.2)

Where:

= initial void ratio [-]

= primary consolidation parameter [-]

= the increase of effective stress in the layer [kPa]


the effective stress at time [kPa]

= the creep parameter [-]

= time after the start of the consolidation starts [s]

= time when creep starts, usually taken as the time primary settlement has ended [s]

The strain is calculated as follows:

(I.3)

The results are presented in Table I.1.

Creep is only taken into account when consolidation has ended. The time necessary to reach the end of

the primary settlements is calculated using the consolidation formula (Barends, 2010):

(

) (I.4)

Where:

= the pore water pressure at time [kPa]

= the initial pore water pressure at the middle of the layer [kPa]

= time after the start of the consolidation starts [s]

= the consolidation coefficient [m2/s]

= the drainage length (the maximum distance a water particle has to travel to reach a

permeable layer) [m]

Usually is said that consolidation is over when

. We know that for the clay layer (which takes

longest to consolidate) , and also that (drainage above and below clay

layer assumed). Thus we can calculate that . Because 228

years is bigger than the lifetime of 50 years creep is not taken into account.

A graph of the total settlement as a function of depth is given in Figure I-5. In this graph can be seen that

the settlement rate decreases with increasing depth. The majority of the settlements occurs in the top

parts of the clay layer.


Table I.1: Hand calculation Bjerrum settlement

Figure I-5: Settlement as a function of depth

-40

-35

-30

-25

-20

-15

-10

-5

0

0 1 2 3 4 5 6

De

pth

be

low

MSL

[m

]

total settlement [m]

Layer number

Layer name

Depth middle layer

Thickness z γsat Cc Cα e0 σ'v,initial Δσ’v e0-e consolidation

e0-e creep

e0-e total

e ε Δh

[-]

[m + MSL]

[m] [m] [kN/m3] [-] [-] [-] [kPa] [kPa] [-] [-] [-] [-] [-] [m]

1 Soft clay -5.75 1.5 0.75 18 0.94 0.013 2 6.0 140 1.303 0.000 1.303 0.697 0.768 1.152 2 Soft clay -7.25 1.5 2.25 18 0.94 0.013 2 18.0 140 0.887 0.000 0.887 1.113 0.420 0.629 3 Soft clay -8.75 1.5 3.75 18 0.94 0.013 2 30.0 140 0.708 0.000 0.708 1.292 0.309 0.463 4 Soft clay -10.25 1.5 5.25 18 0.94 0.013 2 42.0 140 0.598 0.000 0.598 1.402 0.249 0.374 5 Soft clay -11.75 1.5 6.75 18 0.94 0.013 2 54.0 140 0.522 0.000 0.522 1.478 0.210 0.316 6 Soft clay -13.25 1.5 8.25 18 0.94 0.013 2 66.0 140 0.464 0.000 0.464 1.536 0.183 0.274 7 Soft clay -14.75 1.5 9.75 18 0.94 0.013 2 78.0 139 0.418 0.000 0.418 1.582 0.162 0.243 8 Soft clay -16.25 1.5 11.25 18 0.94 0.013 2 90.0 139 0.381 0.000 0.381 1.619 0.145 0.218 9 Soft clay -17.75 1.5 12.75 18 0.94 0.013 2 102.0 138 0.350 0.000 0.350 1.650 0.132 0.198 10 Soft clay -19.25 1.5 14.25 18 0.94 0.013 2 114.0 138 0.323 0.000 0.323 1.677 0.121 0.181 11 Soft clay -20.75 1.5 15.75 18 0.94 0.013 2 126.0 137 0.300 0.000 0.300 1.700 0.111 0.167 12 Soft clay -22.25 1.5 17.25 18 0.94 0.013 2 138.0 136 0.280 0.000 0.280 1.720 0.103 0.155 13 Soft clay -23.5 1 18.5 20 0.94 0.013 2 149.0 135 0.264 0.000 0.264 1.736 0.096 0.096 14 Soft clay -24.5 1 19.5 20 0.94 0.013 2 159.0 135 0.250 0.000 0.250 1.750 0.091 0.091 15 Loose sand -25.75 1.5 20.75 20 0.019 0 0.633 171.5 134 0.005 0.000 0.005 0.628 0.003 0.004 16 Loose sand -27.25 1.5 22.25 20 0.019 0 0.633 186.5 133 0.004 0.000 0.004 0.629 0.003 0.004 17 Loose sand -28.75 1.5 23.75 20 0.019 0 0.633 201.5 131 0.004 0.000 0.004 0.629 0.003 0.004 18 Loose sand -30.25 1.5 25.25 20 0.019 0 0.633 216.5 130 0.004 0.000 0.004 0.629 0.002 0.004 19 Loose sand -31.75 1.5 26.75 20 0.019 0 0.633 231.5 129 0.004 0.000 0.004 0.629 0.002 0.003 20 Loose sand -33.25 1.5 28.25 20 0.019 0 0.633 246.5 127 0.003 0.000 0.003 0.630 0.002 0.003 21 Loose sand -34.75 1.5 29.75 20 0.019 0 0.633 261.5 126 0.003 0.000 0.003 0.630 0.002 0.003 22 Loose sand -36.25 1.5 31.25 20 0.019 0 0.633 276.5 124 0.003 0.000 0.003 0.630 0.002 0.003 23 Loose sand -37 1.5 32 20 0.019 0 0.633 284.5 123 0.003 0.000 0.003 0.630 0.002 0.003

TOTAL 4.6


I.4 Computer calculation of the settlements: example

Weak Vung Tau profile

A computer calculation of the settlements is performed as well. The same layering and parameters are

assumed as for the hand calculation. The input is given in Figure I-6 and the settlement-time diagram of

the profile under the crest of the dike in Figure I-7.

Figure I-6: D-Settlement input Figure I-7: Time-settlement diagram

The computer calculation gives a settlement value after 18250 days (50 years) of 3.318 m. Settlement

after 100.000 days is 4.5 m. With the hand model, which cannot take the time required to reach full

consolidation into account, the settlement after 100.000 days (so after consolidation had ended) is 4.6

years.

I.5 Calculations for the other profiles

Now it is proven that the hand calculation model seems to be working it will be used to calculate the

settlements of the other profiles.

Hand calculations D-Settlement calculations

Estimated full consolidation time [years]

Settlement after end of consolidation [m]

Settlement after 50 years [m]


Infinite settlement [m]

Weak Vung Tau 228 4.6 3.3 4.1 4.5 Strong Vung Tau 410 2.3 1.8 1.9 2.1 Go Cong 198 4.3 2.6 3.1 3.3

Settlement figures in time for the point of maximum settlement for the different profiles are given in

Figure I-8, Figure I-9 and Figure I-10.

D-S

ettle

me

nt 9

.2 : S

ettle

me

nts

Bje

rrum

-10

sla

pp

e k

lei.s

li

Stie

ltjesw

eg

26

28

CK

DE

LF

T

Ph

on

e0

88

- 33

57

20

0

Fa

x

da

ted

rw.

ctr.

form

.

DE

LT

AR

ES

-

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-Annex

-

Input View

Materials

Soft CLAY

SAND, loose

-150,000 150,000

SAND, loose

Soft CLAY

Dam

-150 -100 -50 0 50 100 150 200 250 300 350m

-75

-70

-65

-60

-55

-50

-45

-40

-35

-30

-25

-20

-15

-10

-5

0

5

10

m

D-S

ettle

me

nt 9

.2 : S

ettle

me

nts

Bje

rrum

-10

sla

pp

e k

lei.s

li

Stie

ltjesw

eg

26

28

CK

DE

LF

T

Ph

on

e0

88

- 33

57

20

0

Fa

x

da

ted

rw.

ctr.

form

.

DE

LT

AR

ES

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15-9

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-Annex

-

Time-History

Vertical 4 (X = -9,000 m; Z = 0,000 m)

Method = NEN - Bjerrum with Terzaghi

Depth = 5,000 (-) [m]

Settlement after 100000 days = 4,522 [m]

1 10 100 1000 10000 100000

4,000000

3,000000

2,000000

1,000000

0,000000

Se

ttle

me

nt

[m]

1 10 100 1000 10000 100000

Time [days]

129,772095

129,772100

129,772105

129,772110

Eff

ective

str

ess [

kP

a]


Figure I-8: Settlement in time VT-weak profile

Figure I-9: Settlement in time VT-strong profile

Figure I-10: Settlement in time GC profile

Figures for the settlement over depth after 50 years computed with D-Settlement are given in Figure I-11,

Figure I-12 and Figure I-13. The results of the hand calculation are presented in Figure I-14. D

-Se

ttlem

en

t 9.2

: Se

ttlem

en

ts V

Tw

ea

kB

jerru

m -5

sla

pp

e k

lei.s

li

Stie

ltjesw

eg

26

28

CK

DE

LF

T

Ph

on

e0

88

- 33

57

20

0

Fa

x

da

ted

rw.

ctr.

form

.

DE

LT

AR

ES

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-

Time-History

Vertical 4 (X = -9,000 m; Z = 0,000 m)


Depth = 5,000 (-) [m]


1 10 100 1000 10000 100000

4,000000

3,000000

2,000000

1,000000

0,000000

Se

ttle

me

nt

[m]

1 10 100 1000 10000 100000

Time [days]

129,772095

129,772100

129,772105

129,772110

Eff

ective

str

ess [

kP

a]

D-S

ettle

me

nt 9

.2 : S

ettle

me

nts

VT

stro

ng

Bje

rrum

-5 s

lap

pe

kle

i.sli

Stie

ltjesw

eg

26

28

CK

DE

LF

T

Ph

on

e0

88

- 33

57

20

0

Fa

x

da

ted

rw.

ctr.

form

.

DE

LT

AR

ES

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-

Time-History

Vertical 4 (X = -9,000 m; Z = 0,000 m)


Depth = 5,000 (-) [m]


1 10 100 1000 10000 100000

2,000000

1,500000

1,000000

0,500000

0,000000

Se

ttle

me

nt

[m]

1 10 100 1000 10000 100000

Time [days]

148,183555

148,183560

148,183565

148,183570

Eff

ective

str

ess [

kP

a]

D-S

ettle

me

nt 9

.2 : S

ettle

me

nts

CG

-TB

Bje

rrum

-5.s

li

Stie

ltjesw

eg

26

28

CK

DE

LF

T

Ph

on

e0

88

- 33

57

20

0

Fa

x

da

ted

rw.

ctr.

form

.

DE

LT

AR

ES

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-

Time-History

Vertical 4 (X = -9,000 m; Z = 0,000 m)


Depth = 5,000 (-) [m]


1 10 100 1000 10000 100000

3,000000

2,000000

1,000000

0,000000

Se

ttle

me

nt

[m]

1 10 100 1000 10000 100000

Time [days]

139,046190

139,046195

139,046200

139,046205

Eff

ective

str

ess [

kP

a]


Figure I-11: Settlement with depth after 50

years VT-weak profile

Figure I-12: Settlement with depth after 50 years VT-strong profile

Figure I-13: Settlement with depth after 50

years GC profile

Figure I-14: Hand calculation settlement with depth at end of

consolidation

The hand calculation overestimates the settlements. This is especially in the case of the Go Cong-profile,

where computer calculation gives much lower settlements for the mud layer then the hand calculations.

Also the settlements computed with the computer are lower after 50 years, because in the hand

calculation only the full primary settlement can be calculated, although primary settlement has not

finished after 50 years.

Cleary observable is that especially the top layers settle the most. This is to be expected, because these

layers are prior not consolidated at all because of the lack of vertical effective pressure. The bigger the

difference between the initial vertical effective pressure and the pressure induced by the dam the bigger

the settlements.

D-S

ettle

me

nt 9

.2 : S

ettle

me

nts

VT

we

akB

jerru

m -5

sla

pp

e k

lei.s

li

Stie

ltjesw

eg

26

28

CK

DE

LF

T

Ph

on

e0

88

- 33

57

20

0

Fa

x

da

ted

rw.

ctr.

form

.

DE

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AR

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Depth-History

Vertical 4 (X = -9,000 m; Z = 0,000 m)


SAND, loose

Soft CLAY

0 1 2 3

Settlement [m]

-40

-30

-20

-10

Dep

th [

m]

0 200 400 600

Stress [kN/m²]

-40

-30

-20

-10

Dep

th [

m]

Sigma water Sigma totalSigma effective

D-S

ettle

me

nt 9

.2 : S

ettle

me

nts

VT

stro

ng

Bje

rrum

-5 s

lap

pe

kle

i.sli

Stie

ltjesw

eg

26

28

CK

DE

LF

T

Ph

on

e0

88

- 33

57

20

0

Fa

x

da

ted

rw.

ctr.

form

.

DE

LT

AR

ES

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Depth-History

Vertical 4 (X = -9,000 m; Z = 0,000 m)


SAND, loose

Moderate marine CLAY

Clayey SAND

Soft CLAY

0,0 0,5 1,0 1,5

Settlement [m]

-40

-30

-20

-10

Dep

th [

m]

0 200 400 600 800

Stress [kN/m²]

-40

-30

-20

-10

Dep

th [

m]

Sigma water; Initial stress Sigma effective; Final stressSigma water; Final stress Sigma total; Initial stressSigma effective; Initial stress Sigma total; Final stress

D-S

ettle

me

nt 9

.2 : S

ettle

me

nts

CG

-TB

Bje

rrum

-5.s

li

Stie

ltjesw

eg

26

28

CK

DE

LF

T

Ph

on

e0

88

- 33

57

20

0

Fa

x

da

ted

rw.

ctr.

form

.

DE

LT

AR

ES

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-

Depth-History

Vertical 4 (X = -9,000 m; Z = 0,000 m)


SAND, loose

CLAY, silty

SAND, muddy

MUD

0,0 0,5 1,0 1,5 2,0 2,5

Settlement [m]

-40

-30

-20

-10

De

pth

[m

]

0 200 400 600 800

Stress [kN/m²]

-40

-30

-20

-10

De

pth

[m

]

Sigma water; Initial stress Sigma effective; Final stressSigma water; Final stress Sigma total; Initial stressSigma effective; Initial stress Sigma total; Final stress

-40

-35

-30

-25

-20

-15

-10

-5

0 1 2 3 4 5

De

pth

be

low

MSL

[m

]

total settlement [m]

GC VT-weak VT-strong


I.6 Maintain height calculations

Calculations are performed for the settlements assuming that the dam is continuously maintained at the

same height. The option ‘maintain height’ in the computer programme D-Settlement is used for this

purpose. The increase of settlements are imposed by the extra settlements caused by the settlement

compensation. This extra settlement has to be taken into account when calculating the required height

necessary to compensate the settlements. Results are presented in Table I.2.

Table I.2: Maintain height calculations for the settlements and increase relative to the calculations without

maintaining the profile height


Increase after 50 years [m]


Increase after 100 years [m]

Infinite settlement [m]

Increase infinite settlements [m]

Weak Vung Tau 4.0 0.7 4.9 0.8 5.4 0.9 Strong Vung Tau 1.9 0.1 2.1 0.2 2.3 0.2 Go Cong 2.9 0.3 3.4 0.3 3.6 0.3

I.7 Influence of soil improvement on settlements

Part of the subsoil can be excavated or improved to improve the bearing capacity and to minimise the

settlements. In this section the effect of these soil improvements will be investigated. Assumed is that the

soil is improved such that it is made incompressible. The unit weight of this improved soil is assumed to

be equal to the unit weight of the original soil.

Thus Figure I-15 can be constructed. One might see that especially replacing the upper few meters is

favourable. Increasing the thickness of the soil improvement leads to decreasing efficiency for settlement

reduction. Also only replacing the soft top layer is favourable. Replacing the sand and stiff clay (VT-strong,

-3 m below surface level) or the clayey sand (Go Cong, -7 below surface level) lead to only insignificant

decrease of the settlement.

Figure I-15: Influence of the amount of soil improvement on the total settlements after 50 years

0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

4.0

0 2 4 6 8 10

Tota

l se

ttle

me

nt

in 5

0 y

ear

s [m

]

Thickness soil improvement [m]

VT-weak

VT-strong

GC


Replacing the top weak layer(s) by a loose sand layer with a saturated volumetric weight of

, to increase the bearing capacity, and with a sea bottom level on MSL -5 m gives the more

precise calculated total settlements as given in Table I.3.

Table I.3: Settlements of profiles when the complete soft top layer is replaced by loosely packed sand

Profile Thickness of replaced soft layer [m]

Total settlements after 50 years [m]

Total settlements after infinite time [m]

Weak Vung Tau 20.0 (until MSL -25 m) 0.14 0.14 Strong Vung Tau 3.0 (unit MSL -8 m) 0.49 0.79 Go Cong 13.0 (until MSL -12 m) 0.30 0.31

However, in the report, section 4.2.2 is concluded that for the Weak Vung Tau-profile the application of

vertical drains and preloading is preferable, because 20 m soil replacement is not viable. In Table I.4 for

several preloading times and drain distances the settlements during the lifetime of the dam are calculated

(sea bottom level MSL -5 m, and top of the sand surcharge MSL + 5 m).

Table I.4: influence of preloading on the settlements during the lifetime of the dam

Centre to centre distance drains [m]

Surcharge time [years]

Total settlements after surcharge time [m]

Settlements 50 years after end surcharge time [m]

Settlements during 50 years lifetime [m]

Total settlements after infinite time [m]

1 1 3.69 4.35 0.66 4.55 1 2 3.77 4.36 0.59 4.56 1 3 3.82 4.36 0.54 4.56

2 1 3.32 4.35 1.03 4.55 2 2 3.73 4.36 0.63 4.56 2 3 3.81 4.36 0.55 4.56

3 1 2.25 4.35 2.10 4.55 3 2 3.16 4.36 1.20 4.56 3 3 3.55 4.36 0.81 4.56

Decreasing the centre-to-centre distance and increasing the surcharge time leads to smaller settlements

during the lifetime of the dam. It is chosen to limit the settlements of the subsoil during the lifetime of the

dam to approximately 0.5 m. Therefore a centre-to-centre drain distance of 2 m and a preloading time of

3 years is selected.

This gives a settlement of 3.81 m after 3 years of surcharge. After this the sand surcharge can be built

further to obtain the dam profile and the required crest height. 50 years after construction the dam will

have settled approximately 0.55 m. During the construction of the preload and the time it is applied

special measures has to be taken to prevent the preload from eroding.

Thus for the three profiles, with either soil replacement or drainage and preloading the settlements

during the lifetime of the dam can be minimised to approximately 0.5 m. However, in these calculations

the additional settlement caused by the extra material required to compensate for the settlements is not

taken into account.


I.8 Sensitivity analysis of settlements of weak layers

Because much is unknown about the thicknesses of the weak top layer under the dam, a sensitivity

analysis will be performed to investigate the influence of the bottom depth and the soft soil layer

thickness on the maximum settlement of the layer. The Thu Bo mud parameters are assumed to be

representative for the soft soil layer. The results are summarised in Figure I-16.

The deeper the sea (so the heavier the dam) the more settlements occur. And increasing thickness of the

layer leads to larger settlements.

Figure I-16: Total settlements of the weak top layer as function of the sea bottom depth and the bottom depth of the

soft soil layer

When the settlements after 50 year are calculated for the same case the results (Figure I-17) show much

lower settlements. The percentage of total settlements reached after 50 years are presented in Figure

I-18. It can clearly be seen that the time needed to reach full consolidation increases with increasing

thickness of the layer.

Figure I-17: Settlements of the weak top layer as function of the sea bottom depth and the bottom depth of the soft

soil layer after 50 years

0.5

0.5

1

1

1

1.5

1.5

1.5

2

2

2

2.5

2.5

2.5

3

3

3

3.5

3.5

3.5

4

4

4

4.5

4.5

4.5

5

5

5

5.5

5.5

5.5

6

6

6

6.5

6.5

6.5

7

7

7

7.5

7.5

7.5

8

8

Bottom level of soft layer [m (reference MSL)]

Depth

sea b

ott

om

[m

(re

fere

nce M

SL)]

-40 -35 -30 -25 -20 -15 -10 -5

-12

-11

-10

-9

-8

-7

-6

-5

-4

-3

0

1

2

3

4

5

6

7

8

0.25

0.25

0.5

0.5

0.75

0.75

0.75

1

1

1

1.25

1.25

1.25

1.25

1.5

1.5

1.5

1.5

1.5

1.75

1.75

1.75

1.75

1.75

1.75

2

2

2

2

2

2

2.25

2.25

2.25

2.25

2.25

2.5


Depth

sea b

ott

om

[m

(re

fere

nce M

SL)]

-40 -35 -30 -25 -20 -15 -10 -5

-12

-11

-10

-9

-8

-7

-6

-5

-4

-3

0

0.5

1

1.5

2

2.5


Figure I-18: Percentage of total consolidation reached after 50 years

Also the settlement sensitivity study can be used to estimate the effect of soil improvement. To do this

one can determine the difference in settlements between 2 depths of the dam. For an example, see

Figure I-19.

Figure I-19: Example to demonstrate the effect of soil improvement, Case: bottom depth = -5 m MSL, depth bottom

soft layer = -15 m MSL, influence of 2.5 meters soil replacement on the settlements after infinite time leads to a

decrease of settlements of approximately 0.4 m.

Although soil improvement always leads to decrease of settlements after infinite time, it does not mean

that the settlements after 50 years (when no additional consolidation measure are taken) also decrease: it

could be the case that there is more consolidation because thinner layers consolidate quicker! (The

consolidation time increases quadratic with increasing layer thickness.) A demonstration of this effect can

be observed in Figure I-20. The transition point for settlements after 50 years appears to be at a soft soil

layer depth of approximately 7.5 m.

20

20

20

30

30

30

40

40

40

50

50

50

60

60

60

70

70

70

80

80

80

90

90

90

10

0

10

0


Depth

sea b

ott

om

[m

(re

fere

nce M

SL)]

-40 -35 -30 -25 -20 -15 -10 -5

-12

-11

-10

-9

-8

-7

-6

-5

-4

-3

20

30

40

50

60

70

80

90

100


Figure I-20: Example to demonstrate the negative effect of soil improvement, Case: bottom depth = MSL -5 m, depth

bottom soft layer = MSL -20 m, influence of 2.5 meters soil replacement on the settlements after 50 years leads to an

increase of settlements of approximately 0.3 m.

In conclusion about the influence of soil improvements on the settlements it can be said that:

Soil improvement has a higher efficiency per meter improvement when the total thickness of the

soft layer is thinner.

Soil improvement is only efficient for soft layers.

The efficiency of soil improvement per meter improvement decreases with increasing amount of

soil improvement.

Project Flood Defence HCMC | APPENDIX Stability Page 97 / 140

STABILITY J

J.1 Introduction

Many methods are present to calculate the stability of a slope. For that reason it might be clear that no

model is perfect. In this preliminary design the method of Bishop will be used. By assuming a circular slip

surface the stability factor F (strength/load) for stability is calculated, in which the smallest F is critical.

This method is only valuable for circular slip surfaces and takes not all aspects into account, but it differs

not much with more difficult methods and for several limits it gives the right values. Therefore it is a

method that is used often in geo-engineering.

J.2 Assumed parameters

The assumed parameters can be found in Appendix B The three calculated profiles are:

Weak Vung Tau

Strong Vung Tau

Go Cong with Thu Bo parameters

The bottom level will be at 5 m below Mean Sea Level for a large part of the dam, so this will be taken for

the design. For the water level the tide has only influence in the first meters of the dam. So in the middle

of the dam the water level stays at MSL. The phreatic line during low tide is assumed to be as in Figure J-1.

With sea level rise and storm the phreatic line in the dam is assumed to be maximum MSL +1 m (in the

middle of the dam).

Figure J-1: Phreatic line in the dam

+7.6 m MSL

+4.2 m MSL

+0.0 m MSL

-5.0 m MSL

713.61616.8 15.6 25 8.4 17.5

1:4

1:41:3

1:3.5

20

+2.4 m MSL

Figure J-2: Cross section dam for stability check


As cross section for the dam the cross section of Figure J-2 of the dam with a sand core is used. In this

cross section on the sea side a slope of 1:4 followed from the wave run-up, explained in section 4.1 of the

report. On the lake side above water a slope of 1:3 is chosen, this is the steepest natural slope possible in

sand. Under water the slope has to be the less steep, a slope of 1:3.5 is chosen. It does not have to be the

slope of the sea side, because on this side the water is calmer, the waves are smaller. For all slopes the

stability will be checked.

The position and size of the outer berm and inner berm (with a road on top) are determined by the wave

run up (crest level) and the water level in the dam (see section 4.3.4 in the report). Therefore, these

variables are only checked, and not looked at its influence.

Water levels on the sea side are assumed to be normal (MSL), low (MSL -2.2 m) and high (MSL +3.95 m).

This high water is based on the still water level (including long duration storms, 3.2 m) and sea level rise

(0.75 m). On the lake side the water level is also taken as normal (MSL), low (MSL -2.2 m) and high (MSL

+1 m). The values for the low water levels are based on the mean lower low water level. This value is very

low and will not be reached often. More likely to happen is a water level of MSL -1.6 m when sea level rise

is also taken into account. While in normal circumstances the lake is in open connection with the sea, the

gates will be closed when a water level rise of MSL +1 m is expected in the lake. This is the maximum

water level allowed due to salt intrusion.

J.3 Calculation method

The method of Bishop, as most methods, is based on the circular slip surface. This divides the slip curve in

several slices with a load (Figure J-3). Each slice causes shear stress τ, which together has to withstand the

load of the soil. To be safe, a stability factor F is introduced, when F>1,0 the construction is expected to

be stable. Bishop has taken different aspects in account: weight of the soil, moments, shear strength on

the bottom and the friction between the slices. The formula for the vertical equilibrium will therefore be:

(J.1)

The friction force is with stability factor is:

(J.2)

By substitution the stability factor can be calculated with the next formula:

∑ ( )

(

)

∑

(J.3)

It can be seen that all the slices have to be calculated separately. This calculation has to be made for all

possible slip circles (different mid points). This is a lot of calculation work, so the computer program D-

Geo stability is used to calculate the stability factor. In this program various models can be used, for the

design of the dam the Bishop method is used.


Figure J-3: Slip circle with slices

J.4 Calculation of Weak Vung Tau

The stability of the dam on the soil of the Weak Vung Tau profile is tested in D-Geo Stability. The water

level on the both sides is low: MSL -1.6 m. In the middle the water is at MSL. In Figure J-4 the input is

given.

Figure J-4: Input profile Weak Vung Tau in D-Geo Stability with low water level

The most critical slip circles in this situation are given in Figure J-5 and Figure J-6. The stability factor is

1.19, so the dam is stable, but this stability factor is low. This critical slip surface is located on the right

side of the dam. On the left, the dam is stable with a stability factor of 1.22.


Figure J-5: Critical slip circle on the left side Figure J-6: Critical slip circle on the right side

Different situations are tested in the same way. Due to the width and shape of the dam the two sides do

not correlate with each other for the stability, therefore they are treated seperately. The results of the

different situations are tested, results of the calculations can be seen in Table J.1. The situation with the

sea water level at MSL -2.2 m is unlikely to happen, but also with the water level of MSL -1.6 m the

stability factor is low. Therefore a different slope of the right side is chosen: 1:4 instead of 1:3.5. The

safety factor on the lake side will then be 1.38 with a water level of MSL -1.6 m.

Table J.1 Stability factor profile Weak Vung Tau

Water level sea side [MSL m]

Water level in dam [MSL m]

Stability factor [-]

Water level lake side [MSL m]

Water level in dam [MSL m]

Stability factor [-]

-2.2 0 1.29 -2.2 0 1.10 -1.6 0 1.22 -1.6 0 1.19 0 0 1.53 0 0 1.44 3.95 0 1.70 1 0 1.50 -1.5 1 1.23 -1.5 1 1.14 -1 1 1.34 -1 1 1.15 0.75 1 1.52 0.75 1 1.46 3.95 1 1.70 1 1 1.50

Also the stability of the upper part of the dike is calculated. In Figure J-7 this slip circle can be seen. The

safety factor is 1.74.

Figure J-7: Critical failure slope at top of dam


J.5 Calculations for the other profiles

Also for the two other profiles the same calculations are made, see Table J.2 and Table J.3. This time only

the normal ‘mean’ situation and extreme situations are calculated. In these calculations the soil

improvement chosen before (replacing the upper layer by sand, for profiles Strong Vung Tau and Go

Cong) is taken into account. The conclusion is that at both dams the inner and outer slopes are stable. To

see the effect of the soil improvement, the calculation is also done with the Strong Vung Tau profile,

without soil improvement. The soil improvement has a positive effect on the stability: the stability factor

is much higher. Also, when soil improvement is applied, the slip circle only goes through the dam. Without

soil improvement the slip circle also goes through the soft soil (see Figure J-8 until Figure J-12).

Table J.2: Stability factor profile Strong Vung Tau, with soil improvement (sand)

Water level sea side [m MSL]

Water level in dam [m MSL]

Stability factor Water level lake side [m MSL]


Stability factor

-2.2 0 1.60 -2.2 0 1.39 0 0 2.12 0 0 1.88 3.95 0 2.31 1 0 1.96 -1.5 1 1.93 -1.5 1 1.39 0.75 1 2.10 0.75 1 1.91 3.95 1 2.31 1 1 1.96

Table J.3 Stability factor profile Go Cong, with soil improvement (sand)

Water level sea side [m MSL]


Stability factor Water level lake side [m MSL]


Stability factor

-2.2 0 1.76 -2.2 0 1.56 0 0 2.12 0 0 1.88 3.95 0 2.31 1 0 1.94 -1.5 1 1.88 -1.5 1 1.23 0.75 1 2.07 0.75 1 1.90 3.95 1 2.31 1 1 1.96

Figure J-8: Schematized dam. Water level at MSL -2.2 m


Figure J-9: Slip circle sea side. Profile Strong Vung Tau

with soil improvement: stability factor = 1.60

Figure J-10: Slip circle lake side. Profile Strong Vung Tau

with soil improvement: stability factor = 1.39

Figure J-11: Slip circle sea side. Profile Strong Vung Tau

without soil improvement: stability factor = 1.28

Figure J-12: Slip circle lake side. Profile Strong Vung Tau

without soil improvement: stability factor = 1.14

Project Flood Defence HCMC | APPENDIX MCA scores alternatives Page 103 / 140

MCA SCORES ALTERNATIVES K

Project Flood Defence HCMC | APPENDIX Shearing Page 104 / 140

SHEARING L

When the residual force on the dam due to the water height difference is larger than the maximum shear

force underneath the dam, lateral displacement of the dam (shearing) can occur, see Figure L-1.

+6.5 m MSL

+3.2 m MSL

+0.0 m MSL

71212.8 9.9 25 9.6

1:4

1:41:3

1:3

L

-1.0 m MSL

Friction

Sea

water

pressure

Basin water pressure

Figure L-1: Horizontal equilibrium dam

For the mechanism of shearing no special limit state is defined (Weijers & Tonneijck, 2009). Therefore

basic geotechnical relations are used in combination with safety factors from the Eurocode. The following

safety factor have been chosen for calculating the loss of stability, according to Eurocode 7: ,

, and . For the loading a factor of , because when load is

variable and for a favorable variable load.

For the construction to be safe the following relation should hold:

∫ (

( ) ( )

)

(L.1)

Where:

= the density of water, = 10 kN/m3

= the water depth at the sea side [m]

= the water depth in the basin [m]

= a safety factor on unfavourable loading, = 1.5 (unfavourable, for sea) or 1.0 (favourable, for

basin)

= a safety factor on favourable loading, = 1.0

= a safety factor on the cohesion, = 1.25

= a safety factor on the angle of internal friction, = 1.25

= the (effective) cohesion [kPa]

= the angle of internal friction [°]

= the effective vertical stress [kPa]

In the undrained case this simplifies to:

(L.2)

Project Flood Defence HCMC | APPENDIX Shearing Page 105 / 140

Where:

= A safety factor un the undrained shear strength, = 1.4 [-]

The most unfavourable loading condition occurs when the sea reaches the highest level and the water

level behind the dam is low, in combination with a dam founded on soft soil.

The following assumptions are made:

(highest water level without influence of waves, waves are assumed to

have no influence on lateral equilibrium of forces, because of their short duration)

Clay subsoil, , , (top layer from profile Weak Vung Tau)

The phreatic level in the whole dam is taken as the water level in the sea. This is a negative

assumption, as a higher water table results in less vertical effective stress, and thus in lower

shear forces under the dam

Dam profile: see Figure L-1

Calculation of the drained and undrained case as a function of the depth gives the graph as presented in

Figure L-2. It can be concluded that shearing will not pose a problem for the dam; the residual strength in

both the drained and the undrained case is always larger than 0 for relevant bottom depths.

Figure L-2: Loads and resistances as function of the bottom depth

Project Flood Defence HCMC | APPENDIX Micro instability and seepage Page 106 / 140

MICRO INSTABILITY AND SEEPAGE M

M.1 Introduction

Micro instability is caused by seepage water that reaches the inner slope of a dike. The smaller particles in

the dam can erode from water pressures on the inside. It could be a minor damage, but, if the build-up of

the pressure inside the dike body is big enough, the integrity of the whole construction is in jeopardy.

Micro instability in general can include:

Pushing off of the revetment (if the water pressure is too high). This can happen with a dam with

a permeable core and less permeable top layer.

Shearing of the inner slope revetment. This can happen with a dam with a permeable core and

less permeable top layer.

Shearing of the inner slope as a result of an increase in height of the phreatic line in the dam.

Washing out of soil particles from the inner slope, which might occur when the permeability of

the top layer is approximately equal to that of the core of the structure.

M.2 Seepage flow and chosen parameters

First the flow through the dam has to be calculated. This is done with hand calculations, because no

adequate software was available. The Dupuit formula is used. This formula can be used because vertical

flow can be neglected due to the gentle water table inclination. The formula has the following form

(Barends & Uffink):

(M.1)

Where:

= the total discharge though the dam, in [m3/s per m’]

= the permeability of the core, in [m/s]

= the water level on sea, measured from the bottom, in [m]

= the water level in the basin, measured from the bottom, in [m]

= the drainage length, in [m]

The following parameters are assumed:

Permeability of the sand core ( ). This value corresponds with an

average value for loosely packed sand and the maximum value for the permeability of moderate

sands.

Sea water level: +3 m MSL, and a tidal amplitude of 1 m. A high level is most dangerous. However

the phreatic line in the dam will not follow the water levels very quickly, this can be shown by

calculating the influence length √ ⁄ √ ( ) ⁄ ,

where is the permeability [m/s], the water depth above sea bottom [m], the period of the

wave [s] and the porosity of the material [-], see section 4.3.4 of the report (Barends & Uffink).

This means that at the toe of the tidal amplitude is left. For wind waves this

will even be shorter because of their short period. Therefore the influence of tide and waves on

the head at the inner slope will be neglected and only a storm surge (+2.2 m) and sea level rise

(+0.8 m) will be taken into account.


Basin water level: -1 m MSL. A low level is most critical, as it maximised the head difference, and

thus the discharge. This level is based on the scenario that the level in the basin is kept low to be

able to collect the river discharge during an expected storm on sea.

Dike profile: Figure M-1

MSL-1 m

+3 m

L = 83,5 m

a0

1:31:4

h2 ∙ slope /

(slope + 1)

+6 m

1:4 1:3

7

2512

x

Figure M-1: Assumed dike profile for seepage and micro instability calculation

With these values with Dupuit a total discharge of ( ) ( ) is

calculated.

A slighty more complicated hand calculation method which takes into account the height above the water

table where the water flows out of the dam at the inner slope gives ( ) and

(the height above the water table, see Figure M-1.

The maximum velocity in the dam is estimated with the use of a flow net, see Figure M-2. The total

discharge is divided into 8 streamlines.

Figure M-2: Flow net in dam with head difference of 4 m

At the inner slope the streamlines have a minimum height of approximately 0.3 m. This corresponds with

a specific discharge of

(

). This is the

highest velocity that will be reached in the dam with the assumed profile and head difference.


M.3 Hydraulic gradient at the toe

The hydraulic gradient at the toe is not known, and therefore estimated with the following formula

(Barends & Uffink). The small zone of outflow above the water table at the inner side is hereby neglected.

( ) √

(M.2)

Where:

( ) = the phreatic head at distance from the inner slope, see Figure M-1 [m]

= the phreatic head in the basin, = 4 m

= the phreatic head at sea, = 8 m

= the drainage length, in [m]

When we take the derivation of this formula with respect to an expression for the hydraulic gradient is

found:

( )

√

(M.3)

Calculation of the gradient at the location (at the inner slope) gives a hydraulic gradient

( ).

M.4 Micro instability calculations

Several criteria have to be fulfilled to ensure protection against micro instability for sand core dams with a

sandy inner slope with water at the inner side:

1. Loss of stability of the inner slope

2. Wash out: the transport of sand particles from the dam

These two mechanisms will be checked in the following sections

M.4.1 Loss of stability of the inner slope

The formula to calculate the safety against loss of stability of the inner slope is given by (Schiereck G. ,

2004):

( ) ( )

( ) ( ) (M.4)

Where:

= the slope angle [°]

= Volumetric mass of (wet) soil, = 2000 kg/m3

= Volumetric mass of water, = 1000 kg/m3

= the angle between the water flow and the horizontal .

= Hydraulic gradient, [-]

= the angle of repose (for course sands this is usually 30 – 35°)

When seepage takes place under water, the slope surface is an equipotential line and therefore the flow

will be perpendicular to the surface. So . Thus equation (M.4) simplifies to:


(M.5)

Which gives a minimum angle of repose of .

When seepage takes place above the water table (0.29 m) the slope surface is a flow line, and therefore

the flow will be parallel to the surface. So and . Then equation (M.4) simplifies to:

(M.6)

Which gives a minimum angle of repose of .

This calculated minimum angles are higher than the estimated angle of repose of 30°, so additional works

are needed, like for example:

1. Decreasing the slope angle. Calculation with equation (M.6) gives that a slope of 1:3.5 is

sufficient to prevent shearing. This has only to be applied below the point where the phreatic line

reaches the surface on the inner slope.

2. Applying a drain at the bottom of the dam on the inner side to reduce the hydraulic head and

thus the flow through the inner slope.

3. Application of an impermeable layer on inner of outer slope, to reduce the hydraulic gradient.

Increasing the slope seems to be the easiest method, as the slope has only to be a little gentler.

M.4.2 Wash out

The formula to calculate the safety against wash out of the inner slope below the water table is given by

(Weijers & Tonneijck, 2009):

(M.7)

Where:

= the slope angle, = 18.4°

= total safety, = 2.0 [-]

= A safety factor for the angle of internal friction of the material, = 1.1 [-]

= Volumetric mass of soil, = 2000 kg/m3

= Volumetric mass of water, = 1000 kg/m3

= Hydraulic gradient, = 0.072 [-]

Calculating the minimum slope angle required to prevent wash out with formula (M.7) gives . So a

slope of 1:3 is more than sufficient to prevent washout below the water table.

The test for safety against washout in dikes with a sandy inner slope above the water is given by:

√

(M.8)

Where:

= Total safety, = 2

Calculating the maximum angle with these values gives , so a slope of 1:3 is safe.


M.5 Conclusion

In conclusion it can be said that the dam is safe against micro-instability if the inner slope angle is

decreased to a slope of 1:3.5 below the point where the phreatic line surfaces on the inner slope.

Although the dam will be quite permeable because is consists mainly of sand, no flush out of the particles

will occur.

Project Flood Defence HCMC | APPENDIX Slope revetment Page 111 / 140

SLOPE REVETMENT N

This appendix supports the section 4.3.6.1 of the report.

N.1 First estimate stone size: Input Parameters

An explanation is given on the input parameters of the equation to make a first estimate of the stone size

required for the armour layer.

N.1.1 Stability number,

The stability number depends on the demanded stability of the structure. A stability number smaller than

1 will be chosen for fixed structures where no damage is allowed, like caissons or seawalls. A large

stability number, from 6 to 20, can be chosen for dynamic rock slopes, where the profile of the dam with

small diameter armour stones is developed by the waves.

The dam structure to be built with a rock armour layer is judged to be a statically stable structure. A

statically stable structure is a structure where no or minor damage (stone displacement) to the armour

layer is allowed under design conditions. The stability number of a statically stable structure lies in the

range of 1 to 4.

1 to 4 - Stability number

N.1.2 Significant wave height,

The significant wave height at the toe of the dike is used. This value is determined in Appendix D.

3.5 m Significant wave height

N.1.3 Relative buoyant density,

The relative buoyant density depends on the density of the water and the density of the rock material.

(N.1)

2600 kg/m3 Average rock density

1025 kg/m3 Density of salt sea water

1.5 - Relative buoyant density

N.2 Van der Meer formula: Input Parameters

To apply the van der Meer formula, first a number of parameters have to be determined.

N.2.1 Design wave conditions, and , at the toe of the

structure

The 2 per cent wave height is taken instead of the significant wave height, because the shape of the

spectrum changes and the wave itself becomes more peaked and skewed towards the shore.

The influence of the shape of the wave energy spectra is taken into account by using the spectral wave

period, , instead of the mean wave period from time-domain analysis, .


The values of and are obtained using SWAN with the offshore wave conditions = 10 m

and = 12 s.

m 4.6 s 8.0

N.2.2 Acceptable value of damage level parameter,

The limits of the value of depend mainly on the slope angle of the structure. For armour stone in a

double layer the values in Table N.1 can be used.

Table N.1: Design values of the damage parameter, , for armour stone in a double layer

Slope (cot α) Damage level

Start of damage Intermediate damage Failure

1.5 2 3-5 8 2 2 4-6 8 3 2 6-9 12 4 3 8-12 17 6 3 8-12 17

Although a value of the damage level parameter of = 2 to 3 is often used for design purposes, in some

cases it might be a feasible approach to apply higher values of = 4 to 5. A value of the damage level of

= 4 is chosen.

N.2.3 Number of waves,

The number of waves in a storm can easily be computed from information about the wave period and the

duration of the storm.

(N.2)

- Number of waves h Duration of storm s Wave period s/h Factor, number of seconds in one hour

In the range of validity of parameters in the Van der Meer formulae for shallow water conditions the

number of waves has a maximum value of 3000. With a wave period of 8.0 s, this number of waves

originates from a storm with a duration of 6.7 h. This is a reasonable duration of the time the maximum

storm conditions last. The value of = 3000 is taken for the design of the armour layer of the revetment.

N.2.4 Notional permeability,

The notional permeability of the structure is not defined in a standard way, by using Darcy’s law, but is

rather given as a notional index that represents the global permeability of the structure, or as a ratio of

stone sizes. The notional permeability depends on the size of the filter layers and core.

Van der Meer has the following classification of notional permeability, as given in Figure N-1.


Figure N-1: Notional permeability for different layer building (Van der Meer, 1987)

From this classification can be deduced that the dam body with a rock armour layer, an under layer and

several filter layers will be classified as a structure with a notional permeability of = 0.4.

N.2.5 Surf similarity parameter,

The surf similarity parameter depends on the wave parameters, and , and the slope angle

( ).

√

(N.3)

- Surf similarity parameter - Slope Dam profile - Fictitious wave steepness based on Tm-1.0

The dam profile in the first design will have a slope of 1:4, due to stability reasons.

The fictitious wave steepness based on is the quotient of the water height and the wave length:

(N.4)


4.6 m 2 per cent wave height at toe of structure 10 m/s

2 Gravitational acceleration

8.0 s Spectral wave period at the toe of the structure

0.045 - Fictitious wave steepness based on Tm-1,0

With this calculation, the surf similarity parameter can be calculated:

1:4 - Slope Dam profile 0.045 - Fictitious wave steepness based on Tm-1.0

1.2 - Surf similarity parameter

N.2.6 Critical surf parameter,

The critical surf parameter will be used to compare with the surf similarity parameter. From this

comparison the wave conditions at the toe of the dam can be deduced: plunging or surging.

[

√ ]

8.7 - Plunging coefficient, shallow water calculation

1.4 - Surging coefficient, shallow water calculation 0.4 - Notional permeability

1:4 - Slope of structure

2.6 - Critical value of the surf similarity parameter

N.3 Inner slope protection

N.3.1 Input parameters

The fictitious run-up,

The fictitious run-up is the run-up that would take place on an infinite slope. In this case for 2% of the

waves. The 2 per cent run-up level is determined in section appendices E and F as = SWL +3.1 m.

Crest level relative to still water level at the seaward side of the crest,

The crest level will be constructed at MSL +6.8 m. In this value the Sea Level Rise of 0.75 m and

settlements of 1.0 m are taken into account, as well as a value of 0.25 m for wave surge. The design still

water level is situated at MSL +3.2 m. This value plus the Sea Level Rise and the wave surge leads to an

elevated still water level of MSL +4.2 m. The crest height will subside during time with an estimated total

value (crest and subsoil) of 1 m. This will result in a crest with a height at MSL +5.8 m.

The height of the crest above the still water level thus results in 1.6 m.

Crest width,

The crest width of the dam has been assumed to be 7 m. This is the chosen width of the armour layer at

the crest.

Wave length in deep water,

With the period of deep water waves known, the wave length in deep water can be calculated:

(N.5)

With an offshore wave period of 12 seconds, the deep water wave length is equal to = 225 m.


Coefficients

Empirical coefficients on the crest for regular waves:

0.81 - Coefficient for regular waves

15 - Coefficient for regular waves 0.042 - Coefficient for regular waves

Coefficients based on the slope angle are calculated by:

(N.6)

(N.7)

With a slope of 1:4, this leads to:

0.15 - Coefficient for outer slope 1:4

1.24 - Coefficient for outer slope 1:4

Friction factors

Two types of friction factors are used in the calculations: and . The value of the -factor is the

roughness factor used in this type of equations, with a value of 1 for a smooth surface. For a double rock

layer the value of this roughness coefficient is known to be 0.55. What the value of the other type of

friction coefficient, , should be for this design is not clear. This friction coefficient has a value of 0.02 for

smooth coverage and grass (Bosman, Van der Meer, Hoffmans, Schuttrumpf, & Verhagen, 2008). For this

design with rocks a value of 0.1 is assumed.

N.3.2 Characteristic stone size

For the calculation of the characteristic stone size, , first the characteristic velocity of the water on

the inner slope ( ) is calculated:

(

) (N.8)

The values of , the velocity at the seaward side of the crest and , the layer thickness of the water

on the crest, are calculated below:

(

) (

) (N.9)

3.5 m Significant wave height 0.15 - Coefficient for outer slope 1:4

3.1 m Run up height above SWL 1.6 m Crest height above SWL 0.55 - Friction factor crest (1 is smooth surface) (rock armour)

0.81 - Coefficient for regular waves

15 - Coefficient for regular waves

7 m Crest width 225 m Wave length in deep water

0.14 m layer thickness of water on crest


According to (Bosman, Van der Meer, Hoffmans, Schuttrumpf, & Verhagen, 2008) the velocity on the

outer slope can be calculated by:

√

√

(N.10)

14 ˚ Slope angle outer slope 1.24 - Coefficient for outer slope 1:4

3.1 m Run up height above SWL 1.6 m Crest height above SWL 10 m/s


3.5 m Significant wave height

4.8 m/s velocity outer slope

With the outcomes of these calculations, the velocity on the inner slope can be calculated.

(

) (N.11)

4.8 m/s Velocity on seaward side of the crest (calculated) 0.042 - Coefficient for regular waves

7 m Crest width 0.1 - Estimated friction factor crest (0.02 is smooth surface) (rock

armour) 0.14 m Layer thickness of water on the crest (calculated)

3.9 m/s velocity on landward side of the crest (outcome this calculation)

The calculation of the characteristic stone size necessary for the armour of the inner slope, and calculated

by:

(N.12)

3.9 m/s , elocity on landward side of crest 1.5 - Relative buoyant density 10 m/s


0.34 m Characteristic stone size inner slope

Project Flood Defence HCMC | APPENDIX Toe protection for quarry stone revetment Page 117 / 140

TOE PROTECTION FOR QUARRY STONE REVETMENT O

O.1 General formula

The toe protection provides a stable footing to the armour layer. The stone size that is necessary for the

toe protection can be determined with a formula that was obtained from (Schiereck G. J., 2001):

(

)

(O.1)

Symbol Unit Definition Calculation value

m - m Significant wave height 3.2 m Water depth at structure toe 2.3 m Water depth in front of toe 6.7 - Specific weight of the stones 1.5

O.1.1 Relative buoyant density

The relative buoyant density depends on the density of the water and the density of the rock material.

(O.2)


kg/m3 Average rock density 2600

kg/m3 Density of salt sea water 1025

- Relative buoyant density 1.5

O.1.2 Water depth in front of the toe

For the water depth in front of the toe, the design still water level (MSL +3.2 m) is used. The bottom is

assumed to be located at MSL -2.5 m. 0.75 m of sea level rise is taken into account, as well as 0.25 m of

wave surge elevation. The water depth is therefore 6.7 m.

(O.3)

In the table below the parameters are explained and the calculated values are given:

Parameter Unit Definition Calculated

m Water depth in front of toe 6.7 m Tidal elevation of the Still Water Level 1.0 m Storm surge elevation of the Still Water Level 2.2 m Wave surge 0.25 m Sea Level Rise 0.75 m Bottom level below MSL 2.5

Project Flood Defence HCMC | APPENDIX Toe protection for quarry stone revetment Page 118 / 140

O.1.3 Water depth at structure toe

O.1.3.1 With sublayers

The toe supports the armour layer and the under layer. If the toe is constructed as in Figure O-1, the

height of the toe is 4.4m above the sea bottom. The water depth at the toe is therefore assumed to be 6.7

- 4.4 = 2.3 m.

Figure O-1: Toe protection

O.1.3.2 Thin sublayers

The sub layers may be smaller in reality, if a geometrically open filter or a geotextile is applied. In a

situation without sublayers it is assumed that the toe has a height of 2.4 m (approximately the thickness

of the armour- and underlayer together). The water depth at the toe is therefore assumed to be 6.7 - 2.3

= 4.4 m for this case.

O.2 Results

O.2.1 With sublayers

Stones with a of 1.1 m were calculated with formula (O.1). This stone size is larger than that of the

stones which are used for the outer slope revetment. Because the toe is relatively high above the bed, the

toe acts more like a berm. Calculation should be done similarly to the outer slope revetment. The same

stone size as on the outer slope revetment is an estimate for the toe ( ).

O.2.2 Thin sublayers

For the second case with very thin sublayers, stones with a of 0.44 m are necessary. Because these

are smaller than the stones that are used for the outer slope revetment layer, the stability has to be

checked.

O.3 Conclusion

For the final design the situation will probably be somewhere in between the case with the sublayers as in

Figure O-1 and the situation without sublayers.

In a normal hydraulic situation (water level at MSL), the water depth at the toe is low compared to the

depth in front of the toe. The toe acts like a berm and the same calculation method as for the outer slope

protection is advised. The same stone size as for the outer slope revetment is advised for a preliminary

design of the toe.

Project Flood Defence HCMC | APPENDIX Model for tidal prediction Page 119 / 140

MODEL FOR TIDAL PREDICTION P

In section 5.3 of the report, the storage area approach model is used to calculate the expected velocities

in the closure gap. This model needs input for the storage area, the river discharge, and most important,

for the tide. Harmonic analysis is used in a computer model (Texcel, written by professor Schiereck of TU

Delft) to predict the tide. The harmonic analysis itself will only briefly be discussed in this appendix. For

the full theory, the reader is referred to the textbooks.

In this section, the performance of the model is tested by comparing the computed values with:

Values predicted by the British Admiralty Chart

Measured water levels at Vung Tau

Based on the outcome is evaluated whether the model is sufficiently accurate for this stage of the design.

P.1 Harmonic model

The principle of tidal analysis is that tidal oscillations can be separated into multiple sinusoids. Each

sinusoid has its own amplitude and frequency, which is determined by the force which drives it. The

purpose of tidal analysis is to determine the amplitude and phase of these individual cosine waves. When

the tidal harmonic constants are known, the tide can be predicted. For many places around the world,

these values are known. For instance, the British Admiralty Chart (Admiralty Chart) provides the following

values, see Table P.1. These are the tidal constituents the predictive model uses. Other constituents are

calculated according to the equilibrium theory, which implies that certain constituents are related to

others.

Table P.1: Main tidal constituents at Vung Tau

Tidal constituent Description Nature Amplitude (m) Phase (⁰)

M2 Principal lunar Semi-diurnal 0.79 36 S2 Principal solar Semi-diurnal 0.31 81 K1 Principal declination Diurnal 0.6 312 O1 Principal lunar Diurnal 0.45 263


P.2 Comparison with Admiralty Chart

For comparison, the predicted tide for the period of October 5th

– 11th

is obtained from the website of the

British Admiralty Chart. These charts are used all over the world for navigation. These tidal predictions are

also based on tidal analysis. The results of this comparison are given in Figure P-1.

Figure P-1: Comparison of tidal predictions

From Figure P-1 it can be seen that the harmonic model shows a good fit with the predicted water levels

from the Admiralty tide. The moments of HW and LW seem to fit perfectly, and the differences in water

levels at HW and LW lie in the order of 0.1 m.


P.3 Comparison with measured data

For this comparison, measured water levels at Vung Tau were compared with predictions from the

harmonic model. Figure P-2 shows the results. The moment of HW and LW seems to fit reasonably well.

The model overestimates the tidal range. The predicted water level at HW and LW has a maximum

deviation from measured values in the order of 0.2 m.

Figure P-2: Comparison of computed and measured water levels at Vung Tau

P.4 Conclusions

It can be concluded that the used harmonic model gives results which are comparable with the widely

used Admiralty Charts. This gives enough confidence to use it for the modelling purposes it has been

intended for. The model fits reasonably well with the measured data. The difference in water level

between the two might be caused by other influences than just tidal fluctuations (like meteorological

influences, i.e. storm surges) or by shortcomings in the model. If more tidal constituents were to be taken

into account, the model might have shown a better fit with the measured data. For the purpose of

modelling however, the used harmonic model is satisfactory.

Project Flood Defence HCMC | APPENDIX Determination of grain sizes for the closure Page 122 / 140

DETERMINATION OF GRAIN SIZES FOR THE Q

CLOSURE

In this appendix the required grain size of the rubble, which is used for the closure, is calculated. A full

horizontal closure is assumed.

Q.1 General formula

The stability of rubble in horizontal flow can be described with a Shields relation:

(Q.1)

Q.1.1 Damage parameter

For the damage parameter a value of is recommended (Konter, Jorissen, & Klatter, 1992),

which corresponds with little transport.

Q.1.2 Relative buoyant density

The relative buoyant density is calculated with:

(Q.2)


- Relative buoyant density 1.5

kg/m3 , apparent mass density revetment material (rock) 2600

kg/m3 mass density water 1025

A mass density of 2600 kg/m3 is assumed for the rock material.

Q.2 Area of most damage

During horizontal construction of the final gap only the ends of the dam are loaded by a hydraulic force.

The upstream side is normative.


m Nominal diameter grains (See Q.3) m/s Vertically averaged flow velocity Depends on width closure gap (see Q.3) - Damage parameter 0.04 (see Q.1.1) - Relative density rubble 1.5 (Q.1.2) m

0.5/s Roughness coefficient Depends on Dn (see Q.2.3)

- Load increase factor 1.2 (see Q.2.2) - Slope correction factor 0.63 (see Q.2 )


Figure Q-1: Horizontal closure (Konter, Jorissen, & Klatter, 1992)

In model research (Naylor) is determined that most damage appears at the head of the dam, in the

shaded area in Figure Q-2. For this location, the required stone size is determined in Q.3.

Figure Q-2: Area most vulnerable to damage in horizontal closure (Konter, Jorissen, & Klatter, 1992)

Q.2.1 Slope correction factor

A slope correction factor is included in the Shields relation, because grains on a slope are less stable than

grains on a horizontal plane. The slope correction factor for shaded area can be calculated with:

√

(Q.3)


- Slope correction factor 0.63 ° Slope angle 30 ° Angle of repose 40

Values for the slope angle and angle of repose were found in literature (Konter, Jorissen, & Klatter, 1992).

These are common for horizontal closures with rubble.

Q.2.2 Load increase factor

In (Schiereck G. J., 2001) a load increase factor of 1.2 was advised for round abutments.


Q.2.3 Roughness parameter

For the roughness parameter of the area of most damage, the expression of White-Colebrook is used:

(

) (

) (Q.4)

For the roughness height a value of is used.

The water depth ( ) at the area of most damage depends on the overflow situation. In a sub-critical flow

situation it holds that the water depth at the gap is equal to the downstream water depth. In the storage

area approach calculation (Texcel), the average water depth at the time that maximum velocities occurred

is found to be MSL -0.86 m.

Most damage seems to appear at half the water depth (Schiereck G. J., 2001).

( )

m is used for calculations.

Q.3 Grain size closure

In this section the grain sizes that are necessary in several construction phases are calculated.

Q.3.1 Width final gap

In section 5.3 of the report the velocities that can occur in the final gap are calculated. For a discharge

sluice area of 20000 m² the results are displayed in Figure Q-3. It is assumed that sand closure can be

applied up to a velocity of around 2.5 m/s before too much sand is washed away (Verhagen, d'Agremond,

& van Roode, 2009).

For the VT-GC dam the depth of the final gap is the bottom depth of the Soi Rap navigation channel,

which is located on MSL -9.5 m. The maximum flow velocity is assumed to take place at ebb. The graph for

the storage area approach is made for an assumed depth of 10 m. A speed of 2.5 m/s is found for a

discharge area of m2. The same speed for a depth of 9.5 m will be reached at a width

of m. From a final gap width of 1370 m on bigger grain size than the size of sand have

to be used.

In reality the final gap is probably even wider than 1370 m, because the bottom of the navigation channel

is not located at MSL -9.5 m over the full length of 1370 m. The exact geometry of the final gap area has

to be determined in a later stage of the design.


Figure Q-3: Velocity and discharges with discharge sluice area of 20000 m2

Q.3.2 Grading classes final closure

Using Figure Q-3 and the ratio between the water depth and the final gap area the velocities are

calculated for several gap widths in steps of 300 m. For each step the velocity at the end of the step is

used for calculation (e.g. in phase T0 the velocity for a gap width of 1070 m is used). In section 6.3.3 of the

report a construction speed of 300 m/2 weeks is estimated. In Table Q.1 the results of the calculation of

the required stone class during different phases of the final gap closure are displayed. Formula (Q.1) is

used for these calculations.

Table Q.1: Stone size calculation final gap

Phase Wgap (m) (m/s) (m) Stone class

T0 1370 - 1070 2.6 0.11 80 / 200 mm T1 1070 - 770 3.0 0.17 200 / 350 mm T2 770 - 470 3.3 0.23 10 - 60 kg T3 470 - 170 3.5 0.28 60 - 300 kg T4 170 - 0 4.0 0.44 300 - 1000 kg

With Figure Q-4 and Figure Q-5 the required grading class was determined for each phase of the closure.


Figure Q-4: Size gradings and relationships for the standard fine and light grading classes (Schiereck G. J., 2001)

Figure Q-5: Weight gradings and size relations for the standard light and heavy grading classes (Schiereck G. J., 2001)

Project Flood Defence HCMC | APPENDIX Bottom protection Page 127 / 140

BOTTOM PROTECTION R

A bottom protection is constructed to make sure that a scour hole does not develop too close to the dam.

In this appendix the required stone size for and length of the bottom protection near the final gap are

calculated in case of a full horizontal closure.

R.1 Required stone size bottom protection

For horizontal closure, little information is available on the stability of the top layer of a bottom

protection. To calculate the required stone size for the bottom protection, the Shields relation is used as

was recommended by literature (Konter, Jorissen, & Klatter, 1992).

(R.1)

Compared to the calculation in appendix Q, only the slope correction factor is different. Because the

bottom is assumed horizontal, the slope correction factor for the bottom protection is equal to one.

For the roughness parameter of the area of most damage, the expression of White-Colebrook with is

again used:

(

) (R.2)

The water depth changes during a tidal cycle. For calculations the average water depth at the time that

maximum velocities occurred was. Using the storage area calculation (Texcel) this water depth was

determined: = 8.6 m.

In Table R.1 the results of the calculation for the required stone sizes for the bottom protection near the

final gap are displayed.

Table R.1: Stone size calculation bottom protection final gap

Phase Wgap (m) (m/s) (m)

T0 1370 - 1070 2.6 0.057 T1 1070 - 770 3.0 0.086 T2 770 - 470 3.3 0.11 T3 470 - 170 3.5 0.14 T4 170 - 0 4.0 0.21


m Nominal diameter grains m/s Vertically averaged flow velocity Depends on width closure gap - Damage parameter 0.04 - Relative density rubble 1.5 m

0.5/s Roughness coefficient Depends on Dn

- Load increase factor 1.2 - Slope correction factor 1


R.2 Layer thickness

When the diameter of the original bed material and that of the necessary protection is not too large, one

layer could be sufficient. However, construction of only one layer is not possible. Therefore layers of a few

decimetres are usually applied. (Schiereck G. J., 2001) In (CIRIA, 2007) a layer thickness of 0.5 m is

mentioned as a typical minimum.

R.3 Required bottom protection length

The depth of the scour hole and the bottom material determine how far away from the dam the hole

must be, and therefore the length of the bottom protection. The length of the bottom protection has

again influence on the final depth of the scour hole. Therefore determination of the length is an iterative

process. In this section one iteration will be executed.

Figure R-1: The bottom protection length depends on scour depth and slope angle of the scour hole (Konter, Jorissen,

& Klatter, 1992)

From several hundreds of tests and experimentally established relations an expression for scour behind a

bed protection was derived for the development phase of the scouring process:

( ) ( )

(R.3)


( ) m Maximum depth in the scour hole as a function of time

- Dustbin factor; represents amongst others turbulence and locally increased velocity

m/s Vertically averaged velocity at the end of the protection

m/s Critical velocity m Downstream water depth - Specific weight of the sediment 1.6 h Time

10 m0.9

/s1.3

Non-dimensionless constant 10


The power 0.4 in equation (R.3) holds strictly speaking for 2-dimensional flow situations. For a 3-

dimensional flow situation (horizontal closure) the power is higher in the beginning and a little bit smaller

later on. Because 0.4 is a good first estimate it will be used for the preliminary design of the VT-GC dam.

R.3.1 Downstream water depth

In the storage area calculation (Texcel), the average water depth at the time that maximum velocities

occurred is found to be MSL -0.86 m. = 9.5 - 0.86 ≈ 8.6 m is used as a calculation value.

R.3.2 Critical velocity

The critical velocity for cohesive sediments was given in (CIRIA, 2007).

Table R.2: critical velocity for cohesive sediments

Material Critical velocity [m/s]

Fairly compacted clay (void ratio, =0.50) 0.8 Stiff clay (void ratio, = 0.25) 1.5

Sand with a grain size of 0.3mm will be used for soil improvements. Because the soil improvement

extends a few hundreds of meters beyond the dam (see section 4.2.2 of the report) it is assumed that the

first 8 m of the bottom consist of sand.

For sand, the critical velocity is estimated using a Shields relation:

√

√ (R.4)

The specific weight of the sand is determined with:

(R.5)


- Relative buoyant density 1.6 kg/m

3 ρapp , apparent mass density revetment material (rock) 2650

kg/m3 mass density water 1025

For the roughness parameter of the area of most damage, the expression of White-Colebrook is used:

(

) (R.6)

For the downstream water depth of 8.6 m is used. This value applies when the highest velocities appear

(see R.3.1).


m/s Critical velocity 0.4

m Nominal diameter grains 0.0003

- Damage parameter 0.04

- Relative density 1.6 m

0.5/s Roughness coefficient 94


A critical velocity of 0.4 m/s is found for the first 8 m of the bottom, where soil improvement is applied.

Below that level, the bottom is assumed to consist of clayey sand up to a 25 m below the bottom. The

critical velocity for clay is higher than for sand. A calculation value of 0.4 m/s is used for the critical

velocity because the ratio between clay and sand is unknown.

The critical velocity of 0.4 m/s gives a first approximation of the erosion resistance. It is recommended to

check the estimated critical velocity in a laboratory or to construct a test section.

R.3.3 Factor alpha

R.3.3.1 Horizontal constriction

Factor alpha is determined by the geometry of the dam and turbulence influences. Constrictions, the

length of the bottom protection and the roughness of the bottom protection determine factor alpha.

In Figure R-3 values for alpha are given as function of the horizontal constriction. This graph holds for

L/h3=10.

When determining alpha 1370 m was used for . For this gap width the water level in the basin could

still follow the water level at sea.

Figure R-2: Horizontal constriction

Figure R-3: Values for αL as a function of horizontal constriction for L/h3=10 (Konter, Jorissen, & Klatter, 1992)

R.3.3.2 Length bottom protection

A longer bottom protection leads to lower alpha-values due to dissipation of turbulence.

The -value as a function of relative bottom protection length can be calculated with:

(

) ( ) (

) (R.7)


Symbol Unit Definition

( ) - α-value as a function of relative bottom protection length - α-value for L/h3=10 m Bottom protection length m Downstream water depth

R.3.3.3 Roughness of the bottom protection

The roughness of the bottom protection is also of influence for the -value. For a smooth protection the

velocities near the bed are higher and cause more scour.

R.3.4 Gradually varying flow intensity

For a gradually varying flow intensity the scour depth can be calculated with:

( )

∫ ( )

(R.8)

In Texcel velocities in the final gap were calculated during an average tidal period. Every ten minutes

( ) was calculated for a reference period from 13-09-2011 to 13-10-2011. The average

( ) was calculated for all negative (inflow) velocities. The same was done for all positive

(outflow) velocities to determine the scour depth at the sea side of the barrier. Factor alpha was

determined using Figure R-3 for the lower boundary of each gap width (e.g. for 1070 - 770m α was

calculated for a width of 770 m). Results are displayed in Table R.3.

Table R.3: Flow velocities for in- and outflow for several gap closure stages

Phase Wgap (m)

∫ ( )

Inflow

∫ ( )

Outflow

T0 1370-1070 2.1 1.5 2.4 T1 1070-770 2.7 3.0 4.8 T2 770-470 3.4 5.9 9.3 T3 470-170 3.8 9.2 14.9 T4 170-10 4.2 13.0 20.8

Graphs for the development of the scour hole are calculated using formula (R.8). Results can be seen in

Figure R-4 and Figure R-5.


Figure R-4: Development scour hole due to outflow

Figure R-5: Development scour hole due to inflow

Using these graphs the scour depth at the end of the construction period can be found. Results are:

Scour depth due to outflow (on the land side): 16 m

Scour depth due to inflow (on the sea side): 25 m

R.4 Bottom protection length

The bottom protection must prevent the structure from sliding into the scour hole. The bottom protection

must therefore keep the scour hole at a certain distance from the structure and have a certain length. If a

slide occurs the resulting slope after sliding may not reach the structure itself.


The required length of the bottom protection is determined using the maximum depth of the scour hole

( in Figure R-6) and the slope angle ( in Figure R-6)

Figure R-6: Instability due to bottom erosion (Konter, Jorissen, & Klatter, 1992)

A safe design rule is described in (Konter, Jorissen, & Klatter, 1992):

( ( ) ) (R.9)


m Required length bottom protection Land side: 32m Sea side: 50m

m Depth scour hole Land side: 16m Sea side: 25m

( ) - Factor for slope angle after instability 8 or 20

For the slope angle before instability 1:4 is used in this formula. Steeper slopes can occur downstream of a

horizontal closure, but will not be normative.

The angle of the slope after instability is determined by the bottom material. In (Konter, Jorissen, &

Klatter, 1992) a value of ( ) = 8 is given for sand that is not vulnerable to settlement flow, and a value

of ( ) = 20 for sand that is vulnerable to settlement flow. For clay holds that ( ) = 2.

The bottom consists of sand and clayey sand. The slope angle after instability will lie between ( ) = 2

and ( ) = 8, but closer to ( ) = 8. A value of ( ) = 8 is used for calculations. This results in a

required bottom protection length of 32 m at the land side and 50 m at the sea side.

If a value of ( ) = 20 is used, the required bottom protection length is 128 m at the land side and 200

m at the sea side.

A shorter bottom protection results in a higher value for (see R.3.3.2), resulting in a larger final depth of

the scour hole. Iterations have to be done to calculate the required length of the bottom protection.

The final length of the bottom protection will lie between 32 and 86 m on the land side and between 50

and 86 m on the sea side. The final length will probably be closer to the length that is calculated with

formulas. Therefore the required bottom protection results in:

Land side: 45 m of bottom protection

Sea side: 60 m of bottom protection


For soil that is vulnerable to settlement flow the final length of the bottom protection will lie between 86

and 128 m on the land side and between 86 and 200 m on the sea side. The final length will probably be

closer to the length that is calculated with formulas. Therefore the required bottom protection results in:

Land side: 110 m of bottom protection

Sea side: 180 m of bottom protection

R.5 Velocity

The velocity decreases when a bottom protection is longer. Stone sizes at the end of the bottom

protection may therefore be shorter than in the gap.

The velocity ( ) at the end of the bottom protection follows from the length of the bottom protection

and can in the case of a horizontal closure be determined from the Figure R-7.

Figure R-7: Velocity versus length of bottom protection (Konter, Jorissen, & Klatter, 1992)

Symbol Unit Definition

Lbb m Length bottom protection h3 m Difference between downstream water depth and constriction height U m/s Velocity at the end of the bottom protection U0 m/s Velocity in the final gap Bs m With of final gap Bso m Initial gap width

For the land side it holds that = 5.2. And for the sea side it holds that: = 7.0.

From Figure R-7 can be seen that the velocity decrease at the end of the bottom protection only plays a

role on the sea side, but the reduction won’t be very large.

Project Flood Defence HCMC | APPENDIX Unsteady state salt intrusion models Page 135 / 140

UNSTEADY STATE SALT INTRUSION MODELS S

A lot of research has been carried out on modelling of salt intrusion in estuaries. There are roughly two

different types of models which are widely used: numerical models and analytical models. Numerical

(unsteady-state) models are popular because they can provide more spatial and temporal detail (Zhang,

Savenije, Wu, Kong, & Zhu, 2011). The movement of water and salt is generally described by coupling four

one-dimensional equations: the conservation of momentum and mass, the salt balance and a state

equation (Savenije, 2005):

Momentum:

( )

| |

(S.1)

Continuity:

(S.2)

The salt balance equation:

(

) (S.3)

And the state equation:

(S.4)

In which:

Q = Q(x,t) is the discharge in m3/s;

αs is a shape factor (assumed constant) to account for the spatial variation of the flow velocity

over the cross-section (αs > 1);

A = A(x,t) is the cross-sectional area of flow in m2;

h = h(x,t) is the mean cross-sectional depth of flow in m;

Zb = Zb(x,t) is the mean cross-sectional bottom elevation in m;

g is the acceleration due to gravity in m/s2;

ρ = ρ(x,t) is the density of the fluid in kg/m3;

U = U(x,t) is the mean cross-sectional flow velocity in m/s;

C = C(x,t) is the coefficient of Chézy in m0.5

/s;

B = B(x,t) is the stream width of the channel in m;

Bs = Bs(x,t) is the storage width of the channel in m;

rs = rs(x) is the ratio of storage width Bs to stream width B (rs > 1);

Rs is a source term, accounts for rainfall, evaporation, or lateral inflow in m2/s;

s is the mass concentration, or in this case the salinity in kg/m3;

Equations (S.1) and (S.2) together are generally known as the St. Venant equations. These equations

demands a numerical approach since the set of partial differential equations cannot be solved analytically.

The calculations can be carried out by hydrodynamic models such as MIKE11 and SOBEK-RE. These models

solve the before mentioned system of equations by calculating first the distribution of water level,

velocity and discharge at a certain computed time and specific cross-section, and then solving the salt

Project Flood Defence HCMC | APPENDIX Unsteady state salt intrusion models Page 136 / 140

distribution. MIKE11 solves the differential equations by using a finite differences scheme known as the

Abbott-Ionescu scheme (Parsa, Etemad-Shahidi, Hosseiny, & Yeganeh-Bakhtiary, 2007).

A disadvantage of all 1-D unsteady-state models is that they use the dispersion coefficient as a

calibration coefficient, without making use of a model that describes the dispersion as a function of

geometry, discharge and tide (Nguyen, 2008). Therefore, the predictive capacity of these models is not

very good. A model that uses a predictive equation for is the one of Savenije (Savenije, 2005).

The hydrodynamic models need initial and boundary conditions regarding tidal motion, river flow and

salinity. In general, these models use the upstream discharge as the upstream input, and the tidal

variation as the downstream boundary. For the salinity, the salinity gradient is taken as on

the upstream boundary. The upstream salinity is often set equal to the river salinity. The downstream

boundary condition is more difficult because of varying salinity during a tidal cycle. Often the salinity is set

equal to the salinity at sea during flood. During ebb, the upstream conditions determine the salinity at the

downstream boundary. At the end of ebb, the salt concentration is lower than the salt concentration at

sea. The method of Thatcher and Harleman can be used to determine the transition period for the

boundary condition when the condition changes from water leaving to water entering the model (RIZA,

2005). This period depends on the sea conditions. Initial conditions for tidal motion, river flow and salinity

need to be specified.

Project Flood Defence HCMC | APPENDIX Model and Measurements Page 137 / 140

MODEL AND MEASUREMENTS T

T.1 Run 1 September - Boundary conditions

The upstream boundary conditions are discharges. The discharges for this run are obtained by averaging

the discharges of the last 10 years in September, as the discharge data of 2011 is not available.

Table T.1: Upstream boundary conditions: Discharges

Day Vam Co Tay [m

3/day]

Vam Co Dong [m

3/day]

Dau Tieng [m

3/day]

Tri An and Phuoc Hoa [m

3/day]

1 6.6179 83.2668 11.2 1341 2 5.6911 75.9903 11.2 1320.4 3 5.1716 71.0564 11 1302.4 4 4.8939 67.9846 25.2 1253.7 5 4.6274 66.7159 31 1243.5 6 4.4976 66.2239 31.2 1237.8 7 3.8953 66.1511 22.9 1242.1 8 3.7577 68.9209 11.9 1239.4 9 3.5287 67.267 11.9 1265 10 4.3454 68.1221 11.9 1318 11 4.1193 67.5206 11.9 1219.1 12 6.6338 73.1649 11.9 1259.4 13 10.7935 202.9265 11.9 1272.3 14 32.6041 146.4034 12.7 1304.5 15 19.5521 115.7372 13.3 1338.9 16 12.975 102.7936 13.3 1379.4 17 9.4003 92.3404 13.3 1366 18 14.0689 86.3904 3 1412.6 19 9.9648 82.4335 3 1470.3 20 54.8633 80.9259 1.8 1431.7 21 26.4539 81.1602 3.4 1460.2 22 15.9759 87.6139 3.4 1488.7 23 20.846 83.8919 3.4 1420.1 24 12.1216 83.2044 3.4 1455.2 25 14.0076 88.8387 13.1 1549 26 26.2472 90.9495 12 1778.5 27 21.6242 91.5257 11.3 1852.8 28 16.1298 90.9158 4.1 1807.9 29 15.3443 109.6024 4.1 1899.2 30 20.186 157.3698 4.5 1981.4

The downstream boundary condition is the water level elevation. For this simulation only the tidal

elevation is taken into account, and not the elevation due to wind surges.


Figure T-1: Downstream boundary condition: Water level in Vung Tau

T.2 Run 1 September – Outcome

For the 1st

of September, the velocities in the Phu An reach are predicted with the SOBEK hydraulic model

of the Saigon - Dong Nai delta. The outcome of this simulation is listed in Table T.2.

Table T.2: Mean velocity in Phu An reach - prediction for 01-09-2011 with SOBEK model

Date & Time Velocity mean 38_7 [m/s]

1-09-11 0:00 0 1-09-11 1:00 7.87E-04 1-09-11 2:00 6.82E-02 1-09-11 3:00 0.4126168 1-09-11 4:00 0.6707854 1-09-11 5:00 0.4551059 1-09-11 6:00 -3.72E-02 1-09-11 7:00 -0.3406768 1-09-11 8:00 -0.4130285 1-09-11 9:00 -9.04 E-02 1-09-11 10:00 0.4265332 1-09-11 11:00 0.9417185 1-09-11 12:00 1.233679 1-09-11 13:00 1.238997 1-09-11 14:00 1.015514 1-09-11 15:00 0.4462772 1-09-11 16:00 -0.3247664 1-09-11 17:00 -0.9155796 1-09-11 18:00 -1.2172 1-09-11 19:00 -1.18124 1-09-11 20:00 -0.9298639 1-09-11 21:00 -0.4939313 1-09-11 22:00 0.1477329 1-09-11 23:00 0.7665752


Figure T-2: Mean velocity in Phu An reach - prediction for 01-09-2011 with SOBEK model

The prediction for a couple of days shows more information about the variation of the velocities in the

Phu An reach.

Figure T-3: Mean velocity in Phu An reach - prediction for first 7 days of September with SOBEK model

Project Flood Defence HCMC | APPENDIX Drawings Page 140 / 140

DRAWINGS U

The following drawings are presented on the following pages

1. A3-drawing of the dam with rock revetment

2. A3-drawing of the dam with asphalt revetment, where the toe protection is placed below sea

bottom level

3. A3-drawing of the dam with asphalt revetment, where the toe protection is placed above sea

bottom level

4. A3-drawing of the dam at the location of the final gap closure

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Vung Tau Go Cong Dam Vietnam - TU Delft