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Steel solutions forseismic retrofit and upgrade
of existing constructions
(Steelretro)
Research and Innovation EUR 25894 EN
EUROPEAN COMMISSION Directorate-General for Research and Innovation Directorate G — Industrial Technologies Unit G.5 — Research Fund for Coal and Steel
E-mail: [email protected] [email protected]
Contact: RFCS Publications
European Commission B-1049 Brussels
European Commission
Research Fund for Coal and SteelSteel solutions for seismic retrofit
and upgrade of existing constructions(Steelretro)
A. Braconi, A. TremeaRiva Acciaio S.p.A.
Viale Certosa 249, 20151 Milano, ITALY
G. Lomiento, N. Bonessio and F. BragaUniversità degli studi di Roma ‘La Sapienza’ CERI, Piazzale Aldo Moro, 5, 00185 Roma, ITALY
B. Hoffmeister and M. GündelRheinisch-Westfälische Technische Hochschule Aachen
Templergraben, 55, 52056 Aachen, GERMANY
S. A. Karmanos and G. VarelisUniversity of Thessaly Research Committee Argonauton & Filellinon, 38221 Volos, GREECE
R. ObialaArcelorMittal
Rue de Luxembourg 66, 4009 Esch-sur-Alzette, LUXEMBOURG
P. Tsintzos and D. VasilikisShelter Anonymos Voimichanki Etairia Ependyseon Kai Kataskevon
CHLM Larisas Sykourious 6, 41500 Larisa, GREECE
J. B. Lobo, P. Bartlam and S. C. EstanislauInstituto de Soldadura e Qualidade associação
Avenida do Professor Doutor Cavaco Silva, 33 Parque das tecnologias, 2740 120 Porto Salvo, PORTUGAL
L. Nardini, F. Morelli and W. SalvatoreUniversità di Pisa
Lungarno Pacinotti 43, 56100 Pisa, ITALY
D. Dubina, A. Dogariu and S. BordeaUniversitatea Politehnica Din Timisoara
Piata Victoriei 2, 300006 Timisoara, ROMANIA
G. Bortone, N. Signorini and G. FianchistiRegione Toscana
Via Cavour, 18, 50100 Firenze, ITALY
L. FulopTechnical Research Centre of FinlandVourimiehentie 3, 02044 Espoo, FINLAND
Grant Agreement RFSR-CT-2007-00050 1 July 2007 to 30 June 2010
Final report
Directorate-General for Research and Innovation
2013 EUR 25894 EN
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Table of contents
Final Summary 7
1. Recognition of problems affecting existing buildings 19
1.1. Vulnerability of existing buildings 19
1.1.1. Vulnerability framework 19
1.1.1.1. Dot Vulnerability 19
1.1.1.2. Local and Global Vulnerability 20
1.1.1.3. Vulnerability evaluation tables 21
1.1.2. Main vulnerabilities and typical problems 21
1.1.2.1. Dot vulnerabilities 22
1.1.2.2. Local vulnerabilities 22
1.1.2.3. Global vulnerabilities 23
1.2. Quality of materials in existing buildings: concrete and reinforcement 24
2. Performance based design (PBD) framework 29
2.1. Main concepts on Performance Based Earthquake Engineering 29
2.2. Analysis of existing PBE Framework 30
2.2.1. Building performance objectives 30
2.2.1.1. Combination of structural and non-structural damage levels for the definition of admissible
performance levels 31
2.2.2. Earthquake hazard level 32
2.2.3. Design Strategies 33
2.2.4. Knowledge of the structure to be retrofitted 34
2.3 Performance Based Assessment 34
2.3.1 Analysis methods, modeling and acceptance criteria 34
2.3.1.1. Modeling Parameters and Acceptance Criteria 34
2.3.1.2. Linear – Elastic Analysis 35
2.3.1.2.1 Lateral force method 35
2.3.1.2.2 Modal response spectrum and linear time-history 35
2.3.1.2.3. Acceptance criteria for linear analysis 35
2.3.1.3. Non-linear Analysis 35
2.3.1.3.1. Static – Pushover 35
2.3.1.3.2. Dynamic – Time-history 36
2.3.1.3.3. Acceptance criteria for nonlinear analysis 36
2.3.2 Analysis of Non-linear static procedure 36
2.4. Choice of the intervention technique 37
2.4.1. Structural performance based validation 38
2.4.2. Technical aspects 38
2.4.3. Economic aspects 38
2.5. Complete PBD framework assumed in the project (PBEE/PBA) 39
3. Analysis of existing retrofitting techniques 51
4.1. Description of reinforced concrete benchmark building 51
4.1.1. Materials and general geometry 51
4.2. Definition of the masonry benchmark building 53
4.2.1. Materials and general geometry 53
4.3. Calibration of numerical models 54
4.3.1. Reinforced concrete building 54
4.3.1.1. Non-linear modelling issues adopting SEISMOSTRUCT 54
4.3.1.1.1 Modelling of cross section 55
4.3.1.1.2 Performance criteria 55
4.3.1.2. Non-linear modelling issues adopting SAP2000 55
4.3.1.2.1 RC elements (beams and columns) 55
4.3.1.2.2 Modelling hypothesis 56
4.3.1.3. Non-linear modelling issues adopting DYNACS 56
4.3.1.4. Modelling issues using OPENSEES 57
4.3.1.4.1 Nominal material properties 57
4.3.1.4.2 Modelling of cross sections 58
4.3.1.4.3 Modelling of floor system 58
3
4.3.2. Masonry building 58
4.3.2.1 Material properties 59
4.3.3. Comparison of the results and identification of vulnerabilities in r.c. benchmark building 60
4.3.4. Initial assessment of the masonry building 62
4.3.4.1 Vertical loads 62
4.3.4.2 Horizontal loads 62
4.3.4.3. Deficiencies of the existing building 63
5. Performance analysis of steel solutions for vertical elements 65
5.1. Insertion of new elements in existing vertical systems 65
5.1.1. Analysis phase 65
5.1.2. Evaluation phase 66
5.1.3. Solution phase 66
5.1.4. Optimal sizing and placement 67
5.2. Performance analyses of steel techniques for vertical elements 71
5.2.1. R.C. benchmark 71
5.2.1.1. Buckling Restrained Bracings (BRB) 71
5.2.1.2. Steel and Composite Steel Concrete Shear wall 74
5.2.1.3. Light Gauge Steel panel 81
5.2.1.4. Steel concentric and eccentric bracings 85
5.2.2. Masonry benchmark 88
5.2.2.1. Tying the upper end of walls 90
5.2.2.2. Rigid diaphragm at the roof level 90
5.2.2.3. Rigid diaphragm at roof – LGS strips for external walls at ground floor 92
5.2.2.4. Rigid diaphragm at each floor – LGS strips for external walls at ground floor 93
5.2.2.5. Coupling of steel frames with existing masonry walls 94
5.2.2.6. Strengthening technique: Application of Bracing System 95
5.3. Comparison of analysed retrofitting techniques: structural performance vs. economic aspects
96
5.3.1 Cost analysis of the interventions 99
5.3.2 Practical implications and guidelines 102
6. Performance analysis of steel solutions for horizontal elements 105
6.1 Masonry benchmark structure 105
6.1.1. Intervention Techniques 105
6.1.1.1. Floor systems 105
6.1.1.2. Roof systems 106
6.1.2. Analysis results 106
6.1.3. Connection design for floor and roof systems 107
6.1.3.1. Replacing the existing timber floor system with Reinforced Concrete slab 107
6.1.3.2. Adding horizontal steel bracing systems 107
6.1.3.3. Replacing degraded parts with new steel parts 108
6.1.3.4. Adding trussed perimeter beam 108
6.1.3.5. The ring beam technique 108
6.1.3.6. Adding steel bracing system 108
6.1.3.7. Replacing degraded parts with new steel parts 108
6.2. Retrofitting or upgrading of floors/roofs for r.c. buildings 108
6.2.1. Floor systems in existing r.c. buildings 108
6.2.2. Retrofitting techniques for floor systems in existing r.c. frames 111
6.2.2.1.Post-tensioning of floors 111
6.2.2.2. Steel bracing 112
6.2.2.3. Steel collectors 112
7. Retrofitting technique for foundation system 115
7.1. Analysis of micro-piles for foundation retrofitting 115
7.2. Soil-structure interaction assessment 116
7.3. Influence of foundation retrofitting 119
7.4. Connection system between new elements and existing foundation 120
8. Experimental testing 123
8.1. Experimental investigations on Steel Shear Walls for seismic retrofitting 123
4
8.1.1. Tests on connections between shear panel and boundary elements 123
8.1.2. Tests on welded connections 123
8.1.3. Tests on connections with powder actuated fasteners, steel grade DX51D 124
8.1.4. Tests on connections with powder actuated fasteners, steel grade DX56D 124
8.1.5. Test on Steel Shear Walls 124
8.1.5.1. Loading procedure and measurements 126
8.1.5.2. Test 1: pure RC-frame 127
8.1.5.3. Test 2: Steel Shear Wall with welded shear panel in S235 127
8.1.5.4. Test 3: Steel Shear Wall with shear panel in DX51D fixed by fasteners 128
8.1.5.5. Test 4: RC-frame retrofitted by Steel Shear Wall with welded shear panel in S235 128
8.1.5.6. Test 5: RC-frame retrofitted by Steel Shear Wall with shear panel in DX51D fixed by
Fasteners 129
8.1.5.7. Evaluation of test results according to the ECCS-procedure 129
8.1.5.8. Tests on connection system between Steel Shear Wall and existing structure 130
8.1.6. Tests on connection system between new roofing / floor systems and existing structures 130
8.1.6.1 Test program and test set-up 130
8.1.6.2 Test results 130
8.2. Experimental Qualification of BRB systems for seismic retrofitting of R.C. frames 131
8.2.1. Testing set-up 133
8.2.2. Experimental Results 134
8.2.2.1. Monotonic tests 134
8.2.2.2. Cyclic tests 135
8.3. Experimental testing on novel dissipative bracing element 138
8.3.1. Test setup 140
8.3.2. Gauge system 141
8.3.3. Testing procedure 141
8.3.4. Results 142
9. Application to case studies and design guidelines 145
9.1 Patras House 145
9.1.1 General Description of the building 145
9.1.2 Assessment of the structural vulnerabilities 146
9.1.2.1. The developed numerical model 146
9.1.2.2. Performance of the Un-retrofitted Masonry Structure 147
9.1.3. Intervention techniques selected for the case study 148
9.1.4. Assessment of the retrofitted structure 148
9.2 “Immaculate conception” church 150
9.2.1 General description of the building 150
9.2.2 Assessment of the structural vulnerabilities 151
9.2.3. Intervention techniques selected for the case study 154
9.2.4 Assessment of the retrofitted structure 155
9.3. Bagnone building 156
9.3.1 General description of the building 156
9.3.2 Assessment of the structural vulnerabilities 156
10. Design guidelines 163
10.1. Steel buckling restrained braces 163
10.1.1. BRB system model 163
10.1.2. Specific provisions in design codes 163
10.1.3. Connections 142
10.2. Design guideline for Steel Shear Wall as seismic retrofit measure 167
10.2.1. General description of the retrofitting technique 167
10.2.2. Pre-Design, modelling and assessment rules for Steel Shear Walls 168
10.2.2.1. Pre-Design 168
10.2.2.2. Modelling 169
10.2.2.3. Connection between shear panel and boundary elements 169
10.2.2.3.1. Connection of Steel Shear Wall to existing structure 170
11. Results, general conclusions and perspectives 171
12. References 173
5
Executive Summary
Introduction Structural renovation of historical centres and of existing buildings is one of the most important
concerns of the construction sector, in which social, structural and economic aspects are often to be
considered simultaneously. The problem is particularly serious in earthquake prone areas – typically, in
European, Mediterranean Countries – where existing buildings should withstand seismic action
guaranteeing adequate safety levels for human life.
Modern standards answered to changes of social needs: nowadays, in addition to prevention of
structural collapse and safety of human life in the case of high intensity earthquake, modern seismic
design must guarantee low damage levels for seismic events of low and medium intensity, in order to
reduce the high economic costs due to post-earthquake interventions and interruption of productive
activities. Many approaches have been developed and have been incorporated in different standards that
at disposal of designers could work for driving the seismic retrofitting of construction in a proper way,
increasing safety levels and giving the right tools for taking on board also other aspects. Those design
approaches are generally indicated as Performance Based Earthquake Engineering, and are
characterized by multi-performance and multi-criteria approach in order to calibrated expected
performance of structural systems on defined values of the decision variables.
Nowadays, it is common to recognize the application of intervention techniques of poor quality or not
technologically advanced, hampering the benefits that the application of PBEE could bring in terms of
structural safety but also in terms of economic optimization. This can be partially addressed to the lack
of well-defined technical steel solutions and of their design rules: the high potential of steel solutions is
often unknown in common practice for designers and construction companies, so that in a similar
situation it is obvious that the choice of retrofit solutions in design practice is governed by personal
knowledge of operators.
Research Objectives The scope of the research proposal is to identify and propose steel solutions for seismic retrofit of
existing building – masonry and reinforced concrete buildings - in order to guarantee adequate seismic
safety levels and reduce eventual post-earthquake intervention and, at the same time, increase the
degree of standardisation.
Research plan and work carried out The research was carried through 9 WPs interconnected according to the general flowchart presented in
the figure I.
Figure I. General flow-chart of the research
7
Part 1. Definition of the methodology and tools and techniques pre-selection.
The research started with the recognition and identification of main and more diffused structural
vulnerabilities inside existing buildings; in particular, to execute this analysis an appropriate
vulnerability framework was set up and adopted in order to characterize the extension and the main
aspects of the structural seismic vulnerabilities in existing building both in masonry and reinforced
concrete, figure II. This framework aims to give general criteria that can be used in the identification of
vulnerable zones of buildings, apart from their specific typologies, and in the characterization of the
expected damage. With this aim, damages are primarily referred to the single elements that compose the
resisting structure (“dot vulnerability”) and are successively extended to limited portion of the main
resisting structure (“local vulnerability”) and to the overall structure (“global vulnerability”). The
proposed framework has been employed for discussing main vulnerabilities affecting structural types
examined during the research project and for individuating critical elements that could influence
structural response. In
Figure II. General framework in which the vulnerabilities identification were inserted.
Successively, the research focused the attention on the selection of an appropriate tool for the
application of PBEE design methodology in the choice of most appropriate steel based intervention
techniques in the next research phases (WP2). This process has been followed in order to establish a
common tool to be adopted in all numerical simulations and having so a full comparability of the
results. Different standards have been examined and the PBEE methodology was defined combining
FEMA356 and EN1998; in particular, from FEMA 356 the general framework related to the modelling
and the acceptance criteria has been taken, while design strategy, seismic hazard and the analysis
procedure have been taken from EN1998-1-1 (representative of a generic European Hazard).
8
The PBEE is a multi-criteria approach in which structural, technical and economic aspects are
combined in order to obtain a fully optimized solution. Due to the huge amounts of techniques already
applied in the practice to reinforced concrete and masonry buildings, it has been decided to execute a
previous typological/engineering judgement based matrix analysis focused on the individuation of those
techniques whose technical aspects were more relevant. In such a case, low quality and less performing
technique were immediately not considered for the investigation inside STEELRETRO project. For this
purpose, two decisional tools – matrix/form, figure III.a – were developed in order to rapidly and
extensively analyze existing retrofitting techniques suggested by the state-of-the-art, by the knowledge
and skills of the partners and appropriate for the main and typical structural vulnerabilities individuated
with the vulnerability framework adopted in the WP1. The interventions, analyzed with the decisional
matrix, were classified using summarized tables, figure III.b.
(a)
(b)
Figure III. (a) decisional matrix for the judgment of a single solutions; (b) summarizing tables of
intervention techniques for floor systems.
After a first selection of the intervention techniques, the steel based solutions appeared to be more
competitive in terms of performance, applicability and reversibility. Moreover, some of these steel
techniques selected according to the matrix approach have been designed and analyzed in order to, of
course, establishing their performance but also to have estimation about the materials and field works.
In such a way, on the basis of quantitative information, also the costs due to materials and other
activities (demolitions, temporary structures…) have been considered.
In particular, the work carried out on pre-selected techniques was mainly focused on techniques for
vertical members in both masonry and reinforced concrete elements (WP3), while the intervention
techniques on horizontal elements were analyzed in order to individuate main solution types and
interventions adapt to create optimal seismic conditions: in-plane stiffness for inertia forces re-
distribution (WP4 and WP5). Moreover, concerning the techniques for foundation systems, the micro-
piles technique was judged the unique technique characterized by low-intrusion requirements,
considered as a necessary pre-requisite for containing costs and having a certain feasibility level.
Part 2. Numerical simulations and analyses of the techniques
The analysis of techniques for vertical elements was executed testing them on the same structures: two
benchmark buildings– one masonry building and one reinforced concrete building – were designed
using old structural standard issued in Italy on 1939, in order to have case studies characterized by
typical vulnerabilities individuated in the WP1. These two solutions were assessed in order check their
respective vulnerabilities, see figure IV and figure V.
Structural aspects L M H Mark
Capability to achieve requested performance objective (after building evaluation!)
Compatibility with the actual structural system (no need of complementary strengthening or confinement measures)
Adaptability to change of actions seismic typology (near field, far field, T<>Tic, etc)
Adaptability to change of building typology
Technical aspects L M H Mark
Reversibility of intervention
Durability Operational Functionally and aesthetically compatible and complementary to the existing building
Sustainability Technical capability Technical support (Codification, Recommendations, Technical rules)
Availability of material/device Quality control
Economical aspects L M H Mark
Costs (Material/Fabrication, Transportation, Erection, Installation, Maintenance, Preparatory works)
Stiffness Resistance Ductility
Concrete overlay Yes Yes No
Shotcrete Yes Yes No
Glued fins (floors) Yes Yes No
Post-tensioning (floors) No Yes No
Steel bracing Yes Yes No
Precast element joints No Yes No
Concrete jacketing Limited Yes Yes
Steel jacketing Limited Yes Yes
Glued fins (beams) Limited Yes Yes
Post-tensioning (beams) No Yes Yes
Steel ledger No Yes No
Concrete ledger No Yes No
Local post-tensioning No Yes if part of MRF
9
(a)
(b)
Figure IV. (a) r.c. benchmark building; (b) FEM model of r.c. benchmark for structural assessment
(a)
(b)
Figure V. (a) masonry benchmark building; (b) ABAQUS FEM model for structural assessment.
The numerical analyses executed on masonry building were executed by partners using the same model
developed by one of them and successively distributed; for this reason all results were considered as
comparable. On the contrary the numerical analyses executed on the reinforced concrete building were
executed using different software: SAP2000, OPENSEES, SEISMOSTRUCT and DYNACS; for this
reason, a preliminary benchmarking process was carried out comparing predicted maximum force and
available ductility of the un-retrofitted solution. After this preliminary investigation on the r.c.
benchmark building, several techniques, figure VI, were tested using the benchmark buildings and can
be here summarized:
Steel bracing configurations; (R.C. and Masonry)
parallel steel frames; (Masonry)
BRB bracing configurations; (R.C.)
shear steel walls; (R.C.)
light gauge steel walls; (R.C.)
steel strips. (Masonry)
All the numerical analyses were carried out on benchmark buildings considering the roof and floors
already retrofitted in such a way to have in-plane rigid diaphragm action; moreover, it is important to
underline that this assumption for the r.c. benchmark was near the real condition given that the floor
system was sufficiently in-plane stiff and in-plane strong. Moreover, it was also executed for the
Masonry buildings the influence of the modification of roof and floor in-plane stiffness, in order to
appreciate its influence on the global response of the structure.
The analyses showed that for reinforced concrete building, see figure VI, more effective solutions were
the following:
steel bracing systems (with and without additional dissipative devices);
BRB bracing systems;
Shear steel walls (using low grade steels with low thickness plates).
Secondary beam 3030
Main beam 4060 Main beam 4060 Main beam 4060 Main beam 4060 Main beam 4060
Main beam 4060Main beam 4060Main beam 4060Main beam 4060Main beam 4060
Main beam 3055 Main beam 3055 Main beam 3055 Main beam 3055 Main beam 3055
Roof beam 3020
Roof beam 3020 Roof beam 3020
Roof beam 3020
Top Main beam 3050
Top Main beam 3050 Top Main beam 3050
T foundation beam
10010050
T foundation beam
10010050
Secondary beam 3030
Secondary beam 3030Secondary beam 3030
Secondary beam 3030 Secondary beam 3030
40
4040
40
40
40
40
4040
40
40
4040
40
40
40
30
30
30
30
30
30
30
30
30
30
100
390
340
335
180
50
310
30
310
30
305
30
1015
1195
305
280
20
390
380
360
38
38
38
160
38
16
22
360
38
317
20
36
301
418
398
337
197
0.00
418
816
1153
1350
1
2
3
10
Concerning the masonry buildings, see figure VII, the following techniques presented were judged as
those more effective for the improvement of final seismic performance:
Braced steel frames;
Modification of diaphragmatic actions of roof systems;
Steel strips inside masonry for improving mechanical properties of the wall.
It is also worth underlining that also the foundation system has been considered already retrofitted,
during the numerical analyses carried out in WP3, producing fixed constrains at the base of the
benchmark structures.
(a)
(b)
(c)
(d)
(e)
(f)
Figure VI. (a) hot-rolled steel plates; (b) BRB system; (c) light gauge steel walls; (d) elastic bracings;
(e) eccentric bracing systems; (f) bracing system with additional dissipative devices.
During the development of the numerical simulations at global levels on benchmarks, the retrofitting
techniques on horizontal elements as roofs, floors and foundations were also analyzed using numerical
simulation and typological analyses, for focusing more in detail the techniques and their respective
performance (WP4, WP5 and WP6).
In particular, the foundation systems were analyzed considering, as already said, the micro-pile systems
technique and the first working hypothesis for such analysis was to assume an approximate fixed
condition as expected performance from retrofitted foundation; the input variables for sizing and
designing the foundations were considered the base reactions from WP3 analyses. The study (WP6) was
executed analyzing various micro-piles configurations and considering different soil conditions, figure
VIII. At the end of the study a model for soil-structure interaction was created and adopted for re-
6
1st floor
2nd floor
3rd floor
5
53A
B3C
D1E
Rottura per pressoflessione colonna
11
evaluating some of the WP3 solutions (benchmark buildings + retrofitting solutions), considering the
soil-structure interaction modeled as equivalent springs representative of the deformability of micro-
piles and soil. This beneficial effect, produced by the equivalent springs at the base of the benchmark
buildings, was used to recalculate some of the steel solutions adopted for the r.c. buildings in order to
complete optimize their size and then develop from these specimens to be tested (WP7).
The floor and roofs systems were studied looking for selecting appropriate typological and technical
solutions able to guarantee the hypothesis (i.e. rigid diaphragmatic action) adopted in the execution of
numerical analyses in WP3. Steel based intervention techniques as steel bracings or planar trussed beam
were applied to existing floors; in particular, the floor and roof configuration of reinforced concrete and
masonry were adopted as case studies on which testing various intervention techniques (WP4 and
WP5). After some trials, it was observed that the insertion of bracing elements or steel stiffening
systems were more appropriate and performing in the realization of diaphragmatic conditions assumed
in the numerical analyses carried out in WP3.
(a)
(b)
(c)
(d)
Figure VII. (a) parallel steel frame; (b) braced steel frame; (c) insertion of steel strips inside masonry;
(d) modification of roof diaphragmatic action: very stiff the roof and deformable the floors.
12
(a)
(b)
Figure VIII. Study on techniques for improving existing foundations: (a) micro-pile model; (b)
geotechnical information of soil characterization
(a)
(b)
Figure IX. (a) Floor deformation equipped with different techniques; (b) roof in-plane deformation.
Part 3. Experimental programme for testing selected techniques
The conclusion of the numerical analyses executed in WP3, on reinforced concrete benchmark,
suggested to study by means of experimental programme the following techniques for retrofitting
vertical elements:
BRB systems;
Steel shear walls;
Dissipative bracings (bracing elements + additional dissipative devices).
These three systems were experimentally characterized by means of ECCS procedure, in order to assess
their performance and their capabilities of retrofitting existing buildings, as planned in the research
work-plan.
- A BRB systems was completely developed by CEMSIG laboratory that following a detailed
process took care about designing, assembling and testing all the system components;
subsequently, the complete BRB system was experimentally characterized: strength, stiffness,
ductility and energy dissipation capacities were assessed. After this preparatory tests, some full-
scale tests were carried out, coupling the BRB systems, using different configurations, with a
r.c. frame (1 bay-1 story) extracted from the benchmark used in the WP3, figure X.
- The full-scale specimens of the steel shear wall system were designed considering a novel
connecting system between the shear wall and the existing floor of building: it was decided
where and how horizontal and/or vertical forces are transferred. Additional load transfer beams,
figure XI.a, was found as favourable as they enable to direct the forces to parts of the existing
structure with a sufficient capacity. Insert through anchoring designed was validated as
favourable rigid connecting system for RC-structures due to the high capacity and the
possibility to balance tolerances. Moreover, also in these case the ECCS procedure was
employed for testing the intervention techniques and the material realizing the steel wall were
tested too, figure XI.b and XI.c).
13
- The last experimental programme was that carried out on a novel dissipative bracing system,
developed a completely new system, actually under patenting process, able to integrate
different features. The novel dissipative device was studied considering the feasibility aspects
before the realization of first two prototypes: this work was carried out with the cooperation and
sustain of another research project, contemporary developed, for finding extra resources in
order to complete this purpose (wider and more demanding respect to original aims of the
research). Preliminary tests for calibrating the mechanical systems and final tests were carried
out, figure XIII; the theoretical model indicated that the systems should demonstrate s Flag
Shaped cycle as Hysteretic Device (FSHD) and tests confirms this aspect.
(a)
(b)
Figure X. (a) full-scale testing on BRB +R.C. Frame systems; (b) initial qualification of material
properties.
(a)
(b)
(c)
Figure XI. (a) steel shear wall coupled with r.c. frame; (b) preliminary tensile tests on steel qualities; (c)
tested coupon.
(a) (b(
(c)
Figure XII. (a) testing on steel quality; (b) FSHD system; (c) buckling restraining system for steel fuses.
Material influence in BRB modeling Theoretical Quality Certificate Experimental
Standard EN10025:1993 EN10051 Class A EN 10002-1
BRB steel plate grade S235JRG2 S235JRG2 Specimen Test
Minimum Yield strength Re [N/mm2] 235 255 335
Tensile strength Rm [N/mm2] 340 - 470 360 439
Minimum Elongation % 26 39 28
14
(a)
(b)
(c)
Figure XIII. (a) first tests of FSHD system – not satisfactory behaviour/modification of the system; (b)
and (c) two examples from second series of tests carried out modifying internal properties of the system
(note: to shorten the test procedure only 1 cycle was executed for each displacement level)
Part 4. Applications to case studies and technical guidelines
After this experimental testing programme, the research focused the final activities on the application of
steel based intervention techniques on real case studies; more in the details, four case studies were
considered: an house building located in Patras (GR) and realized with stone masonry; an old historical
building located in Timisoara (RO) – Huniade Castle – and realized with brick masonry for walls,
vaults and pillars; the Immaculate Conception Church located in Brescia (I) representing together with
Huniade Castel an historical case study; the High School building, a reinforced concrete structure
located in Bagnone (I). The project terminated with the definition of short guidelines for two
intervention techniques experimentally tested.
Results, general conclusions and perspectives The research project dealt with the complex problem of defining appropriate intervention techniques for
existing buildings, a not simple task given that in the design practice all retrofitting interventions can be
considered as unique because of particular boundary/environmental conditions that the building has.
Nevertheless, the research consortium tried to face the problem suitably combining different tools and
methods in order to have a systematic approach and at the same time an experimental programme was
also carried out for developing and testing retrofitting techniques to be proposed as valuable solutions to
the practitioners.
In particular, during the research project the following steps (assumed as ‘methodology’) have been
followed in order to systematically treat the seismic retrofitting of existing constructions:
1. defining a framework for surveying existing constructions and recognizing potential
vulnerabilities;
2. choosing a PBEE methodology, composing together design strategy, hazard model, modelling
techniques, simulation method, acceptance criteria (i.e. FEMA 356 and EN1998), technical
aspects and economic model for cost estimation;
3. defining a matrix approach that have been used a first pre-selecting method for analysing most
common techniques (also not steel based) and individuating those that were technically not
convenient (i.e. accessibility, difficulty level for applicability, manpower skill for in-field
works, demolition, previous technical evidences…);
4. defining two benchmark structures on which different steel solutions, pre-selected or derived
from the application of the matrix approach at point 3, have been applied (using chosen PBEE)
and the results of such applications have been so able to be compared;
5. analysis of the structural response at the foundation level, evaluating the required bearing
capacity of the foundations and designing of the intervention techniques;
6. considering the upgraded foundation system applied to the structures, definition of a simplified
soil-structure interaction model and re-analysis of the complete retrofitted structures in order to
secure the reached safety level, previously determined, and eventually optimize the structural
elements in the upper structure.
In general, these steps should be considered as mandatory for every designer engaged in the seismic
retrofitting of the existing constructions, considering that this sequence of steps has been applied to
different structural systems in the research project, confirming the applicability of the methodology.
In particular, the knowledge phase of the structure – step 1 – it is always a fundamental process that is
usually executed in a different way according to personal skills or to different structural types. The step
1 of the methodology adopted in the research could support the designer in this phase, because it faces
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the approach to the structural system irrespectively of the types or of the configuration, in a quite
systematic way. At the end of this logic process, the potential vulnerabilities and the structural parts on
which focusing the investigations can be highlighted and the structural assessment can be executed,
using calculus method that designer considers much more appropriate inside the vulnerability
framework herein adopted.
Another important step is the selection of retrofitting techniques to be analysed and the designers should
look at those techniques that, first of all, are characterized by technical feasibility if examined in the
perspectives of the preliminary information obtained from the preliminary vulnerability assessment of
the existing construction to be retrofitted. Also in this case, practitioners are often used facing the
problem without a general approach or with a partial analysis; the step 2 of the methodology here
proposed tried to answer to his point in a simplified way, applicable in the practice, but maintaining a
systematic approach. The designer can use the matrix approach, considering the (qualitative) variables
that for him have more importance to compare and preselect the techniques before the application of
PBEE that requires a high computational effort.
The steps 3, 4, 5 and 6 are those related to the application of the PBEE and, above all, to the execution
of numerical analyses for sizing the retrofitting techniques, quantifying their effectiveness and
completing the design process. Of course, the step 1 and step 2 are fundamental in the methodology
because their information drive the development of the next phase of the design process.
The application of the methodology to several techniques has allowed, in the first steps, to pre-select
those more interesting and afterwards has allowed the final assessment of seismic performance of those
more performing: Steel bracing configurations; parallel steel frames; BRB bracing configurations; shear
steel walls; light gauge steel walls; steel strips. Moreover, it has been also executed an economic
comparison between different techniques in order to appreciate the impact of costs of the different
solutions.
The complete application of the methodology to those different techniques as allowed also the accurate
analysis of three steel based intervention techniques and the designing of three base cases, sized on the
same benchmark structure – r.c. – that have been subjected to experimental testing. The test
programme, in particular, has been focused on the retrofitting of r.c. concrete structures but the results
and the techniques could be directly extended and applied to masonry structures also.
The three techniques experimentally tested have been:
Buckling Restrained Bracing system; - BRB
Shear Steel wall (with innovative connection system); - SSW
Flag Shaped Hysteretic Dissipative Bracing system with re-centering capabilities. - FSHD
All these three techniques have been selected from the previous numerical simulations because they can
effectively answer to the problems related to the retrofitting of existing constructions, in which strength,
stiffness and ductility deficiencies could be detected contemporary or separately, obliging the designers
for looking at different techniques for addressing such deficiencies singularly, coupled or altogether. In
particular, the development of such techniques and their application to the benchmark structures
allowed verifying their flexibility in grading mechanical properties (i.e. strength, stiffness and ductility),
confirmed also by experimental testing programme carried out in three different laboratories.
Moreover, it also important to stress that one of the major problems of seismic retrofitting is the
localization of stresses/forces that pass from existing structure to the new ones (retrofitting system) and
this phenomena is as much pronounced as less stiffness and strength cannot be controlled into the
retrofitting systems. This aspect has been taken into account; in fact, BRB system and FSHD system do
not localize high level of forces due to their intrinsic possibility of modifying their yielding threshold
and their initial stiffness, through a refined sizing of their internal components. The SSW system in
general are considered as retrofitting techniques characterized by high stiffness (only), high resistance
and by imposing an high resistance demand on surrounding columns, obliging so the designers to costly
and complex local retrofitting actions. These shortcomings from SSW system have been brilliantly
solved defining a novel mechanically composed system in which steel panels can be taken from a wide
variety of qualities (i.e. automotive <1mm to structural >3mm), graduating so the strength and the
stiffness. Moreover, the system is connected to the structure using a beam system connected to the floor
and able to do not create over-turning moments; in such a way, the surrounding columns and the beams
are not overloaded by the retrofitting scheme.
These three techniques represent solutions with a high technological and conceptual contents and their
flexibility proposes those as appropriate for the application of PBEE to the seismic retrofitting of
existing constructions (i.e. grading structural response of retrofitted structures with the different
16
earthquake intensities and correlating them with expected building performance). Moreover, design
guidelines have been developed for BRB system and for SSW system, while the guidelines for FSHD
system are still under development due to the patenting process at which this system has been subjected.
At the end of the research project, some real case studies have been analysed in order to individuate
their vulnerabilities and proposing retrofitting techniques between those analysed during the research.
The STEELRETRO project presented as main general outcome the development of steel based
techniques endowed with high technological content; in particular, two of those are novel techniques
and one of those is subjected to a patenting process.
Moreover, the development of these techniques has required the definition of a ‘real’ and ‘technically
sound’ working environment in order to develop, size and assess these techniques using
applicable/feasible methods and to compare their performance with real or representative demands.
For such a reason, inside the STEELRETRO project a methodology for approaching to the problem of
the seismic retrofitting has been set up, combining together several tools for treating/managing the
various aspect that a seismic retrofitting always involve. In particular, the methodology has been
defined following the logical process that a good practitioner should follow.
17
1. RECOGNITION OF PROBLEMS AFFECTING EXISTING
BUILDINGS
1.1. VULNERABILITY OF EXISTING BUILDINGS A comprehensive evaluation of the vulnerability is the preliminary step to the choice of adequate
retrofit solutions for existing buildings. This chapter aims to give a simple “vulnerability framework”
oriented to find the critical aspects that mainly affect the seismic vulnerability of different typologies of
buildings. The evaluation of the vulnerability is based on the following three steps:
1. definition of performance requirements;
2. quantification of seismic action;
3. evaluation of seismic vulnerability.
The evaluation of seismic vulnerability is obtained from the comparison between demand and capacity
of the construction where:
the demand is the maximum request imposed by actions and loads in terms of stresses and
strains/deformations;
the capacity is the maximum value of the demand parameter that the construction is able to
fulfill.
According to the performance standard framework assumed as reference (EN1998-3), the demand is
organized in a multi-level framework, in which each level is linked to a different intensity of seismic
action, as well as each capacity level is related to a different Limit State. Performance requirements will
be extensively discussed in section 2 “Performance Based Design Framework”.
In the vulnerability framework, general criteria are given in order to identify the critical zones of a
building, given the performance requirements and the seismic action. The critical zones are identified as
those, among all the structural parts of the buildings, in which damages could happen more easily, even
causing the collapse of the building. According to these criteria, results of the vulnerability evaluation
are organized in simplified tables and are used for the choice of the retrofit solutions to be applied. A
synthetic review of most common vulnerabilities and typical problems affecting buildings are given on
the basis of the considered case studies.
1.1.1. VULNERABILITY FRAMEWORK Seismic vulnerability of constructions is evaluated on the basis of the behaviour:
of single parts (structural elements) that compose the structure– “dot vulnerability”;
of structural sub-systems, in which single structural elements are assembled dependently on
the static role of the sub-systems inside the construction – “local vulnerability”;
of constructive typology in which single sub-systems are assembled – “global vulnerability”.
1.1.1.1. Dot Vulnerability In order to perform the evaluation and quantification of “dot vulnerability” (vulnerability of the single
structural element), several criteria oriented to the individuation and classification of more vulnerable
zones (“critical zones”) of single structural elements are given.
Each structural system is composed by different structural elements, each of them with a proper role.
The subdivision of elements on the basis of their structural role is used to identify possible “critical
zones”.
Essentially, all structural elements have two functions:
to collect, generally, loads and actions of different types;
to transfer collected loads and actions to other structural elements.
Besides these functions, each structural element carries its self weight. Finally, foundation elements
have the role of spreading in the ground all collected loads and global self weight.
In the following, elements that collect actions and loads will be named collecting elements while
elements on which loads are transferred by collecting elements will be named supporting elements.
The zones in which the generic element collects the internal actions (collecting zones) and those in
which it transfers actions (transferring zones) to supporting elements can be easily identified.
Collecting zones and transferring zones can be divided into different classes:
“s” zones: surface zones (a collecting or transferring surface is identified);
“l” zones: linear zones (a collecting or transferring alignment is identified)
“p” zones: dot zones (a collecting or transferring point is identified)
Structural elements can be likewise subdivided in classes according to the dimensions needed for the
19
description of their mechanical behaviour:
class “1” elements: 1D elements (beams, columns, arches, rods…)
class “2” elements: 2D elements (walls, slabs, vaults…)
class “3” elements: 3D elements (solid connection, stocky cantilevers, foundation plinths,
anchorage blocks, foundation ground,…)
Four general postulates, for the characterization of structural element working, can be formulated:
1. each structural elements must have at least one collecting zone and one transferring zone;
2. collecting elements are characterized by a number of prevailing dimensions not lower than the
dimensions necessary for the characterization of its collecting zones (e.g. if one element is
delegated to collect “surface” actions, it must be of class “2” at least);
3. contact zones between collecting elements and supporting elements (transferring zones) have a
number of dimensions always lower than the number of prevailing dimensions of collecting
elements (e.g. a plate, class “2” element, can be supported by columns, identifying class “p”
transferring zones, or walls, identifying class “l” transferring zones; a beams, class “1”, can be
supported by columns, identifying class “p” transferring zones, or walls, identifying class “p”
transferring zones);
4. transferring zones of a collecting element becomes collecting zones for the supporting element
on which the load are transferred.
5. The load path, from collecting zones to transferring zones, defines the mechanical behaviour of
the generic collecting element and then the possible critical zones, in which the damage can be
eventually localized.
In order to identify critical zones inside the generic structural element and to classify type of problems
that can occur, the demand and the capacity of the element have to be detected. Demand means the
maximum stresses and strain requested by loads and action to structural element; capacity means
maximum value of the demand, in terms of stresses and strains, that the structural element is able to
fulfill.
The identification of critical zones is made on the basis of the ratio between demand and capacity; as
more demand is approaching capacity, as the examined zone tends to become a critical zone.
In the definition of the effective capacity of a structural element, stress-strain constitutive laws of the
materials have a primary role. As far as ductile behaviour or brittle behaviour, materials have to be
distinguished. Brittle or ductile behaviour of single structural elements directly depends on
brittle/ductile behaviour of constituent materials, as well as on constructive details used in the critical
zones and on the induced capacity/demand ratio.
When the demand reaches the material resistance (capacity in terms of stress), damage occurs in the
critical zone if the deformation demand is still lower than the deformation capacity (ductile behaviour)
until the ultimate deformations are reached and collapse occurs; otherwise, if the material resistance and
the deformation capacity are reached at the same time (brittle behaviour), the critical zone becomes
earlier a collapse zone.
The individuation of the critical zones is executed considering separately every structural element. As
already said, critical zones are defined as zones in which the ratio between demand and capacity, in
terms of stresses, tends to 1.
Two types of critical zones are individuated:
type “a” zone: inside the structural element where, demand/capacity ratio in terms of stresses
tends to 1 along specific load paths; this circumstance takes place in collecting zones, when they have a
number of dimensions lower than number of prevailing dimensions of the collecting element (e.g. at
the connection between columns and plates, i.e. “p” collecting zone on “class 2” collecting element) or
when the load path concerns discontinuity zones of the collecting element (e.g. openings in the wall);
type “b” zone: in transferring zones, that always have a number of dimensions lower that
dimensions of collecting element and, for that reason, are subjected to concentration in stress demand
that can approach material resistance.
Definitively, critical zones of type “a” are localized inside the collecting element, while type “b”
critical zones are at the intersection between collecting element and supporting element and include
adjacent zones of the collecting element and the supporting element, with a dimension at least equal
to the thickness of the element. All retrofit interventions will be finalized to the reduction of
stress/deformation demand in critical zones.
Localization and classification of critical zones in structural elements has to be executed for each
structural type and for each class of structural element, apart from the levels of seismic hazard
20
considered. The demand/capacity analysis will allows to figure out where damage is going to be
expected and its entity. Damages in multiple locations can produce local or global collapse
mechanisms, as discussed in the following section.
1.1.1.2. Local and Global Vulnerability Local vulnerability is interpreted as the vulnerability of a single portion of a structural system
(“structural sub-system”) that accomplishes to a precise static function. Global vulnerability is
interpreted as the vulnerability due to the interaction between different structural sub-systems that can
involve from the most part to the whole structure.
Likewise to what said in the previous paragraph, in which localization and classification of critical
zones of structural elements is defined, it is necessary to clarify the “path” that carries to the evaluation
of local and global vulnerabilities and where and how damages and failures can happen.
Since the modalities that a structure follows in the infringement of considered Limit States are
extremely diversified depending on the different structural behaviour characterizing different structural
types, the evaluation of global seismic vulnerability is finalized to the individuation of typical structural
behaviour for each types.
Considered building types are:
reinforced concrete buildings;
masonry buildings;
historical buildings.
The evaluation of the vulnerability has to be dealt separately for each building type, in such a way to
highlight the peculiarities. Therefore also for each building type, single structural sub-systems and their
reciprocal interactions will be separately dealt.
As a general idea, “dot vulnerabilities” have to be checked for Limit States that involve control of the
damage, while “local vulnerabilities” should be prevented for Limit States addressed to the life safety
and “global vulnerabilities” have to be considered in order to fulfil the collapse prevention.
In order that a local or global collapse could take place, it is necessary to have ductile damages and/or
the as much failures of elements as much high is the redundancy level of the structural sub-system or of
the overall structure, because these have to become mechanisms.
1.1.1.3. Vulnerability evaluation tables Simplified tables are proposed in order to summarize results of the vulnerability evaluation on the
buildings. With this aim, each scheme should be referred to a particular building and the structural sub-
systems are subdivided in:
- roofing and floors systems (including all the structural elements of the building roof and floors);
- vertical resisting system (including all the structural elements supporting the building roof and floors,
e.g. walls in masonry buildings and beams/columns frame in reinforced concrete buildings);
- foundation system (including all the structural elements transmitting the loads to the ground and the
ground itself).
1.1.2. Main vulnerabilities and typical problems Main problems affecting existing buildings in seismic areas were analyzed – subdivided in reinforced
concrete, masonry and historical buildings in areas with low and high seismicity. In this report synthetic
results for the analyzed buildings are presented.
BUILDING SUMMARY: building type, location, age of construction.
VULNER.
TYPE
DESCRIPTION STRUCTURAL
SUB-SYSTEM
CRITICAL ZONES AND
ELEMENTS INVOLVED
LIMIT STATE DEMAND/
CAPACITY
DOT e.g. shear failure in the wall
near openings
e.g. vertical
resisting system
(localization)
e.g. type “a” critical zone (openings),
class “2” collecting element (wall)
e.g. Limit State of
Damage Limitation
e.g. 1,5
LOCAL e.g. out-of-plane failure of
the wall with failure of
wall-to-wall connections
e.g. vertical
resisting system
(localization)
e.g. type “b” critical zones (wall
connections), class “2” collecting
elements (walls), type “l” transferring
zones (wall-to-wall connections)
e.g. Limit State of
Life Safety
e.g. 2,1
GLOBAL e.g. failure of corner
between two orthogonal
walls with following failure
of the roof system
e.g. vertical
resisting systems
and roof system
(localization)
e.g. type “a” critical zones (openings),
class “2” collecting elements (walls),
type “l” transferring zones (wall-to-wall
connections and roof-to-wall
connections)
e.g. Limit State of
Collapse Prevention
e.g. 1,3
21
1.1.2.1. Dot vulnerabilities The main vulnerabilities found at “dot” level have highlighted the following zones as the mostly
common critical zones.
Critical zone type “a” can be found in the following collecting elements:
Class 1 elements:
change of cross-section in beams and columns
beams supporting columns
arches with deflections due to settlements
columns with beams at different levels (e.g. in stairs frames)
frames irregularly filled by masonry infill
Class 2 elements
walls with openings, mostly if with an irregular pattern
floor slabs with openings
slabs supporting columns
Class 3 elements
stocky cantilevers
foundation plinths
foundation ground
Critical zone type “b” can be found in the following transferring zones:
“p” zones
connections between beams and columns
connections between columns and foundation elements
connections between beams and walls
connections between columns and slabs
connections between roof elements and walls/beams
connections between arches and columns/walls
connections between rods and columns/walls
connections between vaults/domes and columns
connections between piles and plinths/slabs
“l” zones
connections between walls and foundation elements
connections between walls and slabs
connections between walls and walls
connections between floor slabs and walls/beams
connections between vaults/domes and beams/walls
“Dot vulnerabilities” are commonly found in different building typologies, while “local vulnerabilities”
and “global vulnerabilities” can significantly differ from one typology to another one.
1.1.2.2. Local vulnerabilities
The “local vulnerabilities” most commonly verified are summarized for the three structural sub-systems
previously defined.
Roofing and floors systems
In seismic design, roofing and floors systems provide diaphragm capacity, serving to interconnect the
building and acting to transmit lateral force to the vertical resisting elements.
Diaphragm forces are derived from the self weight of the diaphragm and the weight of the elements and
components that depend on the diaphragm for lateral support. Any roof, floor, or ceiling can participate
in the distribution of lateral forces to vertical elements up to the limit of its strength. The degree to
which it participates depends on relative stiffness and on connections. In order to function as a
diaphragm, horizontal elements must be interconnected to transfer shear with connections that have
some degree of stiffness.
The most common diaphragm deficiencies in buildings are characterized by:
Extreme flexibility;
Lack of continuity (caused from split level floors and roofs, or diaphragms interrupted by
expansion joints);
Large openings at shear walls;
Plan irregularities (such as extending wings, plan insets, or E-, T-, X-, L-, or C-shaped
22
configurations where large tensile and compressive forces can develop).
The basic function of the diaphragm is to tie the elements of a structure together at a given level and
distribute inertial loads to the various vertical elements of the lateral force resisting system. Diaphragms
which are extremely flexible can result in very large inter-story drifts for supported elements such as
walls subjected to out-of-plane loads. It is important that the diaphragm have adequate stiffness to
prevent excessive inter-story drifts from developing.
Vertical resisting system
In typical concrete buildings structural systems the vertical elements are essentially the beams-columns
frames and the majority of structural deficiencies in concrete columns can be attributed to lack of
transverse reinforcement. This is especially true for buildings in seismically active regions, designed
prior to the enactment of modern seismic codes.
In particular, columns are critical elements in any structural system and their performance during a
seismic event can dominate the overall outcome of the structure. Failure of the reinforced concrete
columns in shear usually takes place at low deformations and is associated with a large and sudden drop
in lateral load resistance. Moreover, the shear strength of a column tends to degrade faster than its
flexural strength with cycling of the lateral load.
Based on as-built information, failure of the reinforced concrete frames are generally associated to lack
of ductility behavior due to deficient design detailing and/or deficient quality of the construction works
that are characterized essentially by:
deficient column bar and beam bar splices;
large column tie spacing and large stirrup spacing;
insufficient anchor lengths;
insufficient joints reinforcing and joints eccentricity;
in-plan and in-elevation irregularities.
In masonry and historical buildings the vertical elements are essentially masonry walls and the majority
of structural deficiencies can be attributed to poor design and/or deficient quality of the material and of
construction works.
Failure of the masonry vertical structural systems are due essentially to:
poor resistance of the materials (mortar and masonry blocks);
insufficient thickness of the panels;
wide presence of openings, especially with irregularly patterns;
separation of walls and gables;
insufficient floor-to-wall connections (especially wooden floor systems but also reinforced
concrete floors are delicate for combined vertical and horizontal loads in case of insufficient
supports);
in-plan and in-elevation irregularities
Foundation system
Foundation deficiencies can occur within the foundation element itself, or due to inadequate transfer
mechanisms between foundation and soil. The failure of one foundation elements is often associated
with the failure of a portion of the whole foundation system.
Element deficiencies basically include:
Inadequate bending or shear strength of spread foundations and grade beams;
Inadequate axial capacity or detailing of piles and piers;
Weak and degrading connections between piles, piers, and caps.
Transfer deficiencies include:
excessive settlement or bearing failure;
excessive rotation;
inadequate tension capacity of deep foundations;
or loss of bearing capacity due to liquefaction.
1.1.2.3. Global vulnerabilities The global vulnerability of a structure is the susceptibility of all the structural elements with direct
participation in the load carrying system (foundations, columns, supporting walls, beams, floor, slabs
and any others), to damage at local level as well as its consequences for the stability of the building
system when subjected to earthquake load.
A deficiency in global strength is common in older buildings either due to a complete lack of seismic
23
design or a design to an early code with inadequate strength requirements. However, it is seldom the
only deficiency and the results of the evaluation must be studied to identify deficiencies that may not be
mitigated solely by adding strength. Also the lack of lateral stiffness may be critical in order to protect
non-structural components of the building.
In reinforced concrete building, local vulnerabilities associated to vertical resisting systems are often
associated with global vulnerability effects. Conversely, in masonry and historical buildings, global
vulnerability is generally associated to collapse mechanisms of portion of the buildings, involving
orthogonal walls and roofing/floor systems.
There are other deficiencies that should be accounted for the global vulnerability of a structure and have
also significant effects on seismic performance. Based on as-built information and among other
deficiencies, the ones that are more common and have more significant effects on seismic performance
are:
presence of adjacent buildings;
deterioration of structural materials.
The issues associated to the adjacent buildings occur when the gap between buildings is insufficient to
accommodate the combined seismic deformations of the constructions, both may be vulnerable to
structural damage from the "pounding" action that results when the two collide. Building pounding can
alter the dynamic response of both buildings and impart additional inertial loads on both structures.
The deterioration and damage of structural materials may have an adverse effect on the seismic
performance of an existing building during a severe earthquake. Deteriorated structural materials may
reduce the capacity of all the vertical resisting systems.
1.2. Quality of materials in existing buildings: concrete and reinforcement The recognition of structural deficiencies and intrinsic vulnerabilities need the study of mechanical
properties of existing materials, basic information for facing each seismic vulnerability assessment.
During the research project, a accurate analysis of material quality was executed investigating, in
particular, steel reinforcement properties, through statistical analysis of testing certificates issued by
official laboratories, and concrete properties, tested by Seismic Regional Service of Tuscany. First of all
an accurate historical investigation of structural codes, issued in Italy assumed as case studies, was
performed, see table 1.1 as example of code issued in 1907, in order to individuate quality classes
across the years. Moreover, the historical investigation using the qualities individuated by old structural
codes analyzed testing certificates organizing the results in terms of steel quality or in term of round
bars or shaped bars, see figure 1.1 where some typical Italian old bars are reported.
Table 1.1. Requirement for steel reinforcement adoption in structural design – 1957-1972
Issue's date
Reinforcing steel
denominationMin Max Min Max Min Max Tension Tension
Aq 42 230 - 420 500 20 - 140 50% yielding
Aq 50 270 - 500 600 16 - 160 50% yielding
Aq 60 310 - 600 700 14 - 180 50% yielding
Special Shaped Steels - - - - 12 - 22050% yielding or
40% tensile
23/05/1957
[MPa] [MPa] % [MPa] (the lowest)
Yielding Stress Tensile Strength Maximum ElongationProposed max working
stress
24
(a)
(d)
(b)
(c)
Figure 1.1. Type of steel ribbed bars analyzed during the data collection: (a)Thor steel; (b) RUMI steel;
(c) star shaped steel; (d) ribbed bar
(a)
(b)
Figure 1.2. Analysis of test certificate produced in 1962 by official laboratory in Pisa: (a) grouping by
bar type; (b) grouping by steel qualities.
The testing certificates were collected and statistically analyzed in order to generally gives a picture of
original mechanical properties of steel reinforcing bars: mean, standard deviation and fractile values of
yielding stress, tensile strength, elongation and hardening ratio, figure 1.3, were determined quite one
thousands of certificates. For example Aq42 steel presented a mean values, in the 1962, equal to 344
MPa and a standard deviation and a lower (5%) fractile respectively to 58.52 and 277 MPa. The
elongation at that time was measured at fracture and so it is higher than elongation usually recorded for
modern steel (TempCore steels); mean value and standard deviation were equal to 27.15% and 4.73%.
This investigation was extended also to other years, giving similar results: the mechanical properties of
steel reinforcing bars as yielding stress, tensile strength and elongation were at the origin (virgin state)
satisfactory, especially for the elongation that was relevant.
Moreover, it was interesting also to look at table 1.1. where a proposed working stress is reported: this
stress was adopted during the structural design, suggesting that a safety (i.e. limited knowledge) factor
equal to 2 was always considered.
Certainly, also if the original mechanical properties seemed to be satisfactory, the actual properties of
steel reinforcement inside concrete members were and are different, modified by the environmental
conditions all around the concrete building.
For this reason, some real samples of steel reinforcing bars were taken from reinforced concrete
building to be demolished. In particular, one building located at Villafranca in Lunigiana (LU) was
demolished and from the ruins, figure 1.4.a, some steel samples were taken: RUMI steel bars; on the
other side, a building was demolished inside ILVA Genova plant during revamping works, figure 1.4.a,
and many rounded bars were taken.
62,49%
37,51%
Smooth rebars
Shaped Rebars8,93%
29,90%
39,29%
17,53%4,35%
High elastic limit Aq42 Aq50 Aq60 Ordinary
25
(a)
(b)
Figure 1.3. Statistical analysis on 1962 production, Aq42 steel: (a) yielding stress, (b) elongation at
fracture.
(a) (b)
Figure 1.4. Demolished buildings: (a) pillars of Villafranca building; (b) Workers Union building
The tensile tests executed on steel reinforcement bars allowed to determine the actual mechanical
properties and it was observed in all cases, only two case as example are reported in figure 1.5, the
resistance was not affected at all while the elongation was deteriorated in some cases reduced by 50%.
(a)
(b)
Figure 1.5. Tensile testing of steel bars sampled from demolished buildings: (a) RUMI steel – end of
‘60s; (b) rounded bars – ‘20s.
Contemporary to the mechanical characterization, also chemical investigation was performed in order to
check the correspondence between chemical composition and mechanical properties and also to check
the weldability of sampled steel bars, see table 1.2. for 6 rounded bars. The tests showed that in general
old bars were always weldable and that it could be possible to use chemical analyses and hardness tests
to assess materials without sampling entire bars, see figure 1.6.
0%
5%
10%
15%
20%
25%
30%
35%
217 235 271 308 344 381 417 453 490 526 563 599 635Yielding Stress [MPa]
Pro
ba
bili
ty o
f o
bse
rva
tio
n [
%]
0%
5%
10%
15%
20%
25%
30%
5 7 10 13 16 19 23 26 29 32 35 39 42Elongation [%]
Pro
bab
ility
of o
bse
rvatio
n [
%]
26
Table 1.2. Chemical, metallographic and mechanical properties compared.
(a)
(b)
Figure 1.6. (a) correlation between tensile strength and Mn content; (b) linear regression between
mechanical properties (measured) and a possible chemical-data based model
The last part of investigation on mechanical properties regarded the assessment of reliable or real value
of concrete strength; the vulnerability assessment carried out in Italy by regional service made available
hundreds of compressive tests, figure 1.7, from which values of mean and standard deviation were
obtained.
(a)
(b)
Figure 1.7. (a) compressive tests on small cylinder; (b) statistical evaluation of the results.
It is evident that the concrete is the weakest material having many samples below 10 MPa. So two main
vulnerabilities at material levels are the elongation of steel bars and the low strength of concrete.
Sample
No.
Hardness
HV
PERLITE
percentageC Mn Si Ni Cu Sn Cr V Re (fy) Rm (fu) Agt
[N/mm2] [N/mm2] [%]
CS1 171 22.98 0.24 0.637 0.188 0.168 0.357 0.046 0.098 0.001 380.41 545.45 18.81%
CS2 159.4 19.09 0.225 0.633 0.199 0.164 0.387 0.0376 0.096 0.0011 413.46 570.16 17.55%
CS3 161 18.75 0.244 0.646 0.197 0.167 0.397 0.0449 0.098 0.0012 363.21 510.85 15.94%
CS4 177.3 19.88 0.154 0.787 0.28 0.152 0.357 0.0544 0.079 0.0018 360.84 500.01 19.76%
CS5 194.2 36.02 0.328 0.896 0.227 0.111 0.285 0.0442 0.099 0.0027 407.83 612.15 14.17%
CS6 162.7 23.28 0.226 0.632 0.196 0.163 0.389 0.0386 0.096 0.001 - - -
Reinforcing bar samples
Correlazione Mn-fu
0
100
200
300
400
500
600
700
0.00 0.10 0.20 0.30 0.40 0.50 0.60 0.70 0.80 0.90 1.00
Contenuto di Manganese
Te
nsio
ne
di ro
ttu
ra fu
[N
/mm
2]
300
320
340
360
380
400
420
440
460
0 100 200 300 400 500
Measured yielding [N/mm2]
Estim
ate
d y
ield
ing
[N
/mm
2]
HVDNiSiMnCrCSnCuBAR 0000e
0
2.5
5
7.5
10
12.5
15
17.5
20
22.5
25
0 26 69 111 154 197 239 282 325 367 410 452 495 538 580Rck: Cubic strength [kg/cm
2]
Experim
enta
l observ
ations
Experimental observations
Log-normal distribution
Normal distribution
27
2. Performance based design (PBD) framework
2.1. Main concepts on Performance Based Earthquake Engineering
Performance-Based Earthquake Engineering (PBEE) implies design and assessment of a building
whose performance complies with objectives expressed by stakeholders (owner, user, and society); in
particular, PBEE considers multi-performance conceptual methodologies where expected structural
response of construction must be calibrated for different levels of actions (i.e. seismic action).
According to this general approach, PBEE is based on the definition of multiple structural performance
levels, identified as damaging levels in structural members or secondary members, reached when the
structure is subjected to multiple levels of earthquake intensity (i.e. peak ground acceleration).
Considering also the operational aspects related to the practical application of the conceptual framework
of PBEE, denominated Performance Based Assessment (PBA), it is possible to individuate the general
flowchart related to the seismic retrofitting or upgrading of existing constructions:
(a) selection of earthquake intensities related to the hazard model associated to chosen area;
(b) definition of performance levels expected from entire building;
(d) knowledge of existing construction to be retrofitted through characterization of structural
vulnerabilities and material properties;
(c) crossing matrix between hazard levels, performance levels and information from structural
knowledge in order to choose a design strategy for the retrofitting;
(f) selection of the intervention technique on the basis of design strategy and structural
knowledge;
(e) execution of numerical analyses in order to evaluate the response of structural model of
existing construction when subjected to different earthquake intensities;
definition of retrofitted structure, cost estimation and evaluation of economic convenience.
Figure 2.1. Performance Based Engineering framework and Performance Based Assessment sub-
framework.
It is clear that the choice of the design strategy – point (C) – to be followed for the retrofitting or for the
upgrading of a structural system and hence of a construction should be suitable addressed for the
considered particular case, on the basis of detected vulnerabilities in existing construction. It is so
obvious that PBEE is the natural operative framework in which retrofitting projects should be
developed and suitably addressed on the basis of initial design information.
As presented in the previous, PBA is the operative core of the PBEE in which modeling techniques,
numerical analyses and technical aspects are interconnected in order to arrive to the final intervention
techniques, while other aspects represents the general set-up of the retrofitting that fix design options
coming from safety levels and minimum structural performance imposed and/or requested by Public
Authorities or, more in general, by stake-holders.
(A) Definition of hazard levels (B) Definition of performance levels
(C) Choice of design
strategy
Performance of entire building structure
(D) Knowledge of existing
construction (deficiencies,
materials,…)
(E) Structural model/ analysis/
evaluation of the response
(F) Selection of the
intervention technique
Retrofitted building
PBA
Evaluation of costs related to
intervention techniques
PBEE
29
The scope of the analysis and the application of PBA inside PBEE is to individuate tools or to give
practical indications in order to contextualize each operative step, only conceptually presented in the
figure 2.1, arriving to the definition of a practical framework for PBEE for retrofitting on the basis of
actually gained knowledge.
In fact, it is worth noting that several standards and codes have been issued during last years for the
regulation of retrofitting and upgrading of structural systems and many of them have been defined
considering a performance based framework. Anyway, in some cases, standards do not furnish to the
designers all necessary rules or guidelines and it would be of a certain interest the integration of
different codes, suitably analyzed and studied, in the operative flowchart depicted in the figure 2.1
properly selecting parts, rules or guidelines to be integrated. So, this chapter would like to discuss
relevant aspects on PBEE and PBA treated in existing codes and standards in order to define the
STEELRETRO procedure to be adopted during the analysis of different retrofitting technique in order
to give them a common playground in which the results are unbiased and so comparable.
2.2. Analysis of existing PBEE Framework
2.2.1. Building performance objectives Performance required from a retrofitted/upgraded construction consists of one or more rehabilitation
goals identified with damage levels occurred to all elements realizing and contained in the construction.
The definition of such performance objectives (i.e. performance of entire building; design targets) is
crucial for the evaluation of the structural safety (i.e. acceptance criteria for evaluating the performance
are tied to performance of structural members and non-structural members): these targets are assumed
during the design of the rehabilitation intervention fixing different damage levels expected from
structural and non-structural elements for different levels of performance.
The FEMA 356 defines four global performance description – Operational, Immediate Occupancy, Life
Safety, Collapse Prevention –, while VISION2000 makes reference to four levels – Fully Operational,
Operational, Life Safe, Near Collapse; the Italian Code for Constructions DM2008 has, as the VISION
2000, four performance levels – Operational, Damage Limitation, Life Safety, Collapse Prevention – ,
while EN1998-3 identified the following three performance levels (targets) – Damage Limitation,
Significant Damage and Near Collapse. In order to be operative, each performance level must be
associated to expected maximum damage levels in the elements, identified as the performance
objectives and so, in general, the building performance where damaging effects in structural and non-
structural members are coupled.
For example, FEMA356 proposes the following building performances described in a general sense:
Operational Level (1-A): minimal or no damage to structural and nonstructural components;
building is suitable for its normal occupancy and use; possibly with some nonessential systems not
functioning; extremely low risk to life safety.
Immediate Occupancy Level (1-B):minimal or no damage to structural elements; only minor
damage to their nonstructural components; following a major earthquake, nonstructural systems
may not function; immediate re-occupancy of the building is possible; some cleanup and repair,
and restoration of utility service; risk to life safety at this performance level is very low.
Life Safety Level (1-C): experience extensive damage to structural and nonstructural components;
repairs may be required before reoccupancy and may be deemed economically impractical; risk to
life in buildings is low.
Collapse Prevention Level (5-E): no consideration of nonstructural vulnerabilities; significant
hazard to life safety resulting from failure of nonstructural components; building itself does not
collapse, gross loss of life should be avoided.
While, EN1998-3 presents three general building performance (i.e. limit states) indentified with
following overall descriptions:
Damage Limitation (DL): light damages; structural elements prevented from significant yielding;
non-structural components with distributed cracking; economic convenience for the reparation;
structural parts without any repair measure.
Significant Damage (SD): significant damages; residual lateral strength and stiffness; vertical
elements capable of sustaining vertical loads; damages in non-structural components not out-of-
plane failed; moderate permanent drifts is present; structure can sustain after-shocks of moderate
intensity; reparation of structure is uneconomic.
30
Near Collapse (NC): heavy damages; low residual lateral strength and stiffness; vertical elements
still capable of sustaining vertical loads; collapsed non-structural components; large permanent
drifts; the structure is near collapse and would probably not survive another, even moderate,
earthquake.
VISION2000 and DM2008 identified building performance which qualitative description is quite
similar to the description proposed by FEMA and by EN1998; the correspondence between damage
levels is also reported in the table 2.1 in which their qualitative description has been reported to the
structural and non-structural performance matrix proposed by FEMA.
2.2.1.1. Combination of structural and non-structural damage levels for the
definition of admissible performance levels The performance objectives of a buildings should be agreed between designers and stake-holders on the
basis of technical, economic and management aspects that many times are not clear at the beginning of
retrofitting process. For this reason, it would be useful to adopt during the design process a performance
matrix in which damage levels expected from structural members (S1…S6) are correlated to damages
expected in non-structural elements (N-A…N-E), see table 2.1 taken from FEMA and specialized for
the analysis carried out. Building performance are presented as suitable combinations of both, through
the alphanumeric code, and it is evident that too disproportioned expected performance between
structural and non-structural elements should not be not recommended in order to reach an optimized
design. Many combinations are possible with the common agreement between designers and stake-
holders that can be different from those proposed by standards, but that can have an high added values
for stake-holders in terms of safety, of course, and of economic convenience.
On this basis, FEMA approach is more complete because treats the building performance as a
composition of single performance, giving to designers and stake-holders quite free-hand in the
definition of expected behavior, suggesting anyway performance levels commonly agreed by
technicians and stake-holders. For sake of completeness, performance levels qualitatively proposed by
other standards have been inserted in the matrix of figure 2.1, showing a substantial agreement with
proposed qualitative damage levels inside members for Life Safety and Collapse prevention (i.e. design
for human life preservation) while some differences are in the definition of performance under frequent
earthquakes where economic losses have a more relevant role on the judgment of a designed solutions.
Table 2.1. Performance matrix for the definition of global building performance.
Non-structural
Performance
Levels
S-1
Immediate Occupancy
S-2
Control Damage
Range
S-3
Life Safety
S-4
Limited Safety
Range
S-5
Collapse Prevention
S-6
Not Considered
N-A
Operational
1-A
Operation(FEMA)
Fully Operational(VISION2000)
OLS(DM2008)
2-A N.R. N.R. N.R. N.R.
N-B
Immediate
Occupancy
1-B
Immediate Occupancy(FEMA)
DL(EN1998)
Operational (VISION2000)
2-B
DLS(DM2008) 3-B N.R. N.R. N.R.
N-C
Life Safety1-C 2-C
3-C
Life Safety(FEMA)
SD(EN1998)
Life Safe(VISION2000)
LLS(DM2008)
4-C 5-C 6-C
N-D
Hazards ReducedN.R. 2-D 3-D 4-D 5-D 6-D
N-E
Not ConsideredN.R. N.R. N.R. 4-E
5-E
Collapse
Prevention(FEMA)
NC(EN1998)
Near Collapse(VISION2000)
CPLS(DM2008)
N.R.
Structural Performance Levels/Ranges
N.R.: Combination of structural/non-structural performance not recommended
Grey cells: Admissible Buildining Performance
(EN1998): Limit States defined by EN1998; (VISION2000): Performance levels; (DM2008): Limit states defined by Italian Code
31
2.2.2. Earthquake hazard level The full application of PBEE needs also the definition of seismic action levels coherent with the hazard
and each level of seismic action must be correlated with an expected building performance defined
according to a matrix method. In general, the intensity of seismic action is fixed in terms of Medium
Recurrence Interval (MRI) or, alternatively, as Probability of Exceedance (PE) in a fixed time interval,
see table 2.2. In particular, all codes fixed a return period equal to 475 year for the rare earthquakes
with a PE equal to 10% in 50 years, while for low intensity earthquakes proposed PE there are some
differences, due probably to same aspects presented in §2.2.1.1. It is also interesting to note the extreme
differences for the very rare earthquakes: FEMA 356 and EC8-3 considers a return PE equal to 2% in
50 years much more demanding than PE considered in VISION2000 and DM2008. Moreover,
VISION2000 and DM2008 with very low MRI for the frequent and occasional earthquakes could lead
to structural solution potentially subjected to disproportioned damages, and hence to relevant economic
losses, that could be effectively limited assuming 225 years as MRI for occasional earthquakes.
Obviously, the complete definition of earthquake levels needs the choice of a hazard model, dependant
from the seismic-genetic characteristic of the area in which the intervention technique has to be applied.
In particular, for the purposes of the project, the earthquake loads are chosen according to a moderate
seismic hazard (largely diffused among European countries) and not to the highest or lowest levels of
seismic hazard in Europe. This choice certainly is going to affect results, but mean reference values and
criteria, useful for suggestions also in extreme situations, can be obtained from this assumption.
For EN1998, the quantification of the earthquake level is given by the maximum ground acceleration ag
on a Type “A” outcrop ground with flat surface, and it has been assumed as an appropriate intensity
measure of the seismic excitation due to its wide acceptance by designers. An hazard curve is given in
order to define earthquake levels for each performance objective (i.e. target or limit state) that has to be
considered according to the Performance Based Design Criteria.
The Ground Acceleration hazard curve relative to a Type A ground and flat topographical surface in
Assisi (Italy) is assumed as reference. The following interpolation rule1 can be used to obtain ag values
associated to Mean Return Periods different from those specified in the table:
1
g 2 R R2g g1
g1 R1 R1
a T Tlog a = log a +log ×log × log
a T T
(4.1)
where TR is the return period (MRI) for which ag has to be determined, and ag1 and ag2 are the maximum
ground acceleration associated to the return period TR1 and TR2 with TR1<TR<TR2.
In order to find the design response spectra, shapes and amplification factors due to local effects from
EN1998 are used, because its parameters are representative for the European seismicity: Response
Spectrum Type 1 with 5% damping; the pick ground acceleration ag for the Live Safety Performance
Level equal to 0.23 g; Ground Type B: S = 1.2, TB = 0.15 s, TC = 0.5 s, TD = 2.0 s.
Figure 2.2 Mean Return Periods (TR, MRI) and expected maximum ground acceleration ag.
1 The interpolation rule is consistent with the hazard curve shapes derived from the usual methods for evaluating
seismic hazard.
ag (g)
TR (years) .
TR ag
[years] [g]
30 0.072
50 0.094
72 0.111
101 0.128
140 0.146
201 0.168
475 0.230
975 0.292
2475 0.390
3050
72101
140201
475
975
2475
0.00
0.05
0.10
0.15
0.20
0.25
0.30
0.35
0.40
0.45
10 100 1000 10000
32
2.2.3. Design Strategies The selection of the performance objectives (expected minimum capacity of the structure) and its
coupling with the intensity of the seismic input, related to the mechanical demand imposed on structural
and non-structural members, define the design strategy for the retrofitting/rehabilitation interventions.
In particular, such phase is the key for PBEE application in which costs and feasibility are faced with
the benefit to be obtained in terms of improved safety in the event of future earthquakes. According to
PBEE, such phase of the designing should foresee the cooperation between designers and stake-holders
for defining the most appropriate strategy: the strategies proposed by analyzed standards are
summarized in the table 2.4 and are represented in the figure 2.2 using a (damage level, MRI) domain.
In particular, from figure 2.2, it is possible to qualitatively appreciate the difference among proposed
strategies.
A qualitative admissible domain for each strategy is represented by the plane portion contained on the
left side of each curve; considered approaches present remarkable differences due to the high difference
in the MRI associated to damage levels correspondent to the structural collapse proximity and to the
structural collapse prevention, see figure 2.2. Moreover, it is also worth noting the high differenced
between Italian Code DM2008 and other standards in correspondence of the Operational performance
objective (identified by level 3).
Ear
thquak
e H
azar
d
Lev
el
Frequency FEMA 356
SEAOC
Vision 2000 EC8-3 DM2008
MRI PE MRI PE MRI PE MRI PE
Frequent 72 50%/50 43 50%/30 - - 72 50%/50
Occasional 225 20%/50 72 50%/50 225 20%/50 140 30%/50
Rare 474 10%/50 475 10%/50 475 10%/50 475 10%/50
Very Rare 2475 2%/50 970 10%/100 2475 2%/50 975 5%/50
Table 2.2 Earthquake hazard level; PE - Probability to exceed; MRI - Medium recurrence interval
Table 2.3 Comparison of the design strategies proposed by different standard.
Damage level 1-A 1-B 2-B 3-C 5-E
Seismic input
50%/30ys
Vision2000
-
-
-
50%/50ys
Vision2000
DM2008
FEMA356
-
30%/50ys
-
DM2008
-
-
20%/50ys
-
-
FEMA356
EN1998-3
10%/50ys
Vision2000
DM2008
FEMA356
EN1998-3
5%/50ys
Vision2000
DM2008
-
-
2%/50ys
-
-
FEMA356
EN1998-3
33
2.2.4. Knowledge of the structure to be retrofitted One of design task during each retrofitting project is made by the collection of the information about the
existing structure, reported in the PBEE flow-chart. This part presents many difficulties from a practical
point of view related to the collection of design data, the real mechanical properties of the building
materials and the history of the building. All these aspect are accurately considered in FEMA356,
DM2008 and EN1993-8, defining appropriate coefficients for modeling the uncertainties level of
information about the existing construction used during the design of the retrofitting intervention. In
such research project the focus has been set on performance of intervention technique and definition of
improved technique for retrofitting, considering the structures as base cases on which applying such
techniques. For this reason, the analysis of the uncertainties of materials properties has not been
considered, assuming the complete knowledge of the structure and its details and the reliability of
mechanical properties of the material (actual values). An analysis of such kind of structural
vulnerability has been executed and its results have been presented in §1.2.
2.3 Performance Based Assessment The options fixed inside the PBEE gives the general framework, the boundaries, in which the design
must be carried out using modeling techniques of structural systems, analysis method for simulating the
structures behavior under earthquake excitation and assuming acceptance criteria for interpreting the
results and judging the fulfillment of expected building performance objectives. In the following a short
description of the methods and techniques is given.
2.3.1 Analysis methods, modeling and acceptance criteria Structures are usually designed to resist earthquake action in the inelastic range of response. The
dynamic nature of earthquake action, and the possible inelastic structural response, implies a nonlinear
dynamic analysis procedure on a three-dimensional model or different bi-dimensional models of the
building structure, depending from its complexity. There are five generally adopted analysis procedures
used for seismic analysis of structures (FEMA 356; Eurocode 8) presented below in a hierarchical
order:
lateral force method (linear static procedure);
response spectrum analysis;
linear time-history analysis;
nonlinear static procedure (pushover analysis);
nonlinear time-history analysis.
Each analysis procedure shall be applied taking into consideration modeling techniques that have to be
consistent with procedure; in particular, FEMA356 and EN1998 consider the same analysis procedures
and give similar indications about the general modeling hypothesis to be considered; therefore, FEMA
indications about the local modeling of the structural members and the role of the secondary members
are more completed and exhaustive than approach followed by EN1998. For such reason, the
indications from FEMA356 for the modeling of the structural systems (modeling parameters) have been
considered in the execution of all models developed during the project.
2.3.1.1. Modeling Parameters and Acceptance Criteria The analysis method allow the evaluation of the demand imposed on the whole structure and on each
single component; in particular, all primary and secondary components shall be capable of resisting
force and deformation actions considering for each of them the applicable acceptance criteria of the
selected performance level. In general, the criteria can be differentiated between those applicable to
brittle elements and those applicable to ductile elements. According to this, all actions, as reported in
FEMA356 and also in EN1998, shall be classified as either deformation-controlled or force-controlled
using the component force versus deformation curves, assuming representative curves as those depicted
in figure 2.3.
34
Figure 2.3 Generalized Component Force-Deformation Relations for Depicting Modeling and
Acceptance Criteria
Elastic stiffness and values for the parameters a, b, c, d, and e that can be used for modeling
components are given; in particular, factor and formulas for those parameters adopted for all the
simulations will be directly presented during the definition of numerical models employed for the
simulations carried out in §4.
The acceptance criteria for brittle and ductile primary members (P) and for secondary members (S)
corresponding to the target Building Performance Levels have been selected according to the adopted
retrofitting strategy, according to the matrix schemes presented in figure 2.1 and 2.4 and assuming
engineering demand parameters (i.e. forces, rotations, displacements,…) from those proposed inside
FEMA356 framework.
2.3.1.2. Linear – Elastic Analysis
2.3.1.2.1 Lateral force method In the case of assessment of existing structure (FEMA 356; EN1998-3), the lateral forces are
determined based on the elastic response spectrum, and not on the design one (reduced by a behavior
factor q determined on the basis of the knowledge of the structural system).
2.3.1.2.2 Modal response spectrum and linear time-history Response spectrum procedure is a generalization of the lateral force method, accounting for more than
one mode of vibration in determining seismic response of the structure.
2.3.1.2.3. Acceptance criteria for linear analysis If linear procedures are used, capacities for deformation-controlled actions shall be defined as the
product of m-factors (modification factor used in the acceptance criteria of deformation-controlled
components or elements, indicating the available ductility of a component action) or q-factor, and
expected strengths, QCE. Capacities for force-controlled actions shall be defined as lower-bound
strengths, QCL.
2.3.1.3. Non-linear Analysis
2.3.1.3.1. Static – Pushover Nonlinear static analysis is an analysis technique in which the non-linear properties of the structures are
modeled, not considering cyclic degradation, being a static method. This method has the value of being
less demanding from a computational point of view than dynamic methods, but the assessment of the
demand imposed by seismic action to the structure must be carried out analyzing results. In particular, it
is necessary to transform, using a procedure (e.g. coefficient method, capacity spectrum method -
FEMA 356; or the N2 method – EN1998), the capacity of the structure in a reference capacity curve
and then comparing it with the seismic demand imposed by the hazard model (i.e. response spectrum):
this approach is generally addressed as the response spectrum approach combined with the non-linear
static analysis. Moreover, it is worth adding that this method is continuing to acquire more importance
into the design practice, being well accepted by engineers.
35
2.3.1.3.2. Dynamic – Time-history Nonlinear time-history analysis represents the most advanced method of analysis for evaluation of
seismic response of structures. Such analysis method needs the definition an accurate structural
modeling in which cyclic behavior of the materials, all the non-linear phenomena and, possibly, cyclic
degradation are taken into account. Despite the complexity of the modeling, whose applicability in the
practice still need to be confirmed, the nonlinear time-history analysis provides results directly
comparable with acceptance criteria without the necessity of using additional procedures as for non-
linear static method. On the contrary, this method is largely time-consuming and the correct choice of
the seismic input to be adopted during the analyses is still a matter of discussion.
2.3.1.3.3. Acceptance criteria for nonlinear analysis If nonlinear procedures are used, component capacities for deformation-controlled actions shall be
taken as permissible inelastic deformation limits, and component capacities for force-controlled actions
shall be taken as lower-bound strengths, QCL.
2.3.2 Analysis of Non-linear static procedure The most appropriate approach seems to be a combination of the nonlinear static (pushover) analysis
and the response spectrum approach, due to its level of modeling accuracy and to its well acceptance
into design practice. Examples of such an approach are: Capacity spectrum method (CSM); Nonlinear
static procedure. The procedure can be summarized according to the following steps:
1. modeling of the structural members and secondary elements (if relevant) using non-linear
technique for considering material and geometrical sources of non-linearity; (figure 2.4.a)
2. execution of pushover (non-linear analysis) subjecting the structural model to one or more set
of horizontal forces, schematizing seismic inertia forces; (figure 2.4.b)
3. individuation of the collapse mechanism of the structure (Soft storey, loss of ductility capacity
in a column, column shear failure, beam-column joint shear failure) for stopping the pushover;
4. definition of the simplified structural model (equivalent bilinear SDOF model - base shear
versus displacement of the participant mass) and of its capacity curve; (figure 2.4.c)
5. evaluation of the seismic demand curve in terms of maximum acceleration and displacement
imposed to the elastic SDOF; (figure 2.4.d)
6. comparison between structural capacity and seismic demand. (figure 2.4.e)
(a)
(b)
(c)
(d)
(e)
Figure 2.4 Complete procedure for applying the non-linear static analysis method and interpreting the
results in terms of capacity and demand.
36
All analysed non-linear procedures presented the same steps and the similar approaches, whose
differences are located in the point 5, of course, due to the different reference hazard curves at which is
code refers, in the point 4, because the schematization of the capacity curve can be executed adopting
different approached and in the point 6, due to the different methods that can be employed for obtaining
the demand diagram. FEMA356 and EN1998 in particular employ different approaches for the
execution of the non-linear procedure: the capacity spectrum method the former and the N2-method the
latter, which act, as expected, in correspondence of the point 4, 5 and 6 of the procedure.
From the analysis of the standards, the N2-method proposed by EN1998 did not present many
disadvantages or weak points respect to FEMA356 approach, unlikely observed before for the structural
modelling and the acceptance criteria. Moreover, N2-method is a technique largely applied across
European countries and many National standards have already implemented it; so it appeared as
appropriate coupling EN1998 procedure with FEMA356, in order to define the PBA tools necessary for
performing the structural assessment inside the PBEE framework.
It is also worth underline again that the application of non-linear procedures, after the analysis of
different standards, appears quite mandatory for having an acceptable assessment level, that linear
techniques cannot guarantee. Determining the nonlinear structural behavior allows significant savings
in seismic retrofit applications for example. Figure 2.5.a shows the typical top displacement vs. base
shear curve obtained from nonlinear pushover analysis of buildings.
Using this curve alone, one can perform a preliminary evaluation of the structure’s seismic safety by
comparing its capacity with the seismic demand determined using the equivalent static load method
described in seismic codes. A better performance evaluation can be performed by converting both the
capacity curve and the seismic demand spectrum to the acceleration-displacement response spectrum
(ADRS) format formed as a relationship of spectral displacement vs. spectral acceleration as shown in
figure 2.5.b.
Figure 2.5 Seismic safety evaluation of buildings using nonlinear analysis
The intersection of the capacity and demand curves shown in figure 2.5.b is called the performance
point of the building, as defined in EN1998. If the performance point is located in the initial portion of
the capacity curve where the inelastic deformations are not significant the performance level of the
building is Immediate Occupancy, which is self-explanatory.
2.4. Choice of the intervention technique The definition of a procedure would allow to testing different intervention techniques in order to
evaluate their performance; moreover, the possibility of repeating different simulations using the same
procedure and the same demand will allow comparing tested solutions and applying additional analysis
criteria to obtained results. In such a way the solution could be determined according to a multi-criteria
method, employing, for example, following criteria.
Technical aspects: Reversibility of intervention, Compatibility, Durability, Corrosion, UV
resistance, Aging, Creep, Local conditions, Availability of material/device, Technical
capability, Quality control
Structural aspects: Structural performance (Strength, Stiffness, Ductility, Fatigue), Response
to fire, Sensitivity to changes of actions/resistances e.g. seismic action, temperature, fire, soil
conditions, Accompanying measures, Technical support (Codification, Recommendations,
Technical rules), Installation/Erection e.g. availability/necessity for lifting equipment
37
Economical and sustainability aspects: Costs, Design, Material/Fabrication, Transportation,
Erection / Installation / Maintenance, Preparatory works
2.4.1. Structural performance based validation Choice of one or another strengthening technique is a multi-criteria problem. One has to select which
technique matches better with assembly of validation criteria. The solution will represent always a
rational compromise among different criteria, because the one criterion based optimization leads, in
general, to an unacceptable choice.
The capacity curve of the strengthened structure, Cs, generally has a higher slope and peak compared to
the capacity curve before strengthening, Cu. In Figure a theoretical situation is considered. Due to the
increased stiffness, which translates into a decreased fundamental period, the seismic demand on the
structure is also increased, as shown by the demand curve for the strengthened structure, Ds, compared
to that for the unstrengthened structure, Du. Although the increase in capacity is partly alleviated by the
increase in seismic demand, the overall performance of the structure is improved as shown by the
locations of the performance points on the spectral displacement axis for before and after strengthening.
(a)
(b)
(c)
a) Effect of structural strengthening; b) Effect of deformation enhancement; c) Effect of enhanced
energy dissipation
Figure 2.6. Analysis of the concept of strengthening solutions
After, depending by the hierarchy between the demand in strength, stiffness and ductility, and also
considered the other complementary criteria of previous section, the final decision can be taken.
2.4.2. Technical aspects The technical aspects related to an intervention techniques are in general related to the boundary or
environemntal conditions that the design of an retrofitting project must take into considerations:
accessbility for installing retrofitting elements; feasibility of partial demolitions; checking if the
intervention could be reversible or not; chekc about the accessing to the foundation level for an eventual
retrofitting; and so on. All these aspects have to be taken into consideration before starting with the
operative design and the application of PBA proccedure; in particular, here an extensive pre-analysis of
all more diffused and known techniques has been executed (§3) using a matrix based approach in which
all technical aspects have been investigated. According to this pre-selection, only the technique
potentially adapt to be employed have been investigated more in the details using PBA.
2.4.3. Economic aspects The economic impact of solutions analyzed using PBA have been determined for some of those tested
in §5; in particular, the variables considered for this assessment are related to the amount of materials
employed for the retrofitting techniques and (where possible) the estimation of the quantity of
demolished materials (i.e. infill walls).
These voices have been estimated for some applied techniques and then economically valorized in order
to appreciate also this impact of the seismic retrofitting and considering this final choice criteria inside
the complete PBA procedure.
38
2.5. Complete PBD framework assumed in the project (PBEE/PBA) Several guidelines concerning performance based seismic evaluation and retrofit of existing buildings
are available. Among these, the most important are FEMA 356 "Guidelines for Seismic Rehabilitation
of Buildings", ATC 40 "Seismic evaluation and retrofit of concrete buildings" and Eurocode 8-3
"Design of structures for earthquake resistance. Part 3: Strengthening and repair of buildings".
This report overviews PBD procedures available in the above guidelines and in literature. Eurocode 8-3
does not offer a complete procedure that can be readily applied to evaluation of an existing structure
and its retrofit solution, without additional knowledge and expertise. Therefore, this report emphasizes
the provisions of FEMA 356, adopting it as the suggested document to be adopted within
STEELRETRO project, in order to have a common basis for evaluation studies undertaken by different
partners. However, several amendments are suggested in order to adapt provisions of FEMA 356 to the
specific objectives of STEELRETRO project and European practice. One of these relates to building
performance objectives to be adopted in the project. Considering that multiple performance objectives
are available in FEMA 356, it is suggested to choose the ones shown in Table within the
STEELRETRO project.
Building performance level
Immediate
Occupancy
Life Safety Collapse
Prevention
Ear
thquak
e
Haz
ard L
evel
Occasional –
MRI = 225 years - -
Rare –
MRI = 474 years - -
Very Rare –
MRI = 2475 years - -
Table 2.4 Building performance objectives for use in STEELRETRO project
Characterization of seismic action is another issue that is believed to need adaptation. It is suggested to
adopt elastic response spectra used in European practice (Eurocode 8-1), adjusted to hazard levels from
table 2.4. Elastic response spectra parameters (peak ground acceleration, soil type, response spectrum
type) to be used for estimation of target displacement within nonlinear static analyses and ground
motion records to be used in nonlinear time-history analyses have been established according to figure
4.8. Moreover, it has been also fixed that the PBA is based on nonlinear static procedure for evaluation
of existing buildings and retrofitting solutions; finally, it has been also fixed the adoption of the
procedure described in annex B of EN1998-1 (N2-method), as being more familiar in European
practice.
39
3. Preliminary analysis of existing techniques Choice of a specific strengthening technique for an existing building is a multi-criteria problem,
involving structural, technical, cultural, social, and economic and sustainability aspects. The designer
has always several solutions at his disposal before starting the design process and it is unrealistic
thinking to analyze all the solutions using numerical analyses which computational and time demand is
high and that can be effectively employed for one retrofitting project in a limited number of cases.
It is without doubt clear that a technical solution should represent always a rational compromise among
different criteria, because a single criterion based optimization leads, in general, to an unacceptable
choice. So, it would be worth considering all the aspects relevant for the conceiving and the design of a
retrofitting as structural aspects, technical aspects and economic aspects.
Regarding the structural aspect, the intervention strategy has to make a choice between increasing the
strength of building or to enhance the deformation capacity (e.g. ductility) or a good balance of both.
The attempt to increase the resistance leads in most of the cases to significant increase in stiffness and
consequently to increasing seismic force and demands. Anyway, the major problem in structures with
limited ductility is deformation capacity (see figure 3.1). Modern retrofitting strategies insist in the use
of intervention techniques optimized for the structural pathologies and intrinsic vulnerabilities; these
aspects necessary require an optimization process for the increasing of performance in the retrofitted
structure. It is also worth noting that at the beginning of the design process the considerations and the
judgments about a retrofitting technique can be qualitative and addressed to the generic type of the
building that shall be retrofitted.
Figure 3.1 Enhance the deformation capacity of the building
With such perspective, it has been decided to define a typological approach allowing a pre-selection on
the intervention techniques on the basis of general qualitative marking criteria, according to the general
scheme presented in figure 2. In particular, in the table 3.1 and 3.2 there are proposed a Decisional
Matrix and a typological form for the selection and the validation of rehabilitation method, inspired by
the typological scheme presented in figure 3.2.
Figure 3.2 Data concerning with intervention techniques using typological analysis
Part 1 – T.C.
Techniques Classification
Part 1 – T.C.
Techniques Classification
StiffnessStiffness
ResistanceResistance
DuctilityDuctility
Typological analysis of an intervention techniqueTypological analysis of an intervention technique
Part 2 – N.S.P.
Non Structural Properties
Part 2 – N.S.P.
Non Structural PropertiesPart 3 – S.C.
Structural Classification
Part 3 – S.C.
Structural Classification
Amount of materialAmount of material
Technological aspectsTechnological aspects
Used space of existing
building
Used space of existing
building
DemolitionsDemolitions
Integration in existing buildingsIntegration in existing buildings
AccessibilityAccessibility
ReversibilityReversibility
MaintenanceMaintenanceMasonry Shear Wall
Cantilever
Masonry Shear Wall
Cantilever
R. C. Frames
Shear Walls
Dual Systems
R. C. Frames
Shear Walls
Dual Systems
Part 1 – T.C.
Techniques Classification
Part 1 – T.C.
Techniques Classification
StiffnessStiffness
ResistanceResistance
DuctilityDuctility
Typological analysis of an intervention techniqueTypological analysis of an intervention technique
Part 2 – N.S.P.
Non Structural Properties
Part 2 – N.S.P.
Non Structural PropertiesPart 3 – S.C.
Structural Classification
Part 3 – S.C.
Structural Classification
Amount of materialAmount of material
Technological aspectsTechnological aspects
Used space of existing
building
Used space of existing
building
DemolitionsDemolitions
Integration in existing buildingsIntegration in existing buildings
AccessibilityAccessibility
ReversibilityReversibility
MaintenanceMaintenanceMasonry Shear Wall
Cantilever
Masonry Shear Wall
Cantilever
R. C. Frames
Shear Walls
Dual Systems
R. C. Frames
Shear Walls
Dual Systems
41
The purpose of such matrix and its accompanying typological form is to schematize and formalize the
engineering judgment and the preliminary evaluation that are the starting point of each design process.
Such treatment (i.e. matrix approach) of subjective and objective data should be followed by the
designers because, before the final selection of the intervention techniques, it would be possible to re-
analyze all decisions and all judgments in a synthetized form in order to check the coherence of the
preliminary decisional process.
Moreover, this phase of the process contains most engineering judgment about the techniques, their
applicability and also qualitative expectations about the costs. So, the possibility of reviewing all these
information for a designer but also for public authorities that manage resources and funds for such
projects could be of a relevant interest. In fact, comparing this information (preliminary) with the final
results of a retrofitting process could create two positive aspects: the designers would continue to
improve their judgment and their skills and their designing/operative practice will be driven to a more
systematic approach; the public authorities could look into these database for preparing more
appropriate tender documents, focused on retrofitting of public/historical value estate and structured for
optimizing the necessary economic resources. In fact, the table 3.1 is very general and contains also
aspects that can be more precisely marked after a complete or a preliminary structural assessment of the
original existing construction or the one equipped with a retrofitting system. It is worth also noting that
the filling of this decisional matrix at the end of the design process (i.e. after cost estimation) could
increase potential benefits of such approach for the technicians and for the stakeholders, allowing a
direct analysis input/output of the design process also.
Table 3.1 Decisional Matrix condensing all
relevant aspects for a preliminary judgment of the
structural intervention technique. Legend for
scoring L = low, M = medium, H = high; Mark –
L (5-6), M (7-8), H (9-10)
Table 3.2. Typological form to be adopted with
the decisional matrix in the preliminary selection
of intervention technique – form filled for ring
beam technique for roof in masonry building
The matrixes previously proposed were used in order to organize all data coming from a typological
analysis carried out on a great number of intervention techniques using both bare steel and reinforced
concrete solutions. These forms were filled in order to arriving at the end of the evaluation process to
delineate some preliminary conclusions about the selection of the intervention techniques. The
investigation was carried out subdividing the analysis in separate interventions techniques groups, each
of them addressed to a different structural element: masonry walls; floors and roofs in reinforced
concrete buildings; foundations systems; vertical elements in reinforced concrete buildings and so on.
Some of analyzed techniques are briefly sketched in the figure 3.2 and, as an example two tables, filled
during the typological analysis, are reported in the table 3.3.
Structural aspects L M H Mark
Capability to achieve requested performance objective (after building evaluation!)
Compatibility with the actual structural system (no need of complementary strengthening or confinement measures)
Adaptability to change of actions seismic typology (near field, far field, T<>Tic, etc)
Adaptability to change of building typology
Technical aspects L M H Mark
Reversibility of intervention
Durability Operational Functionally and aesthetically compatible and complementary to the existing building
Sustainability Technical capability Technical support (Codification, Recommendations, Technical rules)
Availability of material/device Quality control
Economical aspects L M H Mark
Costs (Material/Fabrication, Transportation, Erection, Installation, Maintenance, Preparatory works)
Typological analysis of intervention (horizontal and vertical)
Techniques classification
Stiffening: Yes/No
Resistance: Yes/No
Ductility: Yes/No
Non structural properties
Amount of materials: -
Technological aspects: -
Used space: -
Demolition: -
Accessibility: -
Reversibility: -
Maintenance: -
Structural classification
Masonry:
Reinforced Concrete:
42
(a)
(b)
Table 3.3 (a) typological analysis on micro-piles intervention on foundations; (b) typological analysis
on horizontal bracings for floor/roof stiffening.
The application of the decisional tools (matrix and typological analysis table) to a large number of
intervention techniques commonly used in the practice allowed a preliminary marking and all results
coming from the investigation have been summarized in macro tables where analyzed interventions
have an associated synthetic judgment about its suitability. In particular it has been analyzed the
application and the suitability of different retrofitting techniques to the following elements:
Vertical elements in masonry buildings (walls); (table 3.4)
Flooring elements in masonry buildings; (table 3.5)
Roofing elements in masonry buildings; (table 3.6, 3.8 and 3.9)
Foundation elements in masonry buildings; (table 3.7)
Flooring and roofing elements in reinforced concrete buildings; (table 3.10 and 3.11)
Foundation elements in reinforced concrete buildings; (tables 3.12 and 3.13)
Vertical elements in reinforced concrete buildings (frame elements – global retrofitting). (table
3.14)
The tables presented in the following have been obtained summarizing all the detailed information
coming from the application of table 3.1 and table 3.2; in particular, the tables present a global
judgment about the applicability of analyzed techniques to fixed structural element or parts of the
structure: in table 3.12 for example the applicability of the techniques (i.e. global judgment considering
structural, feasibility and economic aspects) is considered while in the table 3.13 it has been reported
the improvement of failure mechanism of the structural sub-part. In table 3.4, as another example, the
techniques related to the wall masonry are reported summarizing the applicability of the system to
different walls and feasibility aspects. Such extensive work executed on all those different techniques
allowed the realization of a database from which some preliminary evaluation on various techniques
could be prepared.
Concerning masonry walls, analyzing the results it can be argued that techniques as steel tying, steel
pre-tensioning systems or steel strips presented a large applicability while other techniques based on
concrete or carbon/glass fibers presented some limitations. The same conclusions could be derived for
flooring and roofing systems in masonry buildings: the steel solutions resulted as advantageous respect
to concrete ones in terms of cost and applicability; moreover, the high prefabrication levels of steel
solutions guarantee a certain degree of reversibility of the intervention. The same conclusions come
from the tables summarizing the techniques for flooring and roofing systems in reinforced concrete
buildings.
Concerning the vertical systems in reinforced concrete frames, it was clearly recognized that all
analyzed techniques could be applied also to masonry buildings: steel bracing frames, dissipative
bracings and steel walls are techniques that can be easily applied to both systems.
The analysis of the retrofitting techniques for the foundation systems shows that the more performing
technique for upgrading and retrofitting was the micro-piling, applicable to reinforced concrete and
masonry buildings.
According to these considerations coming from the typological/feasibility analysis herein performed the
Typological analysis of intervention Structural aspects L M H Mark
Technique classification
Stiffening: Yes
Resistance: Tension; Compression; Differential Settlements
Ductility: Yes
Non structural properties
Amount of materials: Steel elements/grout
Technological aspects: Need to perform excavations; drilled dowels must be installed
Used space: Depends on the number of micropiles and construction technique used
Demolition: Yes
Integration in existing building: Difficult application
Accessibility: Average/Difficult
Reversibility: No
Maintenance: Not required
Structural classification
Reinforced Concrete: Introduction of micropiles
Capability to achieve requested performance objective (after building evaluation)
Compatibility with the actual structural system (no need of complementary strengthening or confinement measures)
X
Adaptability to change of action seismic typology (near field, far field, T<>Tc)
X
Adaptability to change of building typology X
Technical Aspects L M H Mark
Reversibility of intervention X
Durability X
Operational X
Functionality and aesthetically compatible and complementary to the existing building
X
Sustainability X
Technical capability X
Technical support X
Available material/device X
Quality control X
Economical Aspects H M L Mark
Costs (Material/Fabrication, Transportation, Erection, Installation, Maintenance, Preparatory works)
X
Typological analysis of intervention Structural aspects L M H Mark
Technique classification
Stiffening: Yes
Resistance: Yes
Ductility: Yes
Non structural properties
Amount of materials: Low
Technological aspects: Non-structural members, such as insulation, fill, roofing and partitions may have to be temporarily removed
Used space: Low
Demolition: No
Integration in existing building: Good
Accessibility: Yes
Reversibility: Yes
Maintenance: Not required
Capability to achieve requested performance objective (after building evaluation)
X
Compatibility with the actual structural system (no need of complementary strengthening or confinement measures)
X
Adaptability to change of action seismic typology (near field, far field, T<>Tc)
X
Adaptability to change of building typology X
Technical Aspects L M H Mark
Reversibility of intervention X
Durability X
Operational X
Functionality and aesthetically compatible and complementary to the existing building
X
Sustainability X
Technical capability X
Technical support X
Available material/device X
Quality control X
Economical Aspects H M L Mark
Costs (Material/Fabrication, Transportation, Erection, Installation, Maintenance, Preparatory works)
X
43
following techniques will be analyzed through the execution of numerical simulations according to
different analysis methods:
Steel shear walls; (vertical elements)
Steel bracing elements; (vertical elements)
Steel frames; (vertical elements)
Steel braced frames; (vertical elements)
Steel strips elements; (vertical elements)
Trussed girder; (Horizontal elements)
Steel tying systems; (Horizontal elements)
Horizontal bracing system; (Horizontal elements)
Micro-piles systems. (Foundation elements)
(a)
(b)
(c)
(d)
(e)
(f) (g)
(h)
Figure 3.2. (a) Installation of Near Surface Mounting GFRP bars; (b) Rectangular FRP grids; (c)
Application examples of CAM; (d) New r.c. slab on existing floor deck; (e) Steel braces for stiffening
of floor systems; (f) In-field execution of ring-beam technique; (g) Typical application of reinforced
concrete jacketing to r.c. columns; (h) Reinforced concrete jacketing of beams.
44
(a)
(b)
(c)
(d)
(e)
(f)
Figure 3.3. (a) Realization of new reinforced
concrete shear wall; (b) Buckling Restrained
Brace; (c) application of steel bracings system;
(d) Dissipative steel eccentric bracing; (e)
insertion of external micro-piles with the
addition of reinforced concrete cap; (f)
Micropile Enhancement to Existing Strip
Footing.
45
T
able
3.4
. M
aso
nry
wal
l ty
po
logie
s an
d m
ain l
imit
atio
ns
of
rehab
ilit
atio
n m
ethod
; Y
es -
Poss
ible
to u
se t
he
met
ho
d f
or
bo
th r
esto
rati
on
an
d s
tren
gth
enin
g;
Int
- O
nly
on t
he
inte
rior
surf
ace
of
the
wal
l; *
- If
th
e w
all
had
pla
ster
ing w
hic
h c
an b
e re
mad
e th
an S
or
“-”
; A
– A
pp
lica
ble
; N
A –
Not
Ap
pli
cab
le;
SC
– S
pec
ial
Car
e; G
–
Good;
IM –
Inte
rmed
iate
; P
– P
oor;
M –
Maj
or;
S –
Sm
all;
-
– N
on
e
Single leaf walls
Cavity walls with rubble filled core
Bonded brick-work
Stone masonry walls
Light-weight CMU units
Concrete block walls
Brick column
Stone column
Joints
Applicability on irregular or rough surfaces
Applicability with weak adjacent material
Visibility for workmanship quality control
Chemical and environmental durability
Fire safety
Aesthetic change
Fer
ro-c
em
en
t Y
es
Yes
Y
es
Yes
Y
es
Yes
Y
es Y
es
- A
A
G
IM
G
S
Sh
otc
ret
e
Yes
Y
es Y
es
Yes
Y
es
Yes
Y
es Y
es
- A
S
C
G
IM
G
S
Rei
nfo
rced
pla
ster
Yes
Y
es Y
es
Yes
Y
es
Yes
Y
es Y
es
- N
A
SC
G
IM
IM
S
Gro
ut
inje
ctio
n
Yes
Y
es Y
es
- Y
es
Yes
Y
es
- -
A
A
P
IM
G
-
Dia
gon
al
stee
l st
rip
s Y
es
Yes
Y
es
- -
Yes
-
- Y
es
SC
N
A
G
IM
P
M*
Rec
tan
gu
lar
mesh
of
stee
l st
rip
s Y
es
Yes
Y
es
- -
Yes
Y
es Y
es
Yes
N
A
SC
G
IM
P
M
*
3D
ste
el t
yin
g
Yes
Y
es Y
es
Yes
Y
es
Yes
Y
es Y
es
Yes
A
S
C
G
P
P
M
RC
tie
colu
mn
s a
nd
bea
ms
Yes
Y
es Y
es
Yes
Y
es
Yes
-
- Y
es
A
SC
G
G
G
S
Cen
tre
core
rei
nfo
rcem
en
t Y
es
- Y
es
Yes
Y
es
Yes
Y
es
- -
A
A
P
G
G
-
Inte
rn
al
po
st-t
en
sio
nin
g
Yes
-
- Y
es
SC
Y
es
Yes
-
- A
A
P
G
IM
-
Ex
tern
al
post
-ten
sion
ing
Y
es
Yes
Y
es
Yes
Y
es
Yes
Y
es Y
es
Yes
A
S
C
G
P
P
M
UD
FR
P i
n X
Y
es
Yes
Y
es
- Y
es
Yes
-
- Y
es
NA
N
A
G
G
P
M*
UD
FR
P r
ecta
ng
ula
r g
rid
s Y
es
Yes
Y
es
- Y
es
Yes
-
- Y
es
NA
S
C
P
G
P
M*
BiD
ir F
RP
la
min
ate
Y
es
Yes
Y
es
- Y
es
Yes
Y
es
- Y
es
NA
A
P
G
P
M
*
NS
M F
RP
Y
es
Yes
Y
es
- Y
es
Yes
Y
es
- Y
es
A
SC
IM
G
IM
S
Toe
con
fin
em
en
t Y
es
Yes
Y
es
- Y
es
Yes
-
- -
SC
S
C
P
G
IM
-
TR
M
Yes
Y
es Y
es
- Y
es
Yes
Y
es
- -
A
A
IM
IM
G
-
Poly
mer
gri
d
Yes
Y
es
- -
Yes
Y
es
Yes
-
Yes
A
A
P
G
IM
-
46
T
able
3.5
Flo
ori
ng s
yst
ems
in m
aso
nry
buil
din
g a
nd m
ain l
imit
atio
ns
of
reh
abil
itat
ion
met
hod
T
able
3.6
Roofi
ng s
yst
ems
in m
aso
nry
bu
ild
ing a
nd
mai
n l
imit
atio
ns
of
reh
abil
itat
ion
met
ho
d
T
able
3.7
. F
ou
nd
atio
n s
yst
ems
in m
aso
nry
buil
din
g a
nd m
ain l
imit
atio
ns
of
rehab
ilit
atio
n
met
hod
Met
ho
d
Co
st
Ap
pli
cati
on
S
tren
gth
u
pg
rad
ing
(Ver
tica
l lo
ad
s)
Sti
ffn
ess
up
gra
din
g
RC
sla
b
Hig
h
Dif
ficult
H
igh
H
igh
This
is
a co
stly
met
ho
d e
spec
iall
y w
hen i
t is
co
mb
ined
wit
h t
he
const
ruct
ion o
f ri
ng
bea
ms
at
each
flo
or
level
. S
pec
iali
zed
wo
rkin
g c
rew
is
nee
ded
and
saf
ety m
easu
res
hav
e to
be
ado
pte
d.
Fo
r th
e m
etho
d c
once
rnin
g t
he
add
ing
of
a s
lim
RC
sla
b o
ver
the
exis
tin
g t
imb
er f
loo
r sy
stem
the
dif
ficult
ies
of
app
lica
tio
n a
re m
uch f
ew
er.
Ther
efo
re t
his
so
luti
on i
s m
ore
fea
sib
le,
but
is
exp
ecte
d t
o h
ave
a p
oo
rer
per
form
ance
.
Ho
rizo
nta
l B
raci
ng
Sy
stem
Lo
w
Eas
y
Lo
w
Hig
h
The
bra
ces
sho
uld
b
e p
rop
erly
an
cho
red
in
th
e w
all
co
rner
s.
This
re
quir
es
dri
llin
g
and
spec
iall
y d
esig
ned
ancho
rage p
late
s. D
esp
ite
the
abo
ve,
it
has
a r
elat
ivel
y l
ow
co
st a
nd
it
is
easy
to
be
app
lied
.
Tim
ber
pla
tes
Ver
y L
ow
E
asy
Lo
w
Lo
w
This
is
the
chea
pes
t an
d e
asie
st i
n t
erm
s o
f ap
pli
cati
on t
echn
ique.
Tie
tec
hn
iqu
e
Mo
der
ate
Mo
der
ate
Lo
w
Hig
h
Ther
e is
incr
ease
d c
ost
in t
he
case
of
usi
ng c
entr
al p
rest
ress
ing b
ecau
se o
f th
e d
rill
ing
and
pre
stre
ssin
g
equip
ment
that
nee
ds
to
be
use
d.
The
app
lica
tio
n
of
the
tech
niq
ue
is
no
t
extr
em
ely
d
iffi
cult
b
ut
spec
iali
zed
w
ork
ing
crew
is
nee
ded
.
If
ther
e is
no
p
rest
ress
ing
invo
lved
, th
e ap
pli
cati
on o
f th
e te
chniq
ue
is m
uch
eas
ier
and
the
cost
is
sig
nif
icantl
y l
ow
er.
Met
ho
d
Co
st
Ap
pli
cati
on
S
tren
gth
up
gra
din
g
(Ver
tica
l lo
ad
s)
Sti
ffn
ess
up
gra
din
g
Rin
g b
eam
Hig
h
Dif
ficult
L
ow
H
igh
A v
ery i
mp
ort
ant
safe
ty r
ule
that
has
to b
e fo
llo
wed
is
the
safe
sup
po
rt o
f th
e ro
of
pri
or
any
inte
rventi
on.
That
is,
the
load
s co
min
g f
rom
the
roo
f sy
stem
sho
uld
be
safe
ly t
ransf
erre
d
dir
ectl
y t
o t
he
gro
und
and
no
t to
the
top
flo
or
syst
em
. T
his
fac
t in
crea
ses
the
cost
of
this
tech
niq
ue.
Nev
erth
eless
it
is n
ot
nec
essa
ry t
o h
ave
a sp
ecia
lize
d w
ork
ing c
rew
.
Ad
din
g o
ut
of
pla
ne
bra
cin
g
Lo
w
Eas
y
Lo
w
Hig
h
The
cost
of
this
inte
rventi
on i
s re
lati
vel
y l
ow
co
mp
ared
to
th
e in
crea
se o
f eff
icie
ncy i
t ca
use
s.
No
sp
ecia
lize
d w
ork
ing p
erso
nnel
are
nec
essa
ry.
Rep
laci
ng
Pa
rts
Lo
w
Eas
y
Mo
der
ate
Lo
w
The
cost
of
this
met
ho
d i
s no
t hig
h a
nd
th
ere
is n
o s
pec
ial
need
fo
r q
ual
ifie
d p
erso
nnel.
If
stee
l
pro
file
s ar
e use
d f
or
the
rep
laced
mem
ber
s th
en t
he
cost
is
slig
htl
y h
igher
.
Met
hod
C
ost
A
pp
lica
tion
S
tren
gth
up
gra
din
g
RC
su
b-
fou
nd
ati
on
Moder
ate
Moder
ate
Hig
h
Bec
ause
of
the
dif
ficu
ltie
s of
this
ap
pli
cati
on t
he
pro
cedure
m
ust
be
separ
ated
in
tw
o
stag
es.
At
each
sta
ge,
the
exca
vat
ions
and t
he
const
ruct
ion o
f th
e su
b-f
oundat
ion s
yst
em
should
be
done
only
in t
he
one
side
of
the
wal
l.
Fou
nd
ati
on
stit
chin
g
Hig
h
Dif
ficu
lt
Hig
h
This
pro
cedure
req
uir
es s
pec
iali
zed w
ork
ing c
rew
and t
he
pro
per
mac
hin
ery,
ther
efore
its
cost
is
much
hig
her
.
47
T
able
3.8
Roofi
ng s
yst
ems
in m
aso
nry
bu
ildin
g:
suit
abil
ity o
f re
hab
ilit
atio
n
met
ho
ds
T
able
3.9
Roofi
ng s
yst
ems
in m
aso
nry
bu
ild
ing:
Imp
rovem
ents
du
e to
reh
abil
itat
ion
met
ho
ds
T
able
3.1
0 F
loo
ring a
nd r
oofi
ng s
yst
ems
in r
.c.
buil
din
gs:
Appli
cabil
ity o
f an
alyze
d t
ech
niq
ues
to
flo
or
typ
es.
W
oo
d S
truct
ure
Ro
of
Ste
el S
tructu
re
Ro
of
Slo
ped
Co
ncr
ete
Ro
of
Co
ncr
ete
Arc
h
Thin
-shel
l R
oo
f
Ply
wo
od
over
lay
Y
es
- -
- -
Bo
und
ary c
on
nec
tio
ns,
dia
phra
gm
cho
rd
Yes
Y
es
Yes
Y
es
Yes
Incr
ease
co
nti
nu
ity w
ith s
teel
elem
ents
Y
es
Yes
Y
es
- Y
es
Ad
dit
ion o
f fa
sten
ers
to m
etal
dec
k
- Y
es
- -
-
Ho
rizo
nta
l B
raci
ng
Y
es
Yes
Y
es
Yes
Y
es
Fib
er-R
ein
forc
ed P
oly
mer
Ov
erla
y
- -
- Y
es
Yes
Rem
oval
of
un
nec
essa
ry s
eism
ic
mas
s Y
es
Yes
Y
es
- -
Ro
of
iso
lati
on
Y
es
Yes
Y
es
Yes
Y
es
S
hea
r/fl
exura
l
Str
ength
S
tiff
nes
s
Co
nnec
tivit
y
to v
erti
cal
bea
ring
elem
ents
Co
nti
nuit
y
Red
uct
ion o
f
Sei
smic
Act
ion
Ply
wo
od
over
lay
Y
es
Yes
-
Yes
-
Bo
und
ary c
on
nec
tio
ns,
dia
phra
gm
cho
rd
Yes
-
Yes
Y
es
-
Incr
ease
co
nti
nu
ity w
ith s
teel
elem
ents
Y
es
Yes
Y
es
Yes
-
Ad
dit
ion o
f fa
sten
ers
to m
etal
dec
k
Yes
Y
es
- -
-
Ho
rizo
nta
l B
raci
ng
Y
es
Yes
Y
es
- -
Fib
er-R
ein
forc
ed P
oly
mer
Ov
erla
y
Yes
-
- -
-
Rem
oval
of
un
nec
essa
ry s
eism
ic
mas
s -
- -
- Y
es
Ro
of
iso
lati
on
-
- -
- Y
es
F
lat
sla
bM
ush
roo
m s
lab
Rib
be
d s
lab
Wit
h b
ea
ms
Ho
llo
w c
ore
Co
mp
osi
te f
l.
Co
ncre
te o
ve
rla
yY
es
Yes
Yes
Yes
Yes
Yes
Sh
otc
rete
Yes
Yes
Yes
Yes
Yes
No
Glu
ed
fin
s (f
loo
rs)
Yes
Lim
ited
Yes
Yes
Yes
No
Po
st-t
en
sio
nin
g (
flo
ors
)Y
es
Lim
ited
Lim
ited
Lim
ited
Yes
Yes
Ste
el
bra
cin
gY
es
Lim
ited
Lim
ited
Lim
ited
Yes
Yes
Pre
ca
st e
lem
en
t jo
ints
No
No
No
No
Yes
No
Co
ncre
te j
acke
tin
gN
oN
oN
oY
es
No
No
Ste
el
jacke
tin
gN
oN
oN
oY
es
No
No
Glu
ed
fin
s (b
ea
ms)
No
No
No
Yes
No
No
Po
st-t
en
sio
nin
g (
be
am
s)N
oN
oN
oY
es
No
No
Ste
el
led
ge
rY
es
No
Yes
Yes
Yes
Yes
Co
ncre
te l
ed
ge
rY
es
No
Yes
Yes
Yes
Yes
Lo
ca
l p
ost
-te
nsi
on
ing
Yes
Yes
Yes
Yes
Yes
Yes
48
T
able
3.1
1 F
loo
ring a
nd r
oofi
ng s
yst
ems
in r
.c.
buil
din
gs:
Non S
truct
ura
l P
rop
erti
es o
f an
alyze
d t
ech
niq
ues
T
able
3.1
2.
Suit
abil
ity f
or
fou
nd
atio
n t
yp
olo
gie
s in
r.c
. an
d m
ain
lim
itat
ions
of
reh
abil
itat
ion m
eth
od
; Y
es -
Poss
ible
to u
se t
he
met
hod
for
stre
ngth
enin
g;
A –
Ap
pli
cable
; N
A –
No
t A
ppli
cable
; S
C –
Spec
ial
Car
e; M
– M
ajo
r; S
– S
mal
l; -
– N
one
T
able
3.1
3.
Suit
abil
ity f
or
fou
nd
atio
n t
yp
olo
gie
s in
r.c
. an
d f
ailu
re m
ech
anis
m i
mp
roved
by t
he
rehab
ilit
atio
n m
eth
od
Ma
inte
na
nc
eR
ev
ers
ibil
ity
Am
ou
nt
of
ma
t.T
ech
no
l.as
pe
cts
Us
ed
sp
ac
eD
em
oli
tio
ns
Inte
gra
tio
nA
cce
ss
ibil
ity
Co
nc
rete
ov
erl
ay
Go
od
No
Mod
era
tem
ode
rate
em
issio
ns
Mo
de
rate
Flo
or
fin
ish
Go
od
De
pen
ds
Sh
otc
rete
Go
od
No
Mod
era
teS
kill
ed w
ork
ers
, h
ea
vy e
mis
s.
Mo
de
rate
Ce
ilin
gG
ood
Go
od
Glu
ed
fin
s (
flo
ors
)M
od
era
teL
imite
dLow
Skill
ed w
ork
ers
, fire
pro
tectio
nLow
Ce
ilin
gG
ood
Go
od
Po
st-
ten
sio
nin
g (
flo
ors
)M
od
era
teY
es
Low
Skill
ed w
ork
ers
Low
Flo
or
fin
., c
eili
ng
Go
od
Go
od
Ste
el
bra
cin
gM
od
era
teY
es
Hig
hS
kill
ed w
ork
ers
, lif
tin
g t
ools
Hig
hC
eili
ng
Mod
era
teG
ood
Pre
ca
st
ele
me
nt
join
tsG
ood
No
Low
De
pe
nd
sN
oF
loor
fin
ish
Go
od
De
pen
ds
Co
nc
rete
ja
ck
eti
ng
Go
od
No
Mod
era
teM
anu
al w
ork
Low
Ce
ilin
g lo
ca
lG
ood
Go
od
Ste
el
jac
ke
tin
gG
ood
No
Mod
era
teM
anu
al w
ork
, fire
pro
tectio
nLow
Ce
ilin
g lo
ca
lG
ood
Go
od
Glu
ed
fin
s (
be
am
s)
Mod
era
teL
imite
dM
od
era
teS
kill
ed w
ork
ers
, fire
pro
tectio
nLow
Ce
ilin
g lo
ca
lG
ood
Go
od
Po
st-
ten
sio
nin
g (
be
am
s)
Mod
era
teN
oM
od
era
teS
kill
ed w
ork
ers
, fire
pro
tectio
nLow
No
Go
od
Lim
ite
d
Ste
el
led
ge
rM
od
era
teY
es
Low
Fire
pro
tectio
nLow
Low
Go
od
Go
od
Co
nc
rete
le
dg
er
Go
od
No
Low
Mo
de
rate
ma
nu
al
Low
Low
Go
od
Go
od
Lo
cal
po
st-
ten
sio
nin
gM
od
era
teN
oLow
Skill
ed w
ork
ers
, fire
pro
tectio
nLow
Low
Go
od
Go
od
Isolated spread
footings
Strip Footings
Foundation of new
elements
Mat foundations
Pile foundations
Accessibility and
height restrictions
Impose noise and
vibration
Restrictions
imposed by existing
utilities (gas, water
supply systems)
Restrictions
associated with on
going operations
Sp
read
fo
oti
ng e
nla
rgem
ent
or
rep
lace
ment
Yes
Y
es
Yes
-
- A
S
S
S
Ad
dit
ion o
f a
stra
p b
eam
Y
es
Yes
-
- -
A
S
S
S
Ad
dit
ion o
f m
icro
pil
es
Yes
Y
es
Yes
Y
es
Yes
N
A
M
M
M
Ad
dit
ion o
f sh
allo
w e
lem
ents
to
dee
p
fou
nd
atio
ns
- -
- -
Yes
A
S
S
S
Ad
dit
ion o
f d
riven
Pil
es
Yes
Y
es
Yes
Y
es
Yes
N
A
M
M
M
Over
layin
g m
at f
ound
atio
ns
- -
Y
es
- A
S
S
S
Compression
Tension
Ovetrturning
Differential
Settlement
Fault Rupture
Liquefaction
Differential
Compaction
Landsliding
Sp
read
fo
oti
ng e
nla
rgem
ent
or
rep
lace
ment
Yes
Y
es
Yes
Y
es
- -
Yes
-
Ad
dit
ion o
f a
stra
p b
eam
Y
es
Yes
Y
es
Yes
-
Yes
Y
es
-
Ad
dit
ion o
f m
icro
pil
es
Yes
Y
es
Yes
Y
es
- Y
es
Yes
-
Ad
dit
ion o
f sh
allo
w e
lem
ents
to
dee
p f
ound
atio
ns
Yes
Y
es
- -
- -
- -
Ad
dit
ion o
f d
riven
Pil
es
Yes
Y
es
Yes
Y
es
- Y
es
Yes
-
Over
layin
g m
at f
ound
atio
ns
Yes
Y
es
- -
- -
- -
49
T
able
3.1
4 S
uit
abil
ity f
or
foundat
ion t
ypolo
gie
s in
r.c
. an
d f
ailu
re m
echan
ism
im
pro
ved
by t
he
reh
abil
itat
ion
met
ho
d
Mai
nte
nan
ceR
eve
rsib
ilit
yA
mo
un
t o
f
mat
eri
alU
sed
sp
ace
De
mo
liti
on
sIn
tegr
atio
nA
cce
ssib
ilit
yM
aso
nry
R.C
.
Re
info
rce
d c
on
cre
te
po
st-c
ast
she
ar w
alls
Mo
de
rate
No
Hig
hH
igh
infi
ll w
alls
Mo
de
rate
De
pe
nd
sYe
sYe
s
Ste
el c
on
cen
tric
bra
cin
gsG
oo
dYe
sM
od
era
teM
od
era
tein
fill
wal
lsG
oo
dG
oo
dYe
sYe
s
Ste
el e
cce
ntr
ic
bra
cin
gsG
oo
dYe
sM
od
era
teLo
win
fill
wal
lsG
oo
dG
oo
dYe
sYe
s
Dis
sip
ativ
e b
raci
ngs
Go
od
Yes
Low
Low
infi
ll w
alls
Go
od
Go
od
Yes
Yes
Me
tall
ic S
he
ar P
ane
lsG
oo
dM
od
era
teM
od
era
teM
od
era
tein
fill
wal
lsG
oo
dD
ep
en
ds
Yes
Yes
50
4. Benchmark buildings and calibration of numerical tools
4.1. Description of reinforced concrete benchmark building The old design code assumed in the design process is the Royal Decree n.2229 November 16
th, 1939
issued in Italy for the construction of reinforced and not reinforced concrete building. It was decided to
adopt this old design standard because many reinforced concrete buildings were designed according to
its rules in the ’50 to early ’70 of the XX century in Italy.
4.1.1. Materials and general geometry According to this old regulation, the following material properties were assumed in the design:
“High strength” concrete
Allowable compressive strength 4.5 MPa
Allowable compressive and bending strength 5.0 MPa
Allowable shear strength 6.0 MPa
This concrete can be considered, following the actual classification, as equivalent to a concrete defined
by a characteristic compressive strength of 20 MPa (Rck = 20 MPa).
Mild steel
Allowable tensile strength 140 MPa
Homogeneization coefficient (for “high strenght” concrete) m = 8
This steel class can be assumed in the structural assessment defined by a characteristic yielding strength
equal to fyk = 230 MPa.
Geotechnical parameters
In the design stage it was assumed a foundation soil characterized by an allowable bearing capacity of
0.11 MPa and a modulus of subgrade reaction of 0.01 N/mm3.
Geometrical dimension
The benchmark structure is a three dimensional reinforced concrete frame formed by three storeys, five
to four bays, see figures 4.1, 4.2 and 4.3 The geometrical dimensions of the building are about 23 x 18
m in plant while it has an height of about 10 m at the eaves and about 12 m at the ridge.
(a)
(b)
Figure 4.1. Reinforced concrete benchmark building: (a) first floor plan, (b) second floor plan.
51
(a)
(b)
Figure 4.2. Reinforced concrete benchmark building: (a) third floor plan view, (b) foundations.
Figure 4.3. Typical main frame of the structural scheme in the reinforced concrete benchmark
Figure 4.4. Typical secondary frame of the structural scheme in the reinforced concrete benchmark
52
4.2. Definition of the masonry benchmark building The masonry building has been designed according only to geometrical considerations (as typical at the
beginning of the XX century); this building should be assumed as reference benchmark structure during
the execution of the performance analyses of the steel intervention techniques for the retrofitting of
vertical elements, floors, roofs and foundations.
4.2.1. Materials and general geoemetry The material properties adopted for the structural modelling of the masonry benchmark are drawn by
literature [O.P.C.M. 3431/2005 – “Technical Italian Standards for Design, Seismic Assessment and
Retrofitting of Buildings”]
Walls (stone masonry)
Mean compressive strength fm 1.5 MPa
Mean shear strength 0 5.6 10-2
MPa
Mean elastic modulus Em 1500 MPa
Mean shear modulus Gm 250 MPa
Mean unit weight w 21 kN/mm3
Walls (hollow brick masonry)
Mean unit weight w 11 kN/mm3
Vaults, arches and floors (brick masonry)
Mean compressive strength fm 1.8 MPa
Mean shear strength 0 6 10-2
MPa
Mean elastic modulus Em 1800 MPa
Mean shear modulus Gm 300 MPa
Mean unit weight w 18 kN/mm3
Geotechnical parameters
At the design stage it was assumed a foundation soil characterized by an allowable bearing capacity of
0.11 MPa and a modulus of subgrade reaction of 0.01 N/mm3.
(a)
(b)
Figure 4.5. Masonry benchmark building – plan views: (a) first floor; (b) second floor.
53
(a)
(b)
Figure 4.6. Section views: (a) C-C section; (b) B-B section.
(a)
(b)
Figure 4.7. (a) A-A section view of the building; (b) particular of floor systems at the last floor under
the roofing system.
4.3. Calibration of numerical models A detailed investigation about the retrofitting performance of several steel systems applied on the same
reinforced concrete and masonry structure is a relevant part of the research. According to this,
numerical models defined by partners have been compared in the assessment of seismic vulnerabilities
of structural benchmarks. In particular, the model of reinforced concrete benchmark building has been
developed using four different softwares: OPENSEES, SEISMOSTRUCT, SAP2000 and DYNACS;
their respective results, concerning with the structural assessment of original building, have been
compared. On the contrary, masonry benchmark building is modelled, using software ABAQUS; the
model has been developed by one partner and diffused to all the other involved in the structural
analyses on masonry building. In such a way, the comparability between the numerical results coming
from various simulations is maintained. Both the reinforced concrete and masonry buildings have been
checked against the ultimate limit state load combinations for the static actions (live loads, wind load
and snow load) and against the exceptional load combinations for the earthquake action. Main attention
is focused on the results correlated with the seismic action.
4.3.1. Reinforced concrete building
4.3.1.1. Non-linear modelling issues adopting SEISMOSTRUCT Steel is modeled as bilinear material with the parameters presented in the previous paragraph where the
main information about r.c. benchmark are given . The yield strength was fy=230MPa
54
(ES=200000MPa), and a small value of the strain hardening (1%) was accepted. The possibility of
reinforcement fracture was not included in the model, but the strain level of δs_F=0.04 was monitored as
ultimate strain for the reinforcing steel.
Concrete materials are modelled by a tri-linear material model. The characteristic value of the
compressive strength was taken Rck=20 MPa (EC=29000MPa). The confined concrete (i.e. inside of the
reinforcement cage) retains more significant compression strength, after crushing, then the unconfined
concrete outside the reinforcement cage. For the confined concrete (i.e. inside the reinforcing cage) the
nonlinear concrete model was used, while for the unconfined concrete the tri-linear model.
The remaining compressive strength was set to Rconf=6MPa & Run-conf=2MPa; strain at peak stress has
been fixed to 0.002 for both models (figure 4.8). For the crushing strain of concrete the values of 0.006
(confined) and 0.002 (unconfined) are recommended. However, as the structure is known to be made of
very poor quality concrete, these values have been reduced. More relevant values can be determined
experimentally.
Figure 4.8. (a) Confined (i.e. inside the reinforcing cage) and (b) un-confined (i.e. outside of reinforcing
cage) concrete material properties
4.3.1.1.1. Modelling of cross-sections Cross-sections of the model are divided into 200 fibers. The fibers have one of the properties of
concrete or steel base material. The division is done automatically by SEISMISOFT based on the
geometry of the cross-section and the place of the reinforcing bars. Each fiber is behaving as “confined
concrete”, “unconfined concrete” or “steel”.
4.3.1.1.2. Performance criteria Performance criteria is monitored in the response via the material strains: spalling of the concrete cover
is considered at εsp=-0.002. crushing of the concrete core is εcrush=-0.0035; yielding of the steel
reinforcement fixed to εs_N =fy/Es=0.00115; fracture of the steel reinforcement is εs_F=0.04.
4.3.1.2. Non-linear modelling issues adopting SAP2000 Concrete material was modelled as nonlinear based on Kent and Park model (see figure 4.9) with no
tensile resistance. The concrete was considered as unconfined and the concrete young modulus is set
equal to 29000 MPa. Reinforcement was modelled as modified Park nonlinear using a yield strain of
0.0015 and an ultimate strain from 0.2 to 0.3 corresponding to yield strength of 230 MPa and an
ultimate strength of 350 MPa.
4.3.1.2.1. RC elements (beams and columns) Reinforced concrete elements were modelled as plastic hinges concentrated at the ends of the elements.
With the specification that in case of beams plastic hinges were concentrated in all points were the
rebars change their number from the upper part to the lower part of the cross section and reverse. Plastic
hinges were define as load – deformation relationship following FEMA356 model as a deformation
controlled (ductile) typology.
In the case of beams a moment – rotation relationship for unconfined concrete was described following
acceptance criteria values from FEMA356 tables, basing on efforts obtained from gravity loads (see
figure 4.10). In the same way were defined all plastic hinges for the columns, only that the moment –
rotation relation was defined differently for each direction of column cross section.
55
Figure 4.9. Reinforced concrete material nonlinear model based on Kent and Park; (b). modified Park
nonlinear model of steel reinforcement
Figure 4.10. Deformation controlled action model with nonlinear load-deformation parameters and
acceptance criteria (FEMA356)
4.3.1.2.2. Modeling hypothesis Following FEMA356 table, the stiffness of beams and columns should be reduced by 50%, due to the
fact that beams are nonprestressed and columns have low axial compression, due to design gravity load
lower than 0.3Agfc’. The floor/roofing system defined by thick parallel ribs, the floor/roof was
considered to be as a rigid diaphragm.
4.3.1.3. Non-linear modelling issues adopting DYNACS In the push-over analysis the same model is used than that adopted during the elastic analysis executed
to check the model and to determine the natural periods and participating masses, however with
nonlinear material behaviour. The non-linear moment-rotation springs at the end of the members is pre-
determined under consideration of axial forces.
The effective stiffness of the sections between the plastic hinges is determined in accordance with the
indication furnished by FEMA356. As the axial load in the columns is low, the effective stiffness of all
sections is considered reduced by a factor of 0.5. The assumed properties are: fck,cylinder = 16 N/mm², Ecm
= 29000 N/mm², εcu = 0.35 %, fsm = 230 N/mm² and Es 200000 N/mm². The non-linear moment-
rotation springs of columns and beams are determined by integration of the moment-curvature curves
over an estimated length of the plastic hinge. The moment-curvature curves are determined by a
nonlinear cross-section analysis under the axial load at the maximum lateral load. The plastic hinge
length is obtained in accordance to Paulay and Priestley (1991):
yiiPL fd022.0L08.0L (4.2)
In the figure 4.12 the moment-rotation curve determined by the nonlinear section analysis is compared
with the curves proposed by FEMA 356. The curve in according to FEMA 356 for structural members
with conforming transverse reinforcement fit well with the calculated one, while the curve for beams
with insufficient transverse reinforcement leads to significant smaller rotation capacities. As the
sections in the benchmark building have an insufficient transverse reinforcement, the rotation capacity
may be overestimated by the section analysis.
56
Figure 4.11. Effective stiffness of RC-elements according to the FEMA356
Figure 4.12 Moment-rotation curve for section 1 by section analysis and FEMA 356 with
nonconforming (NC) and conforming (C) transverse reinforcement
4.3.1.4. Modelling issues using OPENSEES The finite element model of r.c. benchmark was developed with OPENSees programme (OPEN System
for Earthquake Engineering Simulation - Pacific Earthquake Engineering Research Center, University
of California, Berkeley) using a fiber based modelling strategy for the cross section of each structural
element (see figure 4.14).
4.3.1.4.1. Nominal material properties The material properties adopted into the model are respectively the Giuffré-Menegotto-Pinto model for
the reinforcing steel and the Popovics model for concrete both implemented in the OPENSees library.
figure 4.13 reported the stress-strain diagrams calculated adopting the main mechanical properties and
material model parameters used for the simulations executed using Dynacs software.
(a)
(b)
Figure 4.13 Stress-strain models adopted in OPENSEES: (a) reinforcing steel; (b) concrete (slightly
confined)
57
Steel is modeled according to the Giuffré-Menegotto-Pinto model (Menegotto and Pinto, 1973)
characterized by a bilinear skeleton with a smooting part that control the transition between the elastic
to plastic branches: post yield tangent modulus was fixed equal to 0.0033 times the elastic modulus.
Concrete material was modeled by the Popovics model (Popovics, 1973).
4.3.1.4.2. Modelling of cross-sections Each structural element of the benchmark frame was modelled using a fiber based approach subdiving
the cross section into longitudinal fibers having the maximum dimension of 20 mm for the concrete
cover and 40 mm for the concrete core (see figure 4.15). Beams and columns are discretized in order to
take into account the changes of reinforcing bars along each element.
Figure 4.14 Cross section fiber subdivision: (a) subdivision in different zones; (b) definition of the
concrete fibres; (c) position of steel reinforcement.
4.3.1.4.3. Modelling of floor system The floor structure was modelled by an equivalent truss system in order to take into account the
diaphragm effect of the r.c. slab; this approach has been used also in SeismoStruct software. The
stiffness of the equivalent truss system (see figure 4.15) was evaluated by the following relationship:
truss 3
conc slab conc slab
1KL L
12E J G A (4.3)
steel trusstruss
truss
E AK
L (4.4)
Figure 4.15 Equivalent truss system for floor modelling.
4.3.2. Masonry building 3D Finite Element Modeling (FEM) has been adopted for the analysis of masonry benchmark building;
in particular, as executed for the reinforced concrete building, a preliminary elastic analysis has been
carried in order to assess main vulnerabilities (to be confirmed by non-linear models) and in a second
step a complete non-linear model has been developed using ABAQUS software. Differently from
reinforced concrete benchmark, all partners has adopted the same model whose input file has been
created by one partner for all. According to this assumption, no benchmarking was executed on the
masonry building.
The refined non-linear model has been suitably calibrated on the basis of the main results coming from
a previous research project carried out at European Level. In particular the material properties of the
model have been chosen according to the simulations executed on a single masonry shear wall.
58
Figure 4.16. Calibration of the constitutive model for masonry in the ABAQUS software
4.3.2.1. Material properties The material properties adopted for the structural modeling of the masonry benchmark are drawn by
literature [O.P.C.M. 3431/2005 – “Technical Italian Standards for Design, Seismic Assessment and
Retrofitting of Buildings”]. The mean tensile strength however was not reported. Therefore it was
assumed to be 10% of the mean compressive strength.
Mean
compressive
strength - fm
Mean
tensile
strength - ft
Mean shear
strength – 0
Mean
elastic
modulus -
Em
Mean shear
modulus -
Gm
Mean unit
weight - w
MPa MPa MPa MPa MPa kN/m3
Walls
(Stone
Masonry)
1.5 0.15 5.6×10-2
1500 250 21
Walls
(Brick) - - - - - 11
Vaults,
arches, floor 1.8 0.18 6×10
-2 1800 300 18
Table 4.1. Mechanical properties of masonry materials in benchmark building
The material model adopted for the masonry building was in-built concrete damage plasticity model of
ABAQUS. The disadvantage of this model is that it cannot handle orthotropic behavior, and therefore is
not very well suited for modeling masonry, which has different properties parallel and perpendicular to
the bed joint. Anyway, the main idea has been to find an equivalent material to replicate the behaviour
of the retrofitted and unretrofitted model arisen. This simplification must be carefully analyzed and
argued.
The advantages of such a model is the possibility to applies the nonlinear analysis and to characterize
the global behaviour of the building in term of drift ratios, which gives the possibility to use the FEMA
356 criteria for validation and performance levels’ characterization.
(a)
(b)
Figure 4.17 FEM model of the benchmark building realized using ABAQUS software.
59
4.3.3. Comparison of the results and identification of vulnerabilities in r.c.
benchmark building Software capabilities have been tested and compared using static pushover analysis applied on
reinforced concrete benchmark building; in particular, this comparison have also furnished information
about the main seismic vulnerabilities of the reinforced concrete benchmark. Performances furnished by
the structure have been presented in terms of capacity spectrum in the ADRS plane. Two pushover
analyses have been executed, in X and Y direction of the benchmark plan, using a 3D model where
accidental torsional effects and member imperfections have been considered also, see figure 4.18.a.
During each push over, as depicted in figure 4.19, the occurrence of local failure modes has been
recorded in order to identify collapse condition at which stop numerical simulations. On the basis of
these results, maximum displacement, required ductility and available ductility have been determined
adopting N2 method, see figure 4.20, for transferring capacity curves on the ADRS plane. First
evidence is related to the fact that different programs have given comparable results in terms of
maximum displacement and maximum force of the different models, see table 4.2; moreover,
information about available and required ductility have been reported also. Results, as expected, are not
coincident and there is some scatter between the different models due also to different personal
approaches followed by each partner. Position of fixing conditions at the bottom of the structure (end of
column or foundation centroid), position of beams at each storey level (beam centroid or floor level)
and other little difference produced the scattering reported in table 4.2, that has been judged not too
high considering that the comparison has been made between non-linear simulations. This result has
been accepted and all four software have been employed by partners for the execution of non-linear
simulations and the application of PB methods in order to tests different intervention techniques.
(a)
(b)
(c)
Figure 4.18 (a) 3D model . deformed shape; deformation in the last captured step: (b) X, (c) Y direction
Figure 4.20. Static pushover curves of the 3D frame in the X and Y direction with identification of
several failure modes
0
500
1000
1500
0.00 0.05 0.10 0.15 0.20 0.25
dn(m)
Fb(k
N)
X
Y
o column
∆ beam
yielding of reinforcement
spalling of concrete cover
crushing of concrete core
60
Figure 4.21. Application of ADRS method for seismic performance assessment in X, Y direction
Problems with the initial reinforced concrete structure individuated after non-linear analyses can be
summarized in the following points:
torsion sensitiveness (TTors ~ TXtrans);
weakness in both X and Y direction;
extreme flexibility in both directions;
high level of compression force (columns tend to fail by crushing of the concrete);
in the X direction the structure is weak-column/strong-beam structure, (the opposite to the one
suggested by design codes).
One of the most disturbing of these problems is the fact that axial forces in columns are very high
compared to the capacity of the columns. This leads to sudden (crushing) failure of the concrete in the
columns, at very low values of the lateral displacement. Even if parallel load bearing systems are
activated below these displacement values, the columns of the frame are still under high compression
and they will fail suddenly at these displacement values. All problems have been reported in the table
4.3 and they have been mainly recognized in all numerical analyses carried out using different software.
Fmax [kN] Dmax [m] required available
x y x y x y x Y
SeismoStruct 750 1093 0.070 0.120 2.5 4.5 1.6 1.9
Dynacs 730 - 0.077 0.070 3.4 3.9 1.3 1.8
OpenSees 800 1197 0.066 0.098 3.7 3.6 1.2 1.4
Sap2000 820 1210 0.065 0.100 2.5 3.2 - -
Table 4.2. Maximum displacement, required and available ductility determined from different software
Nr. Vulnerability Cause
1 Weak in X direction
2 Weak in Y direction Beams are weak
3 Flexible in X direction T*x = 1.25 s Columns are weak
4 Flexible in Y direction T*y = 1.51 s are quite large Beams are weak
5 Not ductile enough in X direction Because failure is local in the 1st floor
6 Not ductile enough in Y direction
7 X direction sudden crushing/failure Existing level of compression force on some
columns
8 In the X direction the structure is weak-
column/strong-beam
9 Torsion sensitive (TTors~TXtrans)
Table 4.3. Recognition of main structural vulnerabilities in the r.c. benchmark
0
2
4
6
8
0 0.05 0.1 0.15 0.2
Sed(m)
Se(m
/s2)
μreq=4.43
T*=1.25
0
2
4
6
8
0 0.05 0.1 0.15 0.2
Sed(m)
Se(m
/s2)
μreq=2.51
T*=1.51
61
4.3.4. Initial assessment of the masonry building
4.3.4.1. Vertical loads Under the vertical loads corresponding to the earthquake load combination, the masonry walls will
develop a stress pattern, which will affect the performance to horizontal pushover. As shown in figure
4.22, there is a strong interaction between the normal and shear stresses in the masonry material. The
stress state under vertical loads is presented in figure 4.22.a for the rigid floor and in figure 4.22.b for
the floor free model.
As it can be seen in figure 4.22 the largest compression stress, corresponding to σy is -0.47N/mm2 and -
0.57 N/mm2 respectively. This means 30% from 1.5 N/mm
2, corresponding to uniaxial compressive
failure stress of the masonry.
(a) (b)
Figure 4.22. (Y-Y) direction stresses in the masonry from vertical loads.
The load vs. vertical displacement curves are also presented for the two models in figure 4.23. It can be
observed that the model using “rigid floor” assumption is slightly stiffer, but no significant difference
has been observed. This is probably caused by the fact that 74% of the vertical load is the mass of the
walls, so the distribution of the remaining 26% load is not crucial.
Figure 4.23. Vertical load vs. vertical displacement.
4.3.4.2. Horizontal loads Given the way floors are constructed, the original building can be considered as floor free, because the
existing floor arrangements ensure very limited diaphragm action. If the 3D, floor free model is
analyzed steadily increasing horizontal forces, the deformed shape presented in figure 4.24 is obtained.
It can be observed that, under this pushover type loading, the failure mode of the structure is always
based on a local mechanism. One main failure mode is due to the separation of the heavy external walls
from the transversal ones (e.g. Point 2 in figure 4.24.a). The second failure mode is by out of plane
deformation of wall segments perpendicular to the loading direction (e.g. Point 3 in figure 4.24.a). As it
can be observed in figure 4.25, both phenomena happens at a very small value of the base shear, below
and around Fb = 800 kN. Keeping in mind that the order of magnitude of the base shear is expected to
0
4000
8000
12000
16000
-0.0015 -0.001 -0.0005 0
Dv_average(m)
Fv(k
N)
3D-Tie- Vertical & Mass only
3D-Free- Vertical & Mass only
62
be in the range of thousands (e.g. 8900 kN is the order of magnitude discussed in) it appear that one of
the goals of the rehabilitation will have to be the tying together of the walls, in order to avoid localized
failure modes.
Possibly, it may be necessary to establish diaphragm effect at each floor level within the structure, in
order to ensure more uniform distribution of the stresses and cracking under the horizontal loads.
Figure 4.24. Plastic-strain/cracking pattern at failure for (a) X direction and (b) Z direction pushover.
a) b)
Figure 4.25. Deformations in the points of figure 4.24. vs. the base shear in (a) X direction and (b) Z
direction loading.
4.3.4.3. Deficiencies of the existing building As far as the current configuration is concerned the following structural properties and potential
deficiencies have been identified:
The structure is almost symmetrical and has similar behavior in the two main directions. Torsion
does not affect the performance.
The largest part of the seismic mass is given by the weight of the wall elements. Both the weight of
the floors and the mass coming from live load is less significant.
In the current configuration the biggest weakness of the structure is the lack of diaphragm effect at
both the level of the floors and at the level of the roof. As consequence the walls are not tied
together and local failure is governing the behavior. Realizing an effective tying between the walls
has to be the main priority of any rehabilitation.
1
2
3
1
2
0
200
400
600
800
1000
0.00 0.01 0.02 0.03 0.04
dn(m)
Fb(k
N)
dx1(m)
dx2(m)
dx3(m)0
200
400
600
800
1000
0.00 0.01 0.02 0.03 0.04
dn(m)
Fb(k
N)
dz1(m)
dz2(m)
63
5. Performance analysis of steel solutions for vertical elements A number of existing buildings have good strength and stiffness levels, but some of their components
may not have adequate strength, toughness, or deformation capacity to satisfy the Performance
Requirements. An appropriate strategy for such structures may be to perform local modifications of
components that are inadequate while retaining the basic configuration of the building’s lateral-force-
resisting system. Local modifications that can be considered include improvement of component
connectivity, component strength, and/or component deformation capacity. But this strategy tends to be
the most economical rehabilitation approach when only a few of the building’s components are
inadequate. Global stiffening and/or strengthening of the structure may be effective retrofit strategy if
the results of a seismic evaluation show deficiencies attributable to a global behaviour in structural
strength and/or to excessive lateral deflection of the building, and critical components do not have
adequate ductility to resist the resulting deformations.
Construction of new braced frames, bracing systems and shear walls within an existing structure are
effective measures for adding stiffness and strength at the same time. By using added structural
components, the threshold of ground motion can raise a level at which the onset of damage occurs.
Shear walls and braced frames are effective elements for increases in strength, but they may be
significantly stiffer than the structure to which they are added, which requires their design to provide
nearly all of the structure’s lateral resistance.
The problem of including stiffening/strengthening systems in vertical structures have been considered
from a theoretical point of view, by defining an optimization algorithm for the added elements in a
building, and then considered by a practical point of view, by considering some retrofit solutions as
case studies. In particular, the first part of this chapter presents the application of an optimization
algorithm to some reinforced concrete frames that have to be retrofitted using bracing elements. The
choice of the frame type and the retrofitting system has been made only for sake of simplicity and
representativeness; the guidelines indicated can be extended to other structural types and other
intervention techniques, given that the procedure works in general terms looking at strength and
stiffness (§5.1).
In the second part of this chapter, the application of steel based intervention technique to the benchmark
base cases has been considered. In particular, the retrofitting technique are presented and assessed in
(§5.2); the insertion of the elements into the structural scheme has been following the general guidelines
derived from the application of the optimization procedure (§5.1). At last, the comparison between the
performance of the applied techniques and their costs (estimated according to a simplified model) have
been reported (§5.3); through this combined analysis between structural performance and economic
costs, general guidelines for designers and suggestions have been derived in order to structure the
practical approach to the problem.
5.1. Insertion of new elements in existing vertical systems The design process of seismic retrofitting intervention usually started with the placement of new
elements inside the upper structural system. Obviously, it is impossible for the designers trying all
possible locations and also when some architectural constraints are present, the possibilities are still
relevant and all of them cannot be tested in order to define an appropriate starting point for the
application of PBA procedure and so the execution of the numerical analyses.
For such reason, all the numerical simulations carried out on the complete structural benchmark have
been carried out considering the application of the analyzed intervention techniques according to a
general preliminary guidance, obtained applying an optimization procedure on some base cases
characterized by many different morphologies. From this analyses six general indications have been
obtained and have been used as guide for the preliminary fixing of some design parameters: in
STEELRETRO method the placement of the elements.
From a practice point of view, the proposed optimization procedure can be divided into three phases:
analysis (§5.1.1), evaluation (§5.1.2) and solution (§5.1.3). In the section §5.1.4 the guideline are
reported.
5.1.1. Analysis phase For the insertion of new elements in existing structural systems, the system without retrofitting elements
is considered as a starting point. Accordingly, the rehabilitated system can be treated as an upgrade of
the initial system and the retrofitting elements are considered as structural components that have to be
65
added to the initial system, whose characteristics are assumed to be not influenced by the added
elements. This statement allows focusing the attention on the additional retrofitting elements. Following
this statement, an optimization procedure has been defined in order to determine the “optimal”
characteristics of retrofitting elements according to the characteristics of the system to be retrofitted.
In the analysis phase of the optimal design methodology, the design variables and performance
parameters of interests are specified. These design variables and performance parameters are used to
express the level of satisfaction of the design criteria in a quantitative manner so that an overall design
performance measure can be computed for each design. In details, the “design variables”, designated by
a vector X, are those parameters of the design which are selected to be varied during the search for an
optimal design. For example, design variables may take the form of geometric information for the
structural members, such as cross-sectional dimensions. On the other hand, performance parameters,
designated by a vector q, represent quantities related to the “performance requirements”, and can take
the form of conventional structural parameters (e.g. Stress, deflection, inter-story drift) or other
parameter (e.g. structural reliability). Obviously, the performance parameters, q(X), are functions of the
current design parameters, X.
Structural performance parameters under “deterministic” (code-based) loads can be computed using a
finite-element model of the structure which is specified by the design parameters. In this case, a
particular set of values X, (reference design values) can be used and the corresponding set of
performance parameter q can be evaluated. Eventually, quantities m directly related to performance
parameters can also be evaluated, so that m(q(X))=m(X).
5.1.2. Evaluation phase The objective of the evaluation phase of the optimal design methodology is to obtain an overall
evaluation measure m(X) for the design specified by the current value of the design variables vector X.
This measure m(X) serves as an objective function which, at the revision stage, is used to determine
improved, or optimal, design.
At the same time, “performance limits” b have to be associated with performance parameters q,
identified in the previous phase. In general, the designer may wish to impose many different
performance requirements. Therefore, since not every performance requirements can be satisfied to its
maximum extent simultaneously with the other requirements, the methodology must allow a trade-off to
occur between conflicting criteria in the optimization process.
Performance requirements are treated as any constraint imposed on the design variables, such as
geometrical constraints. The respect of the requirements can be imposed in deterministic and semi-
probabilistic terms by simple inequality equations: “performance parameters” must not exceed
“performance limits” q(X)<b (a “failure condition” Y=q(X)-b is defined). However, this approach can
result too restrictive, so the requirements can be considered in probabilistic terms: the probability that
“performance requirements” exceed “performance limits” must not exceed an allowable value (the
probability of failure must not exceed the “allowable probability of failure” Pf). This approach can be
easily performed by assuming a probability distribution for the quantity Y=(q(X)-b) and imposing that
P(Y>0)<=Pf.
5.1.3. Solution phase The solution phase consists in the stage in which the quantities and the relations defined in the previous
phases are expressed in a form that can be used in the optimization method, using the mathematical
programming. The choice of the solution technique depends essentially on the ratio between the number
of the variables and the number of the equations, and on the complexity of the problem (linearity or non
linearity of the equations).
A mathematical optimization problem (irrespective of the method of solution) is generally stated as
follows:
Minimize (or Maximize): f (X) (5.1)
Subject To: hi(X)=0 (5.2)
gj(X)≤0 (5.3)
The structural configuration of the system is assumed to be known. From a mechanical point of view,
that is equivalent to consider mass, stiffness and damping matrices ms, ks and cs as known. This
assumption is generally valid when the retrofitting system must be inserted in a structure whose
66
configuration is substantially fixed on the base of aesthetic, economic and functional reasons. Equations
of motion of the same system in which the retrofitting elements are inserted are obtained from the
equations of the motion of the initial system by means of simple adding of the relative terms due to the
retrofitting elements. The two resisting systems develop “in parallel” the inner forces that guarantee the
equilibrium. Accordingly, proposed algorithm determines the optimal global stiffness matrix K (i.e. of
the system structure + resisting-elements), under fixed boundary conditions. With this aim, K is
expressed as a linear combination of the structural stiffness matrix and the stiffness matrices of n
resisting elements with fixed dimensions:
i
n
i rr
T
rt
rttt
Ki
rr
T
rt
rttt
1 ΔKΔK
ΔKΔK
KK
KKαK (5.4)
where K is the stiffness matrix of the system, ΔK is the stiffness matrix of the resisting element,Ki is
the design variable for the i-th resisting element, and the subscripts t and r indicate the degrees of
freedom with mass and without mass respectively.
The dissipative non linear behavior of the resisting element is modeled by means of equivalent linear
damped behavior obtained by linearization method proposed by Kryloff and Bogoliubov:
(5.5)
Where x is the generalized displacement, is the damping coefficient per unit mass, 2 is the linear
stiffness per unit mass, is a dimensional parameter, g is a non-linear function, e is the error term,
subscript eq means equivalent, a is the amplitude of the sinusoid that better approximates the motion
and C is the power dissipated during the motion.
The developed procedure has been validated on several case studies in which elastic and dissipative
braces are inserted: a portal frame, a 3bay×3floors frame and a 3bay×3bay×3floors frame. For each case
study, 5 damping levels have been considered, ranging from 5% to 30% of damping ratio. The
procedure appeared feasible for implementation on real structures.
5.1.4. Optimal sizing and placement Design of retrofit systems requires that the sizes and the placement of stiffening/strengthening elements
are determined. The optimization procedure has been applied to several case studies, which include a
regular building and several buildings with different irregularities, with the aim of giving simplified
general criteria for the choice of the braces placement when irregularities are present. The irregular
buildings are obtained from the regular one only by changing its plant and profile, while material and
structural elements in frame structures have the same characteristics (see Table 5.1). A steel bracing
system has been designed for each building.
Mechanical Characteristics of the Elementary Frame
Column Section 40x40 cm
Beam Section 30x60 cm
Span Length 500 cm
Column heigth 300 cm
Concrete Elastic Modulus E 25000 Mpa
0.2 -
Shear Type Behavior Stiffness K 53.8 kN/mm
Table 5.1. Mechanical characteristics of the elementary frame.
The guiding principles governing the conceptual design of the case studies are here synthetically
described:
the “regular building” is characterized by structural simplicity, uniformity, symmetry and
redundancy; furthermore a bi-directional resistance and stiffness and a torsional resistance and stiffness
are guaranteed, as well as a diaphragmatic behaviour at storey level;
the “dumpbell shaped building” has in plan set-backs (re-entrant corners) exist, with the area
between the outline of the floor and a convex polygonal line enveloping the floor that is 33%>5 % of
67
the floor area;
the “L-shaped building” has an in-plan stiffness of the floors not sufficiently large in
comparison with the lateral stiffness of the frames, so that the deformation of the floor have a large
effect on the distribution of the forces among the frames;
in the “asymmetric re-entrant profile building”, there is a single setback of 50 % of the previous
plan dimension, exceeding 15 % of the total height of the main structural system;
the “symmetric re-entrant profile building” has a setback preserving axial symmetry exist, but
the setback is 33%>20% of the previous plan dimension in the direction of the setback. Layouts of the
buildings are reported in figures 5.1-5.5.
With the proposed algorithm the optimal brace configuration has been found for each building. Results
are obtained by imposing the performance requirements in terms of drift displacements for different
earthquake levels according to the performance based design philosophy, as shown in table 5.2.
Table 5.2. Mechanical characteristics of the elementary frame.
Results can be used as guidelines for designing braces in similar buildings. Especially optimal identified
positions in plan and elevation give criteria for choosing the optimal position in other irregular
buildings.
However, in order to give general criteria, several analyses have been carried out, and results obtained
by the optimization procedure have been interpreted on the basis of those analyses.
Figure 5.1. Optimal bracing configuration for the “regular building” (type 1).
EARTHQUAKE LEVEL DRIFT LIMIT
FREQUENT EARTHQUAKE (operational limit state) TR = - years -%
OCCASIONAL EARTHQUAKE (occupancy limit state) TR = 225 years 0.4%
RARE EARTHQUAKE (life safety limit state) TR = 475 years 1.0%
VERY RARE EARTHQUAKE (collapse prevention limit state) TR = 2475 years 1.5%
68
Figure 5.2. Optimal bracing configuration for the “dumpbell shaped building” (type 2a).
Figure 5.3. Optimal bracing configuration for the “L-shaped building” (type 2b).
69
Figure 5.4. Optimal bracing configuration for the “asymmetric re-entrant profile building” (type 3a).
Figure 5.5. Optimal bracing configuration for the “symmetric re-entrant profile building” (type 3b).
From the analysis of results on selected case studies, the following general criteria have been carried
out:
1. braces are more effective in the central bays of a structural frame rather than in the lateral bays; in
fact, central bays offer higher constraint to the braces and the vertical tension forces induced by braces
are more easily balanced by vertical loads;
2. braces are more effective in external frames of a structural system rather than in the inner frames; in
fact, inner frames are more stiff than external frames due to interaction with adjacent frames. Braces
interact with frames in which are placed and their actual stiffness is lower if they are inserted in more
rigid frames.
70
3. braces are more effective in bays inside the in-plan setbacks; these bays are less stiff than the
adjacent bays. If the braces are placed in setback bays, the adjacent bays guarantee a good constraint
and the actual stiffness of the braces benefits from this condition;
4. braces are more effective in bays adjacent to the in-elevation setbacks; these bays are less stiff than
the adjacent bays. If the braces are placed in setback bays, the adjacent bays guarantee a good constraint
and the actual stiffness of the braces benefits from this condition;
5. braces in the corner of the building increase the torsional stiffness; however, since the braces in
external bays are less effective, the increment in torsional stiffness can be not significant. In the case
presented in figure 5.6, the increment of the torsional radius using braces in corner bays respect than in
central bays is 4% but the increase of torsional stiffness is only 3%. In fact, the actual stiffness of braces
in lateral bays is lower than in central bays (see point 1) due to less effective constraint provided by the
frame.
Figure 5.6. Torsional radius for different bracing configurations (plan view).
6. bracing configurations that allows clear paths for the forces carried by braces are preferred: the
braces should be continuous from the top to the bottom of the building. Furthermore, in order to reduce
the forces induced in the frame, a larger number of smaller braces is preferred. In order to have a better
path of forces in bracing systems, different brace configurations can be effectively used, as shown in
Figure 5.7.
Figure 5.7. Different bracing configurations in terms of path of forces.
5.2. Performance analyses of steel techniques for vertical elements
5.2.1. R.C. benchmark
5.2.1.1. Buckling Restrained Bracings (BRB) The BRB element is characterized by the same behaviour in compression as in tension because of the
core plate which absorbs the loads and by yielding its dissipating seismic energy while the steel tube
and the infill material restrain the buckling of the core plate (figure 5.9). The BRB’s, pinned at the ends,
are installed in the external frames of the RC building, as it can be seen in figure 5.8.
71
(a)
(b)
Figure 5.8. STEELRETRO reference benchmark RC building model and BRB system distribution
b. Elastic and design response spectrum
Figure 5.9 Geometry and components of the tested BRB (CEMSIG)
For the original reinforced concrete structure, a seismic behaviour factor q = 1.5 was used. For the
reinforced concrete structure retrofitted with BRB system, the seismic behaviour factor q amounted to 4
(see Figure 5.8.b). The BRB design was made using a q = 4 and started with a steel core cross section of
3 cm2 (1 cm thickness and 3 cm wide). The following BRB core plate cross section were sized for the
frames in X direction: ground floor = 2 cm x 4 cm; 1st level = 1cm x 4 cm; 2
nd level = 1cm x 3cm. BRB
core plate cross section in Y direction were: ground floor = 2 cm x 3 cm; 1st level = 1 cm x 5 cm; 2
nd
level = 1 cm x 3 cm. The BRB cross section is represented in the model as constant along the length.
Therefore, a reduction of the axial stiffness K [KN/m] is applied (Table 5.3). For this particular case the
BRB cross section was made of S235 steel and the geometry of the core was defined so that all braces
have the same active length of 1.7 m (figure 5.9). Thus, for this active length, the yield displacement
amounts to Δy = 1.9 mm. The estimation of the ultimate displacement Δu was based on the results of the
experimental tests carried on BRB elements. Based on these results, ductility ratios Δu/Δy were
estimated for tension and compression amounted to 22. In order to obtain the adjustment of the design
strengths (maximum compression strength Cmax and maximum tension strength Tmax), the following
formulas were applied:
Tmax
= wRyfyA ; C
max= wbR
yfyA (5.6)
where, fy is the yield strength, Ry is the ratio of the expected yield stress to the specified minimum yield
stress fy (may be considered equal to 1). The values of the compression adjustment factor β=1.2 and a
strain hardening adjustment factor ω=1.9 was obtained from the experimental tests, using the following
formulas:
b =Tmax/ C
max; w = T
max/ f
yscA( ) (5.7)
where: fysc is the measured yield strength of the steel core.
The inelastic behaviour of BRB system was modelled considering the concentrated tri-linear plasticity
curve with strain hardening and strength degradation of 0.8 from maximum capacity, according to
FEMA356 (see figure 5.10)
72
Figure 5.10. BRB tri-linear model: a. on X direction; b. in Y direction
The modelling parameters and the acceptance criteria given by FEMA 356, for steel braces in tension,
were used in the evaluation of the performance of BRB elements. The results of the experimental tests
on BRB specimens showed an available ductility of around 22t, which is twice the value given by
FEMA 356, i.e. 11Dt. The BRB tri-linear model used in the present analysis is characterized by the
following parameters (table 5.3):
Table 5.3: BRB modelling parameters for the final benchmark analysis
Seismic performance of RC structure was computed by means of static nonlinear (pushover) and
compared to the preliminary results obtained using a simplified response spectra analysis. In order to
assess whether the building can achieve the rehabilitation objectives, the following methodology is
applied:
a non-retrofitted frame is analyzed in order to determine the history of plastic hinges;
if necessary, a local retrofitting of the elements (beams, columns) would be adopted until a
favourable plastic mechanism is obtained;
a Global Retrofitted frame is analyzed in order to determine the history of plastic hinges;
if necessary, a local retrofitting of the elements (beams, columns) is adopted until a favorable
plastic mechanism is obtained. It is also checked that the dissipative system (i.e. BRB) be properly
designed. If not, the system is adjusted so as to meet the requirements of a favorable plastic
mechanism;
static nonlinear analysis using N2 method is employed for the evaluation of performance for each
case.
Pushover analysis were performed on 3D models for the initial structure and for the retrofitted
structures (local, global and both). Following the results of the pushover analysis on X direction it may
be seen that the initial structure MRF and the initial structure with local retrofitting MRF + FRP have a
limited ductility and do not attain the displacement demands for LS and CP levels. The benefit of local
retrofitting is reduced. When the global retrofitting is accomplished MRF+BRB, the behaviour is much
improved. The stiffness and the strength increase, and the structure attains the LS performance. The
structure cannot attain the CP level, due to the failure of the concrete structure. The contribution of the
local retrofitting is again very limited (MRF+BRB+FRP). Following the results of the pushover
analysis on Y direction it may be seen that the initial structure MRF has limited ductility and does not
attain the displacement demand for LS level. When the initial structure is retrofitted with FRP (MRF +
FRP), the strength and the stiffness do not change but the ductility increases. The structure attains the
BRB (fy=235 N/mm2) force - displacement - on X direction
-400
-200
0
200
400
-0.08 -0.06 -0.04 -0.02 0 0.02 0.04 0.06 0.08
Displacement [m]
Fo
rce
[K
N]
BRB ground floor [2x4] cm2 BRB 1'st level [1x4] cm2 BRB 2'nd level [1x3] cm2
Compression
Tension
BRB (fy=235 N/mm2) force - displacement - on Y direction
-300
-200
-100
0
100
200
300
-0.08 -0.06 -0.04 -0.02 0 0.02 0.04 0.06 0.08
Displacement [m]
Fo
rce [
KN
]
BRB ground floor [2x3] cm2 BRB 1'st level [1x5] cm2 BRB 2'nd level [1x3] cm2
Compression
Tension
Final Benchmark analysis
Modeling Curve type triliniar (FEMA/ASCE model)
Material steel S235
Aria-core c.s. Ac [cm2] 1x3 (tested cross section)
Core length Lc [m] 1.7
Yielding displacement Δy [mm] 1.9
Ductility displacement µ 22 (cyclic AISC)
IO 0.5Δt
LS 14Δt
CP 18Δt
BRB effective stiffness Ke considered
Compression adjustment
factor β
1.2 (minimum from cyclic
ECCS+AISC)
Acceptance criteria
(modified FEMA356/ASCE41
acceptance criteria for
braces in tension)
BRB properties
Strain hardening adjustment
factor ω
1.9 (minimum from cyclic
ECCS+AISC)
73
displacement demand for LS but not for CP level. When the global retrofitting is accomplished
MRF+BRB, the behaviour is much improved. The stiffness and the strength increase, and the structure
attains the LS performance. The structure cannot attain the CP level, due to the failure of the concrete
structure. The contribution of the local retrofitting is limited (MRF+BRB+FRP).
Figure 5.10.a. Performance of the Benchmark building retrofitted using different techniques (global
approach – BRB – and local strengthening – FRP)
5.2.1.2. Steel and Composite Steel Concrete Shear wall In this part of the report steel and composite shear walls to retrofit and upgrade existing reinforced
concrete buildings are investigated with finite element analysis and compared with analytical models
provided in the literature. Furthermore, the most suitable shear walls are selected and are applied at the
RC-benchmark building to evaluate the obtained structural performance using this strengthening
method.
Three models have been generated to find an optimized solution of a steel plate shear wall, in which
both steel panel and frame are utilized in a similar way. As it is difficult to evaluate the proportions
between steel frame and steel plate, a parametrical study has been conducted varying the thickness (3-8
mm), the width-to-height ratio, the flexibility of connections and typology of vertical surrounding
flanges.
Concerning the RC-benchmark building to which the Shear Wall retrofitting technique has been
applied, its main deficiencies are: low bearing capacity and stiffness in X- and Y-direction, weak storey
failure of the ground floor in Y-direction (strong-beam/weak-column failure), torsion sensitivity (1st
Eigen-period for torsion is in the range of the 1st Eigen-period in Y-direction), inadequate stirrup
spacing of beams and columns for extensive plastic rotation and insufficient anchoring of longitudinal
reinforcement in moment resisting frames.
The demand on the structure is illustrated by the AD-response spectra in figure 5.11. The high capacity,
excellent ductility and sufficient stiffness of steel and composite shear walls provides following
strategies for strengthening the RC-benchmark building: A) increasing strength (continuous line); B)
increasing strength and utilisation of existing ductility (dashed line), C) increasing strength and
increasing ductility by local strengthening (dotted line).
The first option is to increase the capacity of the structure only by strength and provide a sufficient
stiffness. Hence, the structure keeps nearly elastic and the unfavourable ductility of the original
structure do not affect the structural performance. The second strategy is to utilise the existing ductility
of the structure and increase the strength in a limited range. This leads to a lower amount of
strengthening material and the reaction forces for the foundation can be reduced. The last possibility is
to combine the global strengthening of the structure by shear walls with local strengthening methods to
increase the ductility in plastic hinges. This leads to a further reduction of the connection forces,
however the effort for assembling these techniques will rise.
Three different types of steel shear walls were analysed with FE-models: (i) a steel plate in a moment
resisting frame of I-profiles, (ii) a steel plate in a hinged frame of I-profiles and (iii) a steel plate with
vertical flanges as columns. Further a composite shear wall with a moment resisting frame of I-profiles
was analysed. The flexible and the rigid system show similar behaviour concerning the maximum shear
force. The maximum shear force of the system with vertical flanges is about 25 % lower than the
bearing capacities of the systems with I-profiles as columns. Furthermore, the displacements at the peak
forces of the system with vertical flanges are about one half of the displacements of the two other
systems. This system has only low shear capacity and low ductility. The displacements of the flexible
74
system are about 25 % higher for 5 to 7 mm thick plates than the displacements of the rigid system and
for a 3 mm thick plate the displacement of a rigid system is about 37 % higher than that of the flexible
system. Therefore, either systems with moment-rigid connections and thin steel plates or systems with
flexible connections and thicker panels should be used. This decision, however, depends on the bearing
capacity of the frames of the existing building to ensure that the shear walls can develop their strength.
The main common characteristic of composite and steel shear walls in their performance is that an
increasing thickness of the steel plate leads to an increasing bearing capacity but thick infill panels have
a lower bearable displacement than thin one. The maximum shear forces of the composite shear wall are
about 50 % higher than the peak shear forces of the steel shear wall due to the stiffening effect of the
concrete wall. The displacement at peak force however is about 158 mm for the composite shear wall
and about 175 mm for the steel shear wall for an 8 mm thick plate. Furthermore, the stiffness of
composite shear walls is considerable higher than steel shear walls. Again, the application of the steel
shear wall or of the composite shear wall is dependent on the specific requirements of the building,
which has to be retrofitted.
Figure 5.11. Possible strengthening strategies by shear walls for the RC-benchmark building
To obtain a sufficient structural performance of the retrofitted structure, the strength of shear wall
should be at minimum higher than 700 kN even if the local ductility will be increased. The minimum
ultimate displacement of the shear wall for strengthening techniques, which utilize the existing ductility
of the original structure, should be higher than 197/3 = 66 mm in X-direction and 174/3 = 58 mm in Y-
direction. The minimum initial stiffness of the shear wall should be higher than the original structure to
activate them with an adequate displacement (X-direction K > 14 000 kN/m, Y-direction K > 16 000
kN/m).
As shown in the diagrams below, all kind of shear walls are applicable excepting the shear wall with
flanges as frame. As sufficient strength and stiffness can also be reached by steel shear walls, they are
preferred to composite shear walls, which need more effort to assemble them. Furthermore, thins steel
plates with a rigid frame are chosen to obtain a clear failure mechanism in the steel plate instead of an
interaction between steel plate and frame.
In the following analysis the RC-benchmark building is retrofitted by using two different strategies:
Type A: The structure is retrofitted by strength applying steel shear with 3 respectively 4 mm plate
thickness and b x h = 4.0 x 2.8 m respectively b x h = 4.5 x 2.8 m (X- and Y-direction).
Type B: The structure is retrofitted by strength and increasing the local ductility, where steel shear are
used with a plate thickness of 5 mm and dimension of b x h = 1.4 x 2.8 m.
The shear walls are applied over the whole height of the building, symmetrically, at the outer areas but
not in both directions at one.
The capacity curve of the retrofitted structure in X-direction and Y-direction are obtained by a non-
linear pushover analysis. Similarly to the analysis of the original structure the “collapse” of the building
is defined at the maximum base shear ignoring a decreasing branch of the load-displacement curve
(force controlled loading). The shear wall is modelled by a concentric bracing, where the load-
75
displacement curve of the bracings is defined in such a way that the performance is equal to the load-
displacement characteristic of the shear wall obtained by the finite element analysis (figure 5.12).
Figure 5.12. Type of analysed shear walls: steel shear wall with rigid connections (a), with hinged
connections (b), with flanges (c), composite shear wall (d)
Figure 5.13. Possible strengthening with shear walls, ground view.
(a) (b)
(c) (d)
76
Figure 5.14. Possible strengthening with shear walls, Section axis A and E.
Figure 5.15. Possible strengthening with shear walls, Section axis 1 and 6
direction t [mm] frame dimensions
Steel shear wall type A X 4 HEB300 4.0 x 2.8
Y 3 HEB300 4.5 x 2.8
Steel shear wall type B X 5 HEB300 1.4 x 2.8
Y 5 HEB300 1.4 x 2.8
Table 5.4. Parameters of steel shear walls for strengthening strategy A and B
Figure 5.16. Structural model for shear wall Figure 5.17. Load-displacement characteristic of
shear wall
0
500
1000
1500
2000
2500
0 50 100 150 200 250 300 350
displacement [mm]
F [
kN
]
77
The base shear force-displacement curves of the retrofitted structure show a remarkable higher capacity
and stiffness. The performance is similar in X- and Y-direction, even if the ground floor in Y-direction
is still weaker than the storey above (figure 5.18, 5.19, 5.20 and 5.21).
Figure 5.18. Base shear force-displacement curves
in X-direction (4 span), strategy A
Figure 5.19. Storey drift over the height of the
structure in X-direction (4 span), strategy A
Figure 5.20. Base shear force-displacement curves
in Y-direction (5 span), strategy A
Figure 5.21. Storey drift over the height of the
structure in Y-direction (5 span), strategy A
The performance of the retrofitted structure is assed with the N2-method in accordance with EN 1998
Annex B. The Eigen-period of the equivalent SDOF is between TB and TC, hence the capacity diagram
intersects the demand spectra at the upper plateau. This leads to very high base shear forces and
connection forces but low top storey displacements. The structure remains nearly elastic which means
that the required ductility ratio is 1.0. The base shear force-displacement curves with strategy B show a
moderate increase in capacity and stiffness in relation to the original structure. Furthermore, the
ultimate displacement can be enhanced. Again, the ground floor in Y-direction is still weaker than the
storey above.
Figure 5.22. Demand spectra vs. capacity diagram
in X-direction (4 span), strategy A
Figure 5.23. Demand spectra vs. capacity
diagram in Y-direction (5 span), strategy A
78
Figure 5.24. Base shear force-displacement
curves in X-direction (4 span), strategy B
Figure 5.25. Storey drift over the height of the
structure in X-direction (4 span), strategy B
Figure 5.26. Base shear force-displacement
curves in Y-direction (5 span), strategy B
Figure 5.27. Storey drift over the height of the
structure in Y-direction (5 span), strategy B
By utilization of some of the improved local ductility the maximum base shear force can be reduced
significantly, while the maximum storey drift is still acceptable. The required ductility ratio of 1.8 to 2.1
is moderate and can be easily reached by local strengthening techniques.
The main results of push-over analysis and N2-method assessment for the original structure and the
strengthening strategies A and B are summarized in the tables below. Strategy A (strength) as well as
strategy B (strength and ductility) leads to an enhancement of the structure, which fulfil the assumed
seismic requirements. The advantage of strategy A is the very small top displacement and the available
ductility of the structure is sufficient without any local strengthening. However, very high forces have
to be transferred by the connections and into the foundation. Strategy B leads to remarkable smaller
connection and foundation forces, however local strengthening is necessary to achieve the required
local ductility.
In general, the selection of the most suitable shear wall is dependent on the specific requirements of the
building, which has to be retrofitted. The high capacity, excellent ductility and sufficient stiffness of
steel and composite shear walls provides following strategies for strengthening the RC-benchmark
building: increasing strength, increasing strength and utilisation of existing ductility, increasing strength
and increasing ductility by local strengthening.
Figure 5.28. Demand spectra vs. capacity diagram
in X-direction (4 span) for retrofitting strategy B
Figure 5.29. Demand spectra vs. capacity diagram
in Y-direction (5 span) for retrofitting strategy B
79
Two types of steel shear walls are applied at the RC-benchmark building to evaluate the effectiveness
of this strengthening method:
Type A: increasing strength (shear wall: t = 3 resp. 4 mm; b x h = 4.0 x 2.8 m resp. b x h = 4.5 x 2.8 m)
Type B: increasing strength and local ductility (shear wall: t = 5 mm; b x h = 1.4 x 2.8 m)
Strengthening with strategy A as well as with strategy B leads to an enhancement, which enables the
structure to bear the assumed seismic loads. The advantage of strategy A is the very small top
displacement and the available ductility of the structure is sufficient without any local strengthening.
However, very high forces have to be transferred by the connections to the existing structure and into
the foundation. Strategy B leads to remarkable smaller connection and foundation forces, however local
strengthening is necessary to achieve the required local ductility.
Two other solutions, shown in figure 5.30, have been analysed using partial-width shear walls whose
mechanical parameters are listed in Tables 5.4 and 5.5. A part from the steel shear walls, the main
differences between the two solutions is the presence in D configuration of local strengthening
interventions in order to achieve the required local ductility. The results of Nonlinear Static Analysis
performed on the C and D solutions are represented in figures 5.31 and 5.32, showing the better
performance of D configuration able to satisfy the safety assessment also at CP limit state and
presenting a more ductile behaviour.
(a)
(b)
Figure 5.30. Partial-width shear walls: a) configuration C; b) configuration D.
direction storey number t [mm] steel grade Local strength.
Steel shear
wall type
C
Y 1 4 6 S235 no
2 4 6 S235 no
3 4 4 S235 no
X 1 4 6 S235 no
2 4 4 S235 no
3 4 4 S235 no
Table 5.4 Mechanical parameters of shear walls in configuration C.
direction storey number t [mm] steel grade Local strength.
Steel shear
wall type
D
Y 1 2 5 S235 yes
2 2 5 S235 yes
3 2 4 S235 yes
X 1 2 5 S355 yes
2 2 4 S235 yes
3 2 4 S235 yes
Table 5.5 Mechanical parameters of shear walls in configuration D.
80
a) b)
c) d)
Figure 5.31. Nonlinear Static Analysis of C retrofitting configuration: a) and c) ADRS representation
(pushover X and Y); b) and d) interstorey drift profiles (pushover X and Y).
a) b)
c) d)
Figure5.32. Nonlinear Static Analysis of D retrofitting configuration: a) and c) ADRS representation
(pushover X and Y); b) and d) interstorey drift profiles (pushover X and Y).
5.2.1.3. Light Gauge Steel panel The application of LGS walls was considered trying to upgrade the seismic performance of the building
adopting two approaches: one focused into an increasing of the strength and the other in which a ductile
behaviour is considered also, figure 5.33. Obviously, the stiffening of the frame by LGS walls resulted
in an increase of the force demand as the structure is shifted in the lower period range of the spectrum
and this it is a positive aspect considering the deformation capacity of the frame limited to low values –
dlim. The design process was primarily based on strength and less on ductility and then the ductility
involvement was progressively taken into account modifying the proposed solution. The increase of
stiffness can be achieved by one the two schemes presented in figure 5.34. From the theoretical point of
0.000
1.000
2.000
3.000
4.000
5.000
6.000
7.000
8.000
9.000
10.000
0.000 0.050 0.100 0.150 0.200 0.250
ac
ce
lera
tio
n [
m/s
2]
displacement [m]
0.00
1.00
2.00
3.00
0.00% 0.50% 1.00% 1.50% 2.00%
flo
or
Drift
IO
LS
CP
floor1
floor2
floor3
0.000
1.000
2.000
3.000
4.000
5.000
6.000
7.000
8.000
9.000
10.000
0.000 0.050 0.100 0.150 0.200 0.250
ac
ce
lera
tio
n [
m/s
2]
displacement [m]
0.00
1.00
2.00
3.00
0.00% 0.50% 1.00% 1.50% 2.00%
flo
or
Drift
IO
LS
CP
floor1
floor2
floor3
0.000
1.000
2.000
3.000
4.000
5.000
6.000
7.000
8.000
9.000
10.000
0.000 0.050 0.100 0.150 0.200 0.250
ac
ce
lera
tio
n [
m/s
2]
displacement [m]
0.00
1.00
2.00
3.00
0.00% 0.50% 1.00% 1.50% 2.00%
flo
or
Drift
IO
LS
CP
floor1
floor2
floor3
0.000
1.000
2.000
3.000
4.000
5.000
6.000
7.000
8.000
9.000
10.000
0.000 0.050 0.100 0.150 0.200 0.250
ac
ce
lera
tio
n [
m/s
2]
displacement [m]
0.00
1.00
2.00
3.00
0.00% 0.50% 1.00% 1.50% 2.00% 2.50%
flo
or
Drift
IO
LS
CP
floor1
floor2
floor3
81
view there is no significant difference between the two schemes. In practices scheme (a) is more
feasible because consider only an improvement of the strength and of the stiffness, protecting so all
existing r.c. members; the other approach, (b), assumed an involvement of the existing structure in the
resistance and so the local reinforcement of the elements was expected.
In the two main directions of the structure, the steel plates were used in the bays presented in Figure
5.34 (a and b). The first arrangement is idealized, as very often architectural considerations will impede
the use of such symmetrical strengthening scheme. As principle, the shear walls should be placed (i) as
symmetrically as possible in both directions and (ii) as close to the outer frames as possible, in order to
increase resistance to torsion.
Several thicknesses of LGS shear walls have been tried in order to achieve an optimum performance for
the structure. The results presented here refer to the LGS plate dimensions from Table 5.6.
Figure 5.33. Suggested use of the LGS steel shear walls.
a) b)
Figure 5.34. Possible strengthening with LGS shear walls (a) W1, (b) W2.
Dir.
Axis
L
(m) Plates
H
(m)
L
(mm)
α
(deg)
t
(mm)
fy
(N/mm2)
Fplate
(kN)
Kplate
(kN/m)
Fwall
(kN)
Kwall
(kN/m)
X A & E 4.6 4 3.35 1150 38.6 1 350 392 34256 1570 137023
Y 1&6 4.1 3 3.35 1367 41.2 1 350 474 42094 1423 126283
Table 5.6 LGS shear walls in X and Y directions.
The deformed shapes from the two direction pushover are present in Figure 5.35 while the capacity and
demand diagrams are presented in Figure 5.36 for this rehabilitation case (W1). It can be observed that
the soft storey behaviour of the ground floor is preserved in this case. As it can be seen, the strength of
the structure increases in both directions so that the ductility requirements are very low (μreq-x=1.20, μreq-
y=1.13). The LGS walls, together with the RC frame provide sufficient strength almost for an elastic
response; and the strength is enough for a design with q=1.5.
It is important to note that in this case, the shear walls are modelled as simple shear links between the
two levels they connect. This means that shear walls are connected to the frames only in the corner, and
local forces exercised on RC elements are not taken into account. The most important of these local
effects are: (i) the anchoring of the shear wall to the RC elements and (ii) the uplift effect of the wall on
the foundation on the tension side. In order to account for the local effects of the LGS shear walls, a
F
d
Fr.c.
FLGS
s
F
d
Fr.c.
FLGS
dlim dlim
82
more elaborate model was developed where strips play the role of shear wall. Several simplifications
are accepted in this case of modelling too: (i) the strips are made of bi-linear yielding steel material, (ii)
they are very thin, t=1mm, and they can act only in tension, (iii) i.e. they are meant to model the tension
field effect in a very this steel plate, so shear and compression are neglected, (iv) strips are placed at an
angle of 45°, so the presumed tension field is forced to develop at this angle. This is not always the
case, as the tension field in a thin steel plate develops under an angle depending on the dimensions of
the plate.
a) b)
Figure 5.35. Deformed shape before failure from pushover in (a) X and (b) Y directions
a) b)
Figure 5.36. Demand and capacity diagram of the equivalent SDOF system (Annex B, EN 1998)
In these models, at the base of the shear plates has been connected to a IPE500 base girders, which are
supplementary placed between the columns. The modelling of the shear walls as an equivalent shear
element between the floor levels gives a very conservative estimate of the strength and stiffness. This
happens because the used formulations are based on the supposition that the frame bordering the LGS
wall is perfectly rigid and full-strength. However, the deformations of the RC elements also contribute
to the overall displacement, limiting the effectiveness of the LGS wall. Even with the modelling of the
LGS wall as strips, several concerns remain, as: (i) it is supposed that strips do not fail at end
connections and (ii) the transverse compression (and consequent buckling) of the LGS plate can lead to
the formation of important local stress concentrations, and high strains that can further reduce the
capacity of the LGS wall. The W2 model developed for having a more ductile behaviour the technique
of the strip modelling has been adopted, figure 5.37, and the comparison between the two approaches
for the W2 configuration is reported in figure 5.38. In figure 5.39 is reported the structural assessment
performed on the W2 model considering the two modelling techniques for the LGS walls: a and c are
related to the first approach using a single spring for the shear panel; b and d are related to the results of
the strip model. In this last case it is possible to appreciate that the structural capacity in terms of
maximum displacement is larger than the expected performance point and that the structural solution, as
expected exploits larger ductility levels.
0
2
4
6
8
0 0.05 0.1 0.15 0.2
Sed(m)
Se(m
/s2)
μreq=1.20
T*=0.37
0
2
4
6
8
0 0.05 0.1 0.15 0.2
Sed(m)
Se(m
/s2)
μreq=1.13
T*=0.42
83
Figure 5.37. Modeling the LGS shear walls as
inclined strips (W2-Strips)
Figure 5.38. Pushover curves of the W2 and
W2-Strips configurations
a) b)
c) d)
Figure 5.40. Demand and capacity diagram of the equivalent SDOF system (Annex B, EN 1998): (a, c)
X and Y direction of the W2 model, (b, d) X and Y direction of the W2-strip model
As mentioned, one method to rehabilitate the structure would be to make it lighter. The solution of
replacing the roof with a LGS trapezoidal sheeting, and replacing the walls with LGS walls (e.g.
NORDICON walls) is examined in the following section. If the self-weight of the new LGS elements is
presumed to be 25kg/m2 (i.e. down from 200daN/m
2 for roof, and 250daN/m
2 for walls), the structures
mass is reduced in the EQ combination from 1357.6t to 1112.7t. The new distribution of the masses and
horizontal loads is summarized in table 5.7. The capacity and demand curves for this case are presented
in figure 5.41. It is clear that this solution can not improve the performance to the desired level but it is
worth noting that in Y direction could reach expected performance whether the structural members are
largely and extensively subjected to a local retrofitting process. Summary of the data from figure 5.41 is
also in table 5.8. The initial r.c. structure has several potential weaknesses in an eventual earthquake
loading scenario:
0
1000
2000
3000
4000
5000
0.00 0.05 0.10 0.15
dn(m)
Fb(k
N)
X-Strip
Y-Strip
X - W2
Y - W2
0
2
4
6
8
0 0.05 0.1 0.15 0.2
Sed(m)
Se(m
/s2)
μreq=2.09
T*=0.49
0
2
4
6
8
0 0.05 0.1 0.15 0.2
Sed(m)
Se(m
/s2)
μreq=1.71
T*=0.56
0
2
4
6
8
0 0.05 0.1 0.15 0.2
Sed(m)
Se(m
/s2)
μreq=2.32
T*=0.78
0
2
4
6
8
0 0.05 0.1 0.15 0.2
Sed(m)
Se(m
/s2)
μreq=1.81
T*=0.94
84
The stiffness is reduced in both directions, resulting in exaggerated vibration periods (1.25s,
1.51s). If it is accounted that the concrete is in partially cracked state, the vibration periods
would be even higher;
Strength is insufficient in both directions, resulting in large ductility demands (i.e. ductility
factors 4.5 and 2.5);
Ductility is very limited in both directions, mostly because columns are loaded with high axial
forces. In all cases, the failure during the pushover process occurred by crushing of the
compressed concrete in some columns. In fact this phenomenon is limiting the ability of the
structure to deform laterally in the non-linear range;
In the X direction, the structure is a weak column strong beam structure, vulnerable to forming
storey mechanisms.
After identifying these structural problems several methods to rehabilitate the structure have
been tried:
o by using LGS shear walls;
o by making the structure lighter using LGS external walls and roofs;
o by bracketing the columns of the structure in order to increase bending strength and the
ability to sustain plastic hinge rotations.
If presumed that the lateral displacement supply of the structure is unchanged (i.e. no intervention to the
vertical load transmission path is made), it has been shown that the structure can be retrofitted to satisfy
earthquake design criteria only by using stiff horizontal load bearing systems (e.g. shear walls). One
version of LGS shear wall refurbishment has been given as example.
Level mi(t) hi(m) Φi mi×Φi mi×Φi2 hi×mi×Φi
F(%) /
Level
X o
r Y
dir
ecti
on
1 410.8 3.9 0.31 128.7 40.3 501.8 20.5
2 400.6 7.3 0.59 234.9 137.7 1714.7 37.4
3 238.3 10.65 0.86 203.8 174.4 2170.9 32.5
Roof
40.8 11.55 0.93 37.9 35.1 437.3 6.0
22.2 12.45 1 22.2 22.2 276.7 3.5
Total: 1112.7 627.5 409.8 5101.5 100.0
Table 5.7 Distribution of the horizontal loads in the 3D structure
(a) (b)
Figure 5.41. Capacity & demand of structure with LGS wall & roof
5.2.1.4. Steel concentric and eccentric bracings Steel bracing systems for retrofitting r.c. frame structures are widely used and analyzed in last decades
by several authors. Both concentric and eccentric bracing solutions were studied for seismic retrofitting
of the r.c. benchmark structure, analyzing different bracing schemes for the two main directions.
Concentric bracing systems were modelled taking into account geometrical imperfections according to
EN 1993-1-1:2005 introducing a precamber equal to = L/500. In figure 5.42.a and b is illustrated a
simple concentric bracing scheme with the initial precamber and the relative cyclic behaviour. The
0
2
4
6
8
0 0.05 0.1 0.15 0.2
Sed(m)
Se(m
/s2)
μreq=4.40
T*=1.13
0
2
4
6
8
0 0.05 0.1 0.15 0.2
Sed(m)
Se(m
/s2)
μreq=2.32
T*=1.36
85
nonlinear material behaviour of the bracing system was modelled by the Menegotto-Pinto model (see
OPENSees Manual and Uriz and Mahin, 2008) choosing steel grade equal to S235.
Initial RC frame LGS wall (W2 Strip) Light roofs & walls Bracketed Column
X Y X Y X Y X Y
dmax*(m) 0.039 0.100 0.054 0.086 0.037 0.120 0.131 0.143
Fmax*(kN) 490 717 1507 1591 434 687 1323 1229
dy*(m) 0.025 0.053 0.029 0.045 0.022 0.051 0.057 0.074
T*(s) 1.25 1.51 0.78 0.94 1.13 1.36 1.15 1.37
Se-T* (m/s2) 2.76 2.29 4.45 3.66 3.05 2.55 2.99 2.52
Sed-T* (m/s2) 0.110 0.132 0.068 0.082 0.099 0.119 0.101 0.120
qu* 4.44 2.51 2.33 1.82 4.41 2.33 1.78 1.62
dt*(m) 0.110 0.132 0.068 0.082 0.099 0.119 0.101 0.120
μreq 4.43 2.51 1.71 1.81 4.40 2.32 1.78 1.62
μava 1.58 1.90 3.5 1.90 1.63 2.35 2.31 1.93
dt(m) 0.163 0.196 0.101 0.122 0.152 0.182 0.150 0.178
Table 5.8 Summary of the properties of the equivalent SDOF (Annex B, 1998) in all strengthening
cases
Among eccentric bracing systems, the inverted-Y structural scheme (see figures 5.43.a and 5.43.b) was
selected for the seismic retrofitting of the r.c. benchmark, choosing short links according to Italian and
European standards (NTC08, EN1998-1), whose shear and bending behaviours are represented in figure
5.43.c and d. The link was modelled by means of ZeroLenghtSectionElement (see OPENSees Manual),
using also in this case the Menegotto-Pinto material model and a steel grade S235.
a) b)
Figure 5.42. Adopted concentric bracing scheme and cyclic behaviour.
a) b)
c) d)
Figure 5.43. Eccentric bracing systems: a) adopted scheme, b) finite element model, c) shear and d)
bending behaviour of the link.
-750
-500
-250
0
250
500
750
-75 -50 -25 0 25 50 75
Top Displacement [mm]
Fo
rce [
kN
]
86
Several bracing schemes were tested for the seismic retrofitting of the r.c. benchmark structure in the X
and Y directions: for each solution the nonlinear static analysis (N2 method) was performed in order to
improve the structural behaviour. In figure 5.44 are reported the best solutions for X and Y directions,
respectively with HEB 140 steel profile in three braced bays HEB140 (X) and with HEB 120 steel
profile “tree” configuration (Y). In Figure 5.45 and 5.46 are reported the ADRS plane representation,
collapse mechanism and ductility assessment for the two final proposed solutions. The X direction
solution is the most suitable in terms of added stiffness, strength and achieved ductility, giving a
collapse mechanism dominated by the bending of the first floor beam edge sections. In the Y direction,
it can be observed that among the various solutions, the one reported here seems to be the most suitable
providing at the same time a substantial improvement of stiffness, strength and ductility. Compared to
other tentative solutions, the Y dir. configuration is able to provide enough stiffness, strength and
ductility to the retrofitted structure because it interested more columns giving lower values of axial
forces.
a) b)
Figure 5.44. Concentric bracing schemes: a) X direction; b) Y direction.
a)
b)
- Collapse mech.: bending moment of beam edge
section
- Requested ductility 1.27
- Available ductility 1.73
Figure 5.45. X direction retrofitting solution: a) ADRS format representation, b) collapse mechanism
and ductility assessment.
a)
b)
- Collapse mech.: bending moment beam section
- Req. ductility 1.59
- Ava. ductility 2.16
Figure 5.46. Y direction retrofitting solution: a) ADRS format representation, b) collapse mechanism
and ductility assessment.
Concerning the use of Y-inverted eccentric bracings, several configurations have been tested for
seismic retrofitting of the r.c. benchmark frame. The most suitable bracing scheme for the X and Y
Rottura per flessione trave
0
1
2
3
4
5
6
7
8
0.000 0.020 0.040 0.060 0.080 0.100 0.120 0.140 0.160 0.180 0.200
Ac
ce
lera
zio
ne
[m
/s^
2]
Spostamento [mm]
Capacity Spectrum
Struttura non controventata
Struttura controventata
spettro di risposta struttura controventata
spettro anelastico struttura non controventata
=1.73 >req =1.27
Spettro elastico
Spettro anelastico
Bilineare equivalente
Rottura per flessione trave
0
1
2
3
4
5
6
7
8
0.000 0.020 0.040 0.060 0.080 0.100 0.120 0.140 0.160 0.180
Ac
ce
lera
zio
ne
[m
/s^
2]
Spostamento [m]
Struttura non controventata
Struttura controventata HEB120
Spettro anelastico struttura non controventata
spettro Struttura controventata HEB120
Struttura controventata HEB140
spettro struttura controventata HEB140
Profili HEB140 =2.16 >req =1.59
Spettro elastico
Spettro anelasticoBilineare equivalente
Profili HEB120 =2.32 >req =1.84
Profilo HEB160: Rottura colonna in trazione
Profilo HEB140: Rottura trave secondo solaio
Profilo HEB120: Rottura trave secondo solaio
87
direction are shown in Figures 5.47 and 5.48 in which are also reported the link properties. Figures 5.49
and 5.50 illustrated the capacity curve, the equivalent bilinear model and the ADRS plane assessment
with ductility properties. In particular, it can be observed that among the various solutions, the Y
scheme seems to be the most suitable providing at the same time a substantial improvement of stiffness,
strength and ductility and the best displacement profile with respect to the other solutions.
a)
Link profile = HEA260
e = 400 mm
Vy = 241 kN
y = 0.67 mm
Vu = 362 kN
u = 32 mm
My = 104 kNm
y = 0.00048 rad
Mu = 104 kNm
u = 0.08 rad
Figure 5.47. X1 eccentric bracing scheme and link properties.
Link profile = HEA260
e = 400 mm
Vy = 241 kN
y = 0.67 mm
Vu = 362 kN
u = 32 mm
My = 104 kNm
y = 0.00048 rad
Mu = 104 kNm
u = 0.08 rad
(ground and upper floor)
Link profile =
HEA300
e = 400 mm
Vy = 318 kN
y = 0.67 mm
Vu = 477 kN
u = 32 mm
My = 157 kNm
y = 0.00041 rad
Mu = 157 kNm
u = 0.08 rad
(first floor)
Figure 5.48. Eccentric bracing schemes analyzed in the Y direction with adopted link properties.
a)
b)
- Collapse mechanism: limit shear deformation
in the upper floor link
- Requested ductility 1.41
- Available ductility 2.24
Figure 5.49. X retrofitting solution: a) capacity curve in ADRS format representation, d) collapse
mechanism and ductility assessment.
5.2.2. Masonry benchmark Once the main structural vulnerabilities of the masonry building have been individuated using the linear
model developed using SAP2000, non-linear analyses have to be carried out in order to examine the
performance of steel based intervention techniques in which the coupling of existing masonry walls
with new steel structures will be evaluated.
According to the assumption that the calibration executed in the PROHITECH research project can be
considered valuable also for the STEELRETRO project (adjusting the mechanical values of resistant
properties of masonry elements according to the previously presented values), a refined model of the
masonry benchmark building has been defined using ABAQUS software, figure 5.51.
Rottura per pressoflessione colonna
0.000
1.000
2.000
3.000
4.000
5.000
6.000
7.000
8.000
0.000 0.020 0.040 0.060 0.080 0.100 0.120 0.140 0.160 0.180 0.200
Ac
ce
lera
zio
ne
[m
/s^
2]
Spostamento [m]
Capacity Spectrum
Struttura controventata
Spettro anelastioco Struttura non controventata
Struttura non controventata
Spettro struttura controventata
=2.24 >req =1.41
Spettro elastico
Spettro anelastico
Bilineare equivalente
Raggiungimento scorrimento limite Link
88
As far as the current configuration is concerned the following structural properties and potential
deficiencies have been identified: The structure is rather symmetrical and has similar behavior in the
two main directions. Torsion does not seem to affect the performance. The largest part of the seismic
mass is given by the weight of the wall elements. Both the weight of the floors and the mass coming
from loads is less significant. In the current configuration the biggest problem of the structure is the
lack of diaphragm effect at both the level of the floors ad at the level of the roof. As consequence the
walls are not tied together and local failure is governing the behavior. Realizing an effective tying
between the walls has to be the main priority of any rehabilitation. If floor diaphragm action is realized
the structure would have satisfactory performance in the Z direction. However in the Z direction
supplementary intervention is most probably required.
a)
b)
- Collapse mech.: combined axial force and
bending moment of the column base section
- Requested ductility 1.86
- Available ductility 2.18
Figure 5.50. Y retrofitting solution: a) capacity curve in ADRS format representation, d) collapse
mechanism and ductility assessment.
(a)
(b)
(c)
(d)
Figure 5.51 Abaqus model of masonry benchmark building: (a) 3D model: (b) deformed shape at
collapse; (c) constitutive law in compression; (d) constitutive law in tension.
0.000
1.000
2.000
3.000
4.000
5.000
6.000
7.000
8.000
0.000 0.025 0.050 0.075 0.100 0.125 0.150 0.175
Ac
ce
lera
zio
ne
[m
/s2]
Spostamento [m]
Struttura non controventata
Spettro anelastico struttura non controventata
Spettro struttura controventata
Struttura controventata
Spettro elastico
Spettro anelasticoBilineare Equivalente
=2.18 >req =1.86
Rottura per pressoflessione colonna
89
Based on the observations concerning the behavior of the structure the following rehabilitation
techniques have been tested:
Tying, using tension only ties, of the upper part of the walls.
Establishing rigid diaphragm at the top of the walls.
Rigid diaphragm at roof level, coupled with reinforcement of external ground floor walls with
horizontal LGS strips.
Rigid diaphragm at each floor level, coupled with reinforcement of external ground floor walls
with horizontal LGS strips.
Coupling of steel structures to existing walls
5.2.2.1. Tying the upper end of walls As mentioned earlier one of the problems of the initial structure is that walls are not tied at the top.
Therefore the first rehabilitation solution proposes the full tying of the upper end of the walls, but
without realizing any diaphragm in the structure. Figure 5.52 presents the deformed shape of the
structure, under distributed pushover loads, when the top of the walls have been tied using 24mm,
fy=350N/mm2 steel bars.
Figure 5.52 Pushover deformations with 24mm, fy=350N/mm2 tying at the top of the walls
As it can be observed in figure 5.53 the tying solves part of the problems of the initial structure, namely,
“unzipping” off the walls at vertical connections is mostly eliminated. “Unzipping” (i.e. tension
cracking at the vertical connections) can still be observed at the X direction pushover, at the height of
the second floor slab. This indicates that tying should be available not only at the top of the walls, but
also at intermediate levels in order to completely effeminate unzipping. Whatever, a more acute
problem of the structure is the out of plane bending of walls; which was not eliminated by the tying.
The pushover curves using this configuration are presented in figure 5.54. It can be noted that base
shear force has approximately doubled compared to the initial curves, but out of plane bending of the
walls is not solved by this solution. It appears that the only solution in order to eliminate out/of plane
failure of the walls is to introduce bending stiffness at the midspan of the walls.
5.2.2.2. Rigid diaphragm at the roof level The level to rehabilitation of the structure could be not only to provide tying, but to establish full
diaphragm action at the top of the walls. This, of course, is both more technically challenging and more
expensive procedure compared to just tying; and it supposes the poring of a r.c. slab or the realization of
a horizontal steel truss system at the top of the walls. This case has been modeled by ABAQUS
providing a “rigid body” constraint for nodes at the top of the walls. As it can be observed the cracking
of the walls is uniformly distributed over large areas, which is definitely an advantage of the solution.
However, localized failures are still present: (a) unzipping of vertical wall connection at the level of the
second slab (figure 5.54) and, (b) out of plane failure of an entire wall at the second floor level (figure
5.54). Local intervention and strengthening is supplementary (i.e. besides the roof diaphragm) required
to eliminate these failures.
90
Figure 5.53. Pushover curves of structure tied at top with 24mm, fy=350N/mm2 ties. (a) X (b) Z
direction
Figure 5.54 Views of the deformed shape and distribution of tension cracking for (a) X and (b) Z
direction pushover
The overall performance of the structure is very advantageous in this configuration. As one can observe
from the curves in figure 5.55, the rehabilitated structure possesses sufficient strength and ductility to
withstand the design earthquake load in both X and Z direction. As observed from figure 5.57, this
rehabilitation method providing less strength, but substantially more ductility, than the one involving
rigid diaphragm at each floor level. Also, the disadvantageous soft/storey failure mode, observed in
chapter 4 is completely avoided.
91
Figure 5.55 PSASD plot vs. pushover curve transformed in SDOF format (a) X & (b) Z direction
5.2.2.3. Rigid diaphragm at roof – LGS strips for external walls at ground
floor Even though the previously presented rehabilitation technique seems to provide sufficient performance
in order to fulfill the earthquake design requirements, it has been decided to try to further improve the
properties of the building by strengthening selected walls with horizontal LGS steel strips. The
proposed technical solution is presented in figure 5.56, and it involves the placing and gluing of steel
strips (Astrip=20mm2) in precut slots of 50mm depth. The slots are supposed to be cut at 200mm
intervals. This proposal is inspired from the so called surface-mounted FRP solutions, frequently used
for masonry strengthening; but it is hoped that the steel strips would have better performance due to the
larger elastic modulus of steel compared to FRP. Therefore at small strains of the masonry larger
stresses could be transmitted to the reinforcing strips.
The logic of placing the strips horizontally is illustrated in figure 5.56.b. It is expected that the
interaction between the masonry and strips will provide additional tension strength in the X direction.
Therefore, it is expected that the initial isoshear surfaces (i.e. magenta lines in figure 5.56.b) will be
extended in the positive direction of the X axis (i.e. dashed blur lines in figure 5.56.b), and the shear
strength of the masonry will be increased. Undoubtedly, to test the efficiency of such LGS steel solution
both further theoretical study and testing would be necessary.
Figure 5.56 (a) Technical solution for horizontal LGS strips and (b) expected working principle
92
Figure 5.57. Deformation shapes and distribution of tension cracks for LGS model. (a) X direction and
(b) Z direction pushover
The pushover curves from the models without, and with LGS strengthening, are compared in figure
5.58. It is clear from the figure that, even if the cracking pattern is slightly modified, the overall
performance of the building has not been fundamentally changed by the LGS strengthening. In order to
have a performance improvement, the LGS strips should probably be extended, all the way up to the
roof slab where they can interact with the rigid diaphragm at that level.
Figure 5.58. Comparison of pushover curves
without and with LGS strengthening of selected
external walls (i.e. diaphragm provided only at
roof level)
Figure 5.59. Comparison of pushover curves
without and with LGS strengthening of selected
external walls (i.e. diaphragm provided only at
each slab)
5.2.2.4. Rigid diaphragm at each floor – LGS strips for external walls at
ground floor Finally, an attempt to combining the LGS strips with rigid diaphragm at each floor level has been made.
As previously, LGS strips have been applied only to ground floor, external walls. The deformed shape
and the tensile cracking pattern from pushover loads, in the X and Z directions, are presented in figure
5.57. The most notable difference between this deformation shapes, and the ones presented in figure
5.62 (i.e. same structure but without LGS strengthening), is that, under Z direction forces the initial two
floor mechanism has changed into a single floor mechanism on the second floor. This can undoubtedly
be attributed to the gain of strength of the ground floor caused by the LGS strengthening.
93
Figure 5.60. Deformed shape and tensile cracking pattern for (a) X and (b) Z direction pushover
The comparative pushover curves, from the structure without and with LGS strengthening are presented
in figure 5.59. As it can be observed, the effect of the LGS strengthening is more significant then in the
case presented in 5.58. A measurable improvement of the performance can be observed, both in terms
of strength and ductility, especially in the X loading direction. Based on these results, it can be
appreciated that using LGS strips, in the presented horizontal configuration can bring benefits to the
performance of masonry structures.
5.2.2.5. Coupling of steel frames with existing masonry walls The first strengthening technique examined is the attachment of steel frames at the exterior part of the
walls. These frames were considered fully connected to the structure at each beam to column
connection. At this stage, the connections were not examined in detail. The profiles used for the frames
were HEA 400 and the steel grade was assumed as S275. In the figure 5.61 the application of this
technique on the ABAQUS model is depicted.
Figure 5.61. Scheme of retrofitting technique: coupling of masonry building using steel elements
From the original structure pushover curve (figure 5.62.a) it is clear that the masonry vertical walls
develop their maximum base shear at the area of 5 mm displacement. After the application of the steel
frames, the pushover curve for the masonry and for a single steel frame was created separately as shown
in figure 5.64.a and figure 5.64.b.
It is clearly seen from the above figures that the vertical masonry walls and the steel frame are reaching
their maximum base shear at different top displacements. The steel frame is fully activated after the 20
mm top displacement (figure 5.62.b). At this top displacement, masonry walls have already failed
94
(figure 5.62.a). This strengthening technique seems capable to provide ductility to a structure that is
originally semi-ductile and not as stiff as the masonry benchmark.
The contribution of the steel frames to the stiffness of the retrofitted building is quite poor. The
maximum base shear developed on the retrofitted structure is at the range of 5300 KN and refers to top
displacement of 4 mm. After the maximum shear is reached, the pushover curve drops to base shear at
about 3900 KN for top displacement up to 10 mm (5.62.c). Beyond this value, the curve is intensively
oscillating due to the use of Dynamic Explicit Analysis. These results are neglected.
For the evaluation of the adding steel frames strengthening technique, the Demand-Capacity curve
according to EC8 was created (figure 5.63). With the contribution of the steel frames to the lateral
stiffness of the structure, the retrofitted building is not able to reach the Life Safety Demand curve.
(a)
(b)
(c)
Figure 5.62 Retrofitting technique using coupled steel Moment resisting frames. (a) masonry (b) steel
(c) masonry and steel
Figure 5.63 Demand-Capacity diagram according the EN1998-1-1 spectrum.
5.2.2.6. Coupling of braced frames with existing masonry walls The last retrofitting technique analyzed for the masonry benchmark is the application of vertical bracing
systems fully connected to the masonry walls. The layout of bracing frames is shown in figure 5.64.
Two types of steel S275 profile have been used: HEA 200 for columns and beams, box 80x80x8 for
diagonal elements.
a) b)
Figure 5.64. Application of vertical bracings: a) 3d view; b) lateral view of the bracings.
95
The structural behaviour of the retrofitted structure in the X direction is reported in figure 5.65.a as
ADRS representation of the N2 method assessment, while in figure 5.65.b is shown the intersorey drift
profile for Immediate Occupancy, Life Safety and Collapse Prevention limit state. In figure 5.66 are
illustrated similar curves for the retrofitting in the Y direction. It can be observed that in both cases the
retrofitting solution is very stiff and strong with a very low level of ductility, sufficient enough to
satisfy also the CP assessment. It should be also noticed that the intestory drift profile are rather
different in the two direction: in fact in X dir. there is an high demand at the bottom floor, while in the
Y dir. the request is more graduated.
a) b)
Figure 5.65. X retrofitting solution: a) capacity curve in ADRS format representation, d) collapse
mechanism and ductility assessment.
a) b)
Figure 5.66. Y retrofitting solution: a) capacity curve in ADRS format representation, d) collapse
mechanism and ductility assessment.
5.3. Comparison of analysed retrofitting techniques: structural performance
vs. economic aspects In Figure 5.67 a comparison of the analyzed techniques for the seismic retrofitting of r.c. benchmark
structure is provided in terms of structural performance (i.e. ability of fulfilling expected target
displacement). The analyzed techniques are:
1) Buckling Restrained Braces (BRB);
2) Concentric Braced Frames (CBF);
3) Eccentric Braced Frames (EBF);
4) Light Gauge Shear Walls (LGSW);
5) Steel Shear Walls (SSW).
For each technique, the ADRS representation of N2 method assessment (EN 1998-1) in both X and Y is
reported in the figures 5.67, comparing the equivalent bilinear SDOF curve with the elastic and inelastic
spectra corresponding to Immediate Occupancy, Life Safety and Collapse Prevention. The red spectrum
is related to CP, the black one to LS and the blue one to IO. All the techniques satisfy the IO, LS and
CP requirements, as foreseen by the adopted design strategy, with the only apparent exception of BRB:
in such a model the effects of confinement and the improved ductility on existing column is not
considered and so the premature failure in local sections occurred. In the other models, on the contrary,
it has been clearly considered and this fact has leaded the simulation till the expected target point.
In particular, the following general considerations on the interventions techniques can be extrapolated
from the structural performance assessment carried out and the layout of the technical solutions:
0.000
1.000
2.000
3.000
4.000
5.000
6.000
7.000
8.000
9.000
10.000
0.000 0.050 0.100 0.150 0.200 0.250
acc
ele
rati
on
[m
/s2]
displacement [m]
0.00
1.00
2.00
3.00
0.00% 0.10% 0.20% 0.30% 0.40% 0.50%
flo
or
Drift
IO
LS
CP
floor1
floor2
floor3
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3.000
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acc
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rati
on
[m
/s2]
displacement [m]
0.00
1.00
2.00
3.00
0.00% 0.10% 0.20% 0.30% 0.40% 0.50%
flo
or
Drift
IO
LS
CP
floor1
floor2
floor3
96
the techniques employing the shear walls have been optimized in order to having the lowest
possible level of yielding in order to reduce the demand on the foundations and aiming at a
ductile behaviour;
the techniques employing shear walls due to the presence of few elements (only 4 – 2 along X
and 2 along Y – for the ductile solutions) presents also a low lateral stiffness if compared to the
other solutions more diffusely distributed among the bays of the exterior frames (CBF,
EBF,BRB and LGSW using a resistance upgrading approach – see §5.2.1.4: 8 LGSW have
been used for having a more resistant and stiff structure);
solutions using bracing systems, after many design iterations, presented articulated structural
paths for transferring the inertia forces to the foundations; in particular, CBF solutions along Y
direction and EBF solutions require the insertion of many element in different bays inside the
external structural frames, producing some potential architectural constraints (not considered in
the actual analysis as a design parameter);
EBF solution has been defined adopting inverted V configuration with stub profile between the
r.c. beam and the steel braces, in order to reducing the drilling operations and connections
between the steel elements and the floor; the inclination of the braces is not favourable and a
high amount of steel elements are required for stiffening and strengthening the structure; (more
steel is employed for EBF than for CBF);
BRB configuration has a quite clean layout and require less bracing elements respect to EBF
and CBF, lowering the intrusion level of new elements inside existing structure.
X direction
Y direction
Figure 5.67.a Performance obtained using BRB technique in an optimized application.
0.00
1.00
2.00
3.00
4.00
5.00
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acc
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[m
/s2]
displacement [m]
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ac
ce
lera
tio
n [
m/s
2]
displacement [m]
97
Figure 5.67.b Performance obtained using CB technique – limited ductility / more strength – in an
optimized application.
Figure 5.67.c Performance obtained using EBF technique –ductility / strength – in an optimized
application.
Rottura per flessione trave 0.000
1.000
2.000
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Rottura per pressoflessione colonna
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2]
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98
Figure 5.67.d Performance obtained using LGS technique –ductility / strength – in an optimized
application.
Figure 5.67.e Performance obtained using Shear Wall technique –ductility / strength – in an optimized
application.
5.3.1 Cost analysis of the interventions The safety requirements in the seismic retrofitting intervention techniques are mandatory so, often,
economic requirements or feasibility aspects are those that more condition the final choice between
different techniques. In such part of the report a cost analysis is reported in order to contextualize the
various technique in economic terms, trying at the same time to analyze the cost breakdown between
the different ‘elements’ of an intervention technique.
The analysis here presented considers the following cost sources:
Wall demolition – (m3)
Ground digging – (m3)
Concrete removing – (m3)
0.000
1.000
2.000
3.000
4.000
5.000
6.000
7.000
8.000
9.000
10.000
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acc
ele
rati
on
[m
/s2]
displacement [m]
0.000
1.000
2.000
3.000
4.000
5.000
6.000
7.000
8.000
9.000
10.000
0.000 0.050 0.100 0.150 0.200 0.250
acc
ele
rati
on
[m
/s2]
displacement [m]
6
1st floor
2nd floor
3rd floor
5
53A
B3C
D1E 0.000
1.000
2.000
3.000
4.000
5.000
6.000
7.000
8.000
9.000
10.000
0.000 0.050 0.100 0.150 0.200 0.250
acc
ele
rati
on
[m
/s2]
displacement [m]
6
1st floor
2nd floor
3rd floor
5
53A
B3C
D1E 0.000
1.000
2.000
3.000
4.000
5.000
6.000
7.000
8.000
9.000
10.000
0.000 0.050 0.100 0.150 0.200 0.250
acc
ele
rati
on
[m
/s2]
displacement [m]
99
Steel new system – (kg)
New concrete – (m3)
Core drillings – no. of holes
N° employed micro-piles
All the additional works for rebuilding the infilling walls and to rebuild the ground floor of the building
after the retrofitting intervention on foundations have been not considered because often they are
correlated to other architectural parameters and final details that are not structural related. The unitary
costs utilized in this analysis are reported in the table 5.9.
Table 5.9 Costs for each voice obtained from the Italian prices of Commerce Chambers.
The cost of the steel elements considers the base material/products supply, the working of the
material/products according to the design specifications, the delivery of finished element to the
construction site (about 200 km maximum distance) and the installation of the elements in the existing
structure. The cost for the realization of the local reinforcement is higher because its realization is made
on site using pre-heating and welding approach of the elements.
The cost of the micro-piles consider the following contributions: drilling phase for holes with 200 mm
of diameter (maximum); supplying of steel parts and reinforcement; installations of the elements;
completion of the micro-pile with the concrete grouting. The cost for removing the existing concrete
considers: demolition of concrete; cutting and removing of the steel reinforcement; loading and
transport of demolished parts. The cost of the ground digging has been considered adopting a mixed
approach: the 50% of the ground can be removed using machine from the external side when the other
50% of the ground can be removed working inside the building and using only workmanship and no
high capacity machines. In particular, looking at the commerce chamber prices for such type of work it
has been obtained: 10€/mc for digging from the exterior and 170€/mc for digging from the interior.
The total cost of the interventions are reported in the table 5.10 and table 5.11 (total costs and relative
incidence on the total); looking at these values the following considerations can be argued: in all the
interventions the foundation cost represents about the 50% of the total; after the foundations, the most
relevant costs voices are the construction of the new steel systems, the local strengthening of the
existing elements and the demolition of infill walls (here reported according to their decreasing
relevance in the total cost estimation). These first four cost sources represent the more valuable
economic indicators for the examples here considered, and their estimation acquires according to this
perspective a relevant role in the designing of each seismic retrofitting intervention.
Table 5.10 Total cost and cost breakdown for all the optimized solutions
Wall demolition 270 €/mc
Ground digging 90 €/mc
Concrete removing 400 €/mc
Steel new system - main
structural elements3,5 €/kg
Steel elements for the local
reinforcement of beam and
columns
5,5 €/kg
New concrete 140 €/mc
Core drillings 14 €/each
N° micropiles 1400 each
SolutionTotal
cost
Wall
demolition
Ground
digging
Concrete
removing
Steel new
system
New
concreteCore drillings Micropiles
Local
strengthening
No. Reinforced
columns
Cost of
intervetion
€ € € € € € € € €/mq
Ductile SSW 356727 21060 7560 5600 113470 3780 1186 167872 39000 8 431
LGS SW -
strength234821 21060 6120 1600 76825 3080 3030 123106 284
LGS SW -
ductility243262 11340 6080 12160 28000 2100 1976 123106 58500 12 294
CB System 263888 36383 7200 2000 51279 4200 128702 34125 7 319
EBF System 311037 35978 7200 2000 98833 4200 128702 34125 7 376
BRB 262522 22680 6480 2000 54620,3 3444 134297 39000 8 317
Average 278710 24750 6773 4227 70504 3467 2064 134297 40950 8 337
100
Table 5.11 Relative influence of each single voice on the total
The high cost of the foundation upgrading is in general expected during the design practice, but in this
case its incidence is so relevant because the ground adopted for the design of the intervention technique
using micro-piles (§7) has been a class C soil with relevant bearing problems. Obviously, a better
ground quality could mitigate this effect.
Another aspect to be considered according to the perspective of the economic convenience is the
ductility: the solutions designed for exploiting relevant ductility properties of the steel system
automatically call into the working scheme also existing elements, requiring so a relevant economic
contribution for their local strengthening (r.c. columns and beams).
The geometrical configuration of the steel elements in the bracing schemes has also a relevant impact in
the steel consumption: the EBF scheme could be an economic effective solutions (e.g. low impact on
foundations), but the scheme of braces require big sections for satisfying stiffness requirements
increasing the incidence of steel cost to level equal to shear wall systems: the inclination of inverted V
scheme does not allow braces working properly in the stiffening effect
The most convenient intervention technique considering the total cost is the shear wall technique that
use light gauge steel products while the more expensive technique is the shear wall using structural
plates: in particular, the strong difference between the two technique is in the foundation costs, imposed
by the demand at the foundation system for transferring all the upper structure reactions to the soil.
Moreover, the ductile SSW has been developed using an articulated steel frame surrounding the SSW in
order to transfer load mainly through the floor slab. This solution produced a very high amount of steel
consumption that was reflected in the total cost of the solution.
All bracing schemes arrive to comparable total costs but it is interesting to note that the cost for CBF
solution and BRB system are similar while the cost of EBF is higher; this relative differences can be
individuated, mainly, in the performance of the steel bracing elements. In fact, an inverted V scheme
with low dissipative capacities require much more material than a similar geometrical scheme endowed
with a clear ductile behaviour; on the other hand, a more pronounced ductile behaviour of the retrofitted
structures will necessarily require a relevant upgrading of existing members and their relative
foundations.
(a)
(b)
Figure 5.68 Total cost of the intervention for sm of useful floor area and costs of the four selected
economic parameters.
All simulations carried out considered foundations and horizontal elements (i.e. floors) as already
retrofitted; in particular, this has been diffusely treated for the foundations in the cost analyses being a
relevant parameter in the judgement of the retrofitting scheme. On the contrary, the flooring systems
SolutionTotal
cost
Wall
demolition
Ground
digging
Concrete
removing
Steel new
system
New
concreteCore drillings Micropiles
Local
strengthening
€ € € € € € € € €
Ductile SSW 5,90% 2,12% 1,57% 31,81% 1,06% 0,33% 47,06% 10,93%
LGS SW 8,97% 2,61% 0,68% 32,72% 1,31% 1,29% 52,43% 0,00%
LGS SW 4,66% 2,50% 5,00% 11,51% 0,86% 0,81% 50,61% 24,05%
CB System 13,79% 2,73% 0,76% 19,43% 1,59% 0,00% 48,77% 12,93%
EBF System 11,57% 2,31% 0,64% 31,78% 1,35% 0,00% 41,38% 10,97%
BRB 8,64% 2,47% 0,76% 20,81% 1,31% 0,00% 51,16% 14,86%
Average 8,88% 2,43% 1,52% 25,30% 1,24% 0,74% 48,19% 14,69%
101
have not been considered directly given that their cost, if necessary, it has to be summed to all
techniques as a fixed cost. Anyway, an estimation of this on the global cost of the retrofitting has been
executed. In particular, according to the results presented in §6, the solution that furnished the best
result in terms of stiffening has been the steel bracings system and it has been assumed to adopt this
solution to in-plane stiffening and strengthening the floor (i.e. diaphragmatic action).
The type of intervention for the floor assumed in order to obtain an economic estimation has been
characterized by the following data: (1) in each floor field two 16mm bracing elements are placed; (2)
the connection system between the bracing and the existing parts are realized using steel plates; (3) the
connection system is localized at the corners of each floor field, using 500×100×10mm steel plates; (4)
the connection between the new and the old structure is realized using mechanical fastening with bolts.
The total cost of the interventions for the floor stiffening has been estimated about 11500€, about
14€/m2 of the total floor area of the building; in particular, 7500€ is the cost of the steel elements
(braces and connections) to be installed under the floor while 4000€ is the realization of the holes for
connecting the elements with the existing parts. Compared to the cost of the global retrofitting solutions
floor intervention incidence is between 3 and 5% maximum, and it can be considered a parameter that
can be considered on a second step after the analysis on the previous four more relevant cost sources:
steel consumption; foundations; walls demolition; local strengthening of existing members.
(a) (b)
(c) (d)
(e) (f)
Figure 5.69 Influence of each voice on the total costs of the intervention techniques.
102
5.3.2 Practical implications and guidelines
Structural and economic analyses
The relevant economic implications of steel consumptions, demolition of existing parts and the
upgrading of the foundations suggest that a correct evaluation of the design technique necessarily
require the design of interventions in all structural parts: upper structure and foundations. Partial
analysis of the upper structure only could give some preliminary indications about costs, but as
presented in the previous economic analysis, it is the foundation system that strongly influence the total
cost, modifying the preliminary estimations.
Excavation works and drilling works for realizing the connection between new system and old structure
do not represent in such analysis a relevant part of the cost, suggesting as economic parameters of the
retrofitting design the following four sources: steel consumption (i.e. total cost = material supplying,
material working, delivery on site and installation); demolition of existing structural parts for installing
new elements; local strengthening of existing elements in the upper structure; works for upgrading the
foundation systems.
Another aspect to be carefully considered is the balance between strength and ductility; it has been
shown comparing the intervention techniques with LGSW that adopting an approach mostly devoted to
the strength improvement rather than ductility improvement can be a valuable solutions. In the
presented case, anyway, it is worth recalling that the accuracy of the models and the expertise of the
designers produced two solutions characterized by a similar economic impact and compared in the
previous cost analysis; moreover it is worth underlying also that increasing the strength level of the
retrofitted structure help the protection of existing parts reducing so intervention limiting as much as
possible the local retrofitting of the elements. On the other hand, a too severe internal loading level
would mean an high level of forces to be transferred to the foundations if not properly taken into
account during the design.
It is so clear from the previous analyses and from these last considerations that a good starting point in
the design strategy is related to the choice of a solution able to develop its main beneficial effects at
displacement level compatible with the existing structure in order to limit as much as possible the
intervention on existing elements.
On the other hand the exploitation of a certain amount of ductility could have for sure a positive impact
in the reduction of internal forces and on the forces to be transferred to the foundation; for such reason
flexible techniques as those examined are strongly suggested due to their capacity of regulating strength
levels and the ductility exploitation (from the existing structure side). In particular, bracing systems and
shear walls using light gauge (and weak) walls seem to be more appropriate.
Technical aspects
The adoption of retrofitting techniques only in the exterior frames it is herein suggested as a convenient
approach for the intervention allowing to minimize the works inside the structure and to guarantee a
certain level of reversibility of the intervention; moreover, placing the new elements, when possible, in
the exterior frames guarantee a higher torsional stiffness in the retrofitted construction. The realization
of connecting system with the existing structure will require the execution of the holes and the
realization of the steel details for the mechanical connection. Also if the costs for realizing the holes, the
steel details and the installations have been found as not relevant, the connecting points between the
new installation and the old structure should be limited in order to reduce the amount of work but at the
same time the extension of this connecting zones should be enough ‘large’ to reduce local strength
demand on the existing material. In particular, pre-tensioned mechanical connecting systems that do not
require the drilling in existing main structural members (tested in §8) could work in this sense.
Figure 5.70 Connection technique between braces and existing elements using pre-tensioned elements
and limiting the holes drilling inside main structural elements.
103
6. Performance analysis of steel solutions for horizontal elements
6.1 Masonry benchmark structure For the analysis of the horizontal elements of the Masonry Benchmark building a detailed model was
created in SAP2000 v.10 In general, shell elements were used to simulate the vertical elements of the
structure as well as the floor and roof covers. Moreover, beam elements were used to simulate the floor
and the roof support system, figure 6.1.
(a)
(b)
Figure 6.1. (a) 3D model of the masonry benchmark; (b) model of the floor system
As presented in the general drawings, three floor types were used at all the levels of the structure. Floor
type 1 is a masonry based floor type which covers the whole ground floor and most of the elevation of
the first floor. Floor type 2 consists of a timber beam supporting system and a masonry tile cover. It is
mainly used in the second floor level. Finally, floor type 3 consists of steel beams that support a block
cover. This floor type is used only in a small part of the second floor level and it is probably a result of
a prior strengthening intervention.
The roof consists of a main timber beam system that supports all the secondary beams and the tile
cover. All timber parts and their exact geometry in space were modeled in detail.
6.1.1. Intervention Techniques Taking into consideration the strengthening techniques presented in the WP2 for floors and roofs and
the structure loading calculated according to the provisions of EC 1 and EC 8, as described in the
corresponding report of WP4-5, the following intervention techniques were examined.
6.1.1.1. Floor systems Replacing the existing timber floor system with Reinforced Concrete slab: By replacing the
existing timber floor system with reinforce concrete slab, the earthquake performance of the
structure is highly improved due to the diaphragmatic action the concrete slab introduces at each
floor level. The corresponding horizontal deflections of the surrounding walls are significantly
decreased, resulting to an also decreased development of stresses up to 25%. The vertical load
bearing capacity is also increased. For the application of this technique a concrete slab of thickness
t=15 cm was created. The concrete grade was assumed to be C 25/30 and the reinforcement steel
grade was assumed as S400.
Adding horizontal steel bracing systems: An alternative way to improve the diaphragmatic
behavior of the floor system is to insert a horizontal steel bracing system under the existing timber
floor system. This technique does not increase the vertical load bearing capacity of the existing
floor. For the application of this technique, steel bars of S400 steel grade and diameter 12mm and
105
16mm were used.
Replacing degradated parts with new steel parts: In order to apply this technique it was assumed
that the main supporting beams at each floor part were degradated and had to be replaced. The
overlapping timber plates on the beams were kept in place. The profiles used were IPE 140 and IPE
160 of steel grade S275 and cold-formed C210-30 of steel grade 350G.
Adding trussed perimeter beam: A trussed perimeter beam is inserted under the existing timber
floor system. This is an alternative to the steel or concrete ring beam. It is commonly used in order
to improve the diaphragmatic behavior of the floor system. This technique is not expected to
increase the vertical load bearing capacity of the existing floor. For the application of this technique,
the trussed beam was formed with TUBO 60X60X5.4 cross-sections. The width of the trussed beam
is 1.46 m. The steel grade was assumed as S235.
6.1.1.2. Roof systems The ring beam technique: A RC/steel ring beam constructed at the roof level is one of the most
effective measures to prevent the out-of-plane collapse of masonry walls. Dislocation of the roof
structure is prevented by anchoring its elements into the ring beam. The ring beam contributes to the
reduction of the out-of-plane stresses on the upper part of the roof supporting wall but it does not
retrofit the roof to withstand additional vertical loading.
Adding steel bracing system: In order to improve the roof’s bearing capacity against vertical
loading, a steel truss has been inserted underneath the existing roof. Despite the fact that the main
purpose of this intervention is to upgrade the bearing capacity of the roof under vertical loading, a
significant horizontal deformation reduction is also observed. The cross-section used for the truss is
TUBO 100x100x10 of S275 steel grade.
Replacing degradated parts with new steel parts: One of the most traditional methods for
repairing roofs is the replacement of the degradated parts with new timber parts or steel profiles. If
steel profiles are selected, the roof’s bearing capacity and efficiency against vertical loading is
improved. The steel profile adopted for the examination of this technique in the present study was
IPE 220 of steel grade S275.
6.1.2. Analysis results For the evaluation of the performance of each technique adopted, several check points were selected on
the second floor and the roof level, as illustrated in the following pictures. At these points the horizontal
and vertical displacements were monitored and compared with the reciprocal displacements of the
original structure.
Figure 6.2. Check Point at Floor – Roof
The performance of each retrofitting technique regarding the reduction of the horizontal and vertical
deflections is depicted in the following graphs. Concerning the vertical deflections at the middle span of
the floor, it is shown that the use of steel members can reduce the developed deflection in an effective
manner (Figure 6.3). Depending on the steel profile adopted, the deflection reduction can reach up to
25%.
The effectiveness of the adopted retrofitting techniques as far as the horizontal deformation reduction is
concerned is also presented in the following diagram (figure 6.4.a). If the criterion of 10% difference
from the infinite-diaphragmatic action limit is applied, then from the following chart it is observed that
only the use of 16 mm steel braces can provide adequate diaphragmatic action. The ring beam
techniques (steel/concrete, trussed) do not provide adequate diaphragmatic action; nevertheless they
106
decrease the developed stresses on the walls. A similar graph referring to a single wall is also depicted.
This graph refers to the wall between joints 1 and 2 (figure 6.4.b). The intermediate points are placed
every L/3, where L is the distance of joints 1 and 2.
Figure 6.3. Deflection reduction of Floor, (a), and Roof (b) systems . comparison
(a)
(b)
Figure 6.4. Horizontal displacement reduction – Floor systems
6.1.3. Connection design for floor and roof systems In this paragraph all the connection types used for each strengthening technique are discussed in detail.
The dimensioning was based on the acting forces and moments resulting from the model analysis.
6.1.3.1. Replacing the existing timber floor system with Reinforced Concrete
slab In order to connect the new RC slab to the existing wall system, steel anchors of diameter 12mm were
used. The anchor length was protruding from the wall side 60cm into the RC slab to provide sufficient
anchorage. Along the wall, anchors were placed every 1.5m in order to distribute the tensile forces
resulting from the wall-RC slab interaction. At the exterior part of the wall, the anchors were bolted
over steel plates of nominal dimension 100x100x10 mm (figure 6.5.a).
6.1.3.2. Adding horizontal steel bracing systems For the connection of the horizontal steel bracing systems to the wall corner, two connection types have
been examined. Both consist of angle-formed steel plates with nominal dimension 900x500x10 mm and
450x500x10 mm placed on the exterior and interior face of the wall correspondingly. These angle-
plates are well connected with steel anchors of nominal diameter 12 mm spaced every 100 mm and
passed through drilled holes along the wall height (figure 6.5.b)
EQ1
-2.5
-2
-1.5
-1
-0.5
0
1 2 3 4Joints
Ho
rizo
nta
l D
isp
lacem
en
t [m
m]
RC slab (diaphragmatic)
brace 16 mm
concrete ring beam
trussed perimeter beam
unreinforced
107
6.1.3.3. Replacing degradated parts with new steel parts The new steel beams that replaced the existing masonry had to be inserted to the wall for at least 15 cm
in order to support the overlaying timber floor system. One steel flange of nominal dimension
100x80x10 mm was bolted to the web flange of the IPE section with two M 12 8.8 bolts. Then the steel
flange was formed and connected to a steel anchor of nominal dimension 12 mm. At the exterior part of
the wall the anchor was bolted over a steel plate of nominal dimension 100x100x10 mm. (figure 6.5.c)
6.1.3.4. Adding trussed perimeter beam In order to connect the trussed perimeter beam to the existing wall system, steel anchors of diameter
12mm were used. The anchors had to be inserted to the wall for at least 10 cm in order to provide
sufficient anchorage. Along the wall, anchors were placed at every truss joint. (figure 6.5.d)
6.1.3.5. The ring beam technique After the formation of the ring beam at the top of the supporting walls, a connection between them had
to be established. Steel dowels of nominal dimension 12 mm every 50 cm were used to connect the ring
beam with the supporting walls. The roof was connected to the ring beam with a combination of
anchors and dowels. (figure 6.5.e)
6.1.3.6. Adding steel bracing system The connection consists of a steel plate of nominal dimension 400x300x20 mm welded at the end of the
main truss. The plate is drilled at six locations where M20 8.8 steel anchors are inserted. The anchor’s
overall dimensions are 150x50 mm. Moreover a dowel of length 100 mm and nominal dimension 14
mm was used. (figure 6.5.f)
In order to verify that the support will not fail, a local strengthened area has to be formatted. The
common practice is to form a cavity on the wall and replace the masonry elements with a reinforced
concrete block. The concrete quality was considered C25/30 for calculations. (figure 6.5.f)
6.1.3.7. Replacing degradated parts with new steel parts The support of the new steel profiles that replaced the main timber beams on the roof is formed as
shown in figure 6.5.g. The connection consists of a steel plate of nominal dimension 350x300x20 mm
welded at the end of the main beam IPE 220. The plate is drilled at four locations where M20 8.8 steel
anchors are inserted. The anchor’s overall dimensions are 400x50 mm. Moreover a dowel of length 100
mm and nominal dimension 20 mm was used. (figure 6.5.g)
6.2. Retrofitting or upgrading of floors/roofs for r.c. buildings
6.2.1. Floor systems in existing r.c. buildings In existing r.c. buildings floor systems are commonly made of in-situ or prefabricated reinforced
concrete or floors with precast reinforced concrete joist and lateritious and reinforced concrete slab.
Over the years a wide variety of floor systems have been developed. Some examples of floor systems
usually present in existing buildings can be found in Table 1.
In ordinary buildings rib and pan floor systems have been usually adopted due to reduced weight in
comparison with flat r.c. slab and relatively quick erection time. A large percentage of the current
constructed one-way slabs are partial prefabricated floors, where the prefabricated lower surface
includes the whole tensile reinforcement and substitutes the formwork.
Under operating loads floors are mainly subjected to vertical loads. The diaphragm action of the floor is
exploited under wind loads, where the floor connects the vertical members. Seismic loads normally do
not govern the design of floors. The primary purpose of floors as diaphragms in the overall seismic
system is to act as a horizontal beam spanning between lateral force-resisting elements. The stiffness
and capacity of the floor must provide a sufficient transfer of the seismic load to the vertical bracing
elements.
108
Figure 6.5. Details of connecting systems for application of intervention techniques
Inside a frame structure, floor systems have two main structural functions: the “out-of-plane” and “in-
plane”. The primary function of floor and roof systems is to support gravity loads and to transfer these
loads to other structural members such as columns and walls (“out-of-plane” behaviour), whereas under
earthquake loadings, floor systems play a central role in the distribution of seismic forces to the vertical
elements of the lateral load resisting system, such as frames and structural walls (in-plane behaviour).
Concerning the “out-of-plane” behaviour, floor systems in existing r.c. buildings are often modelled
according different strategies depending on floor structure:
- continuous beams (precast r.c. concrete floor joists + r.c. slab; lateritious reinforced floor joists
+ r.c. slab);
- supported beams (precast r.c. concrete floor joists + r.c. slab; lateritious reinforced floor joists
r.c. slab);
- Plates (solid flat slab).
Regarding the “in-plane behaviour”, floor systems are often modelled as rigid diagrams even that they
do not satisfy the code requirements about minimum thickness and reinforcement. Otherwise they can
be modelled as flexible diaphragm:
- series of composite beams (floor joists + concrete slab);
- equivalent shell elements (isotropic or orthotropic).
In any case it should be noted that in-plane floor flexibility can play an important structural role only in
particular stiff r.c. structures such as wall-system frames, where the floor displacements due to in-plane
floor flexibility is of the same magnitude of floor displacements due to vertical load-bearing system
flexibility (Barron and Hueste, 2004).
Deficiencies affecting the primary purpose of floors are typically inadequate shear or bending strength,
stiffness, or inadequate reinforcement around openings or re-entrant corners. Insufficient local shear
109
transfer to lateral force-resisting elements or missing and inadequate collectors are categorized as load
path deficiencies.
In the seismic field the deficiencies of monolithic concrete diaphragms are closely correlated with the
type of floor.
In the case of reinforced concrete or post-tensioned concrete diaphragms principal deficiencies are:
- inadequate in-plane shear capacity of the concrete diaphragm;
- inadequate diaphragm chord capacity;
- excessive shear stresses at the diaphragm openings or plan irregularities.
In the case of precast or post-tensioned concrete planks, tees, or cored slabs it’s possible to add:
- Inadequate in-plane shear capacity of the connections between the adjacent units.
Therefore the main objective to retrofit and upgrade floor systems is to establish a sufficient diaphragm-
action and therefore the increase of strength and stiffness. This includes the transfer of forces from the
floors to lateral force-resisting elements as well.
Floor system made of cast-on-site joist,
usual spacing 50 cm (depending on the pan
dimension);
Original configuration: without distributing
slab
Bottom reinforcement in the ribs
Floor system made of prefabricated on site
lateritious pan joist, spacing 20 cm;
Original configuration: without distributing
slab
Bottom reinforcement: 16 bars for meter (
5-6) + additional bars in the ribs
Floor system realized by prefabricated
joists, spacing 25 cm
Only 2 bottom rebars (increased diameter)
Upper concrete slab (3-4 cm).
Lateritious prefabricated joists with
incorporated rebars, heigth 16 - 20 cm,
spacing 50 cm; + Lateritious pan
Additional rebars inserted in the concrete
ribs
Upper concrete slab 3-4 cm (not always
present)
Prefabricated r.c. joists, variable spacing 50
- 100 cm;
Thin hollow masonry tiles
Higher load bearing capacity
wire mesh not always present
110
Partially prefabricated joists with trussed
rebar systems, easier casting
Light and easy to mount
Several typologies
Collaborating slab, light pan
Different hollow tile system
Possibility to reduce the self-weight
introducing polystyrene blocks
Prestressed precast joists
Floor joists with higher rebars number
Different lightening blocks
Upper r.c slab
Partially prefabricated panel floor
Additional rebars in the rib
Upper r.c. slab
Reinforced concrete slab
Mushroom slab
6.2.2. Retrofitting techniques for floor systems in existing r.c. frames Strengthening techniques for floor systems can be grouped in:
techniques to strength the floor directly;
techniques to strength downstand elements of the frames;
techniques to add supplemental vertical-resisting elements (shear walls or braced frames);
In this context only the techniques belonging to the first group are analyzed.
6.2.2.1.Post-tensioning of floors Post-tensioning is an excellent method to increase the capacity of many different reinforced concrete
elements. The main objective is to increase the bending and shear capacity by axial forces. For post-
tensioning of floors straight tendons are used in two layers at the lower and upper side of the floor (see
Figure 1). Further applications of tendons are the connection of new vertical bracing systems e.g. shear
walls, staircases and lift shafts with the existing structure. External post-tensioning can also be used as
ties along floor edges to reach a sufficient diaphragm action.
111
The design of retrofit solutions by post-tensioning includes the choice of tendons type (strands, wires
and bars), their arrangement and the introduction of the post-tensioning forces into the existing
structure. Bars are preferred for short tendons (5 – 10 m) or to simplify the erection (e.g. connection
with coupler). Short strands have relative high loss of clamping forces due to the slip in the anchorage
and wire with headed ends needs an exact cutting to length. The advantage of the higher steel strength
of strands and wires in case of creep and shrinkage are of minor importance for existing buildings.
Internal post-tensioning is normally not applicable due to the limited depth. Hence, external post-
tensioning is used in two layers at lower and upper side of the floor to avoid eccentricities, while
ducting are necessary to prevent second order effects in case of deflections. Corrosion protection is
obtained by PE-coating and grease. The anchoring and load introduction of post-tension forces can be
provided by steel trusses. The existing structure has to be verified or strengthened for local and global
lateral tension forces caused by the post-tensioning.
Figure 6.6. External post tensioning.
Figure6.7. Additional steel bracings.
6.2.2.2. Steel bracing Providing a horizontal braced frame as a diaphragm strengthening technique is useful if concrete
overlays add too much mass or lead to other construction complications. The new horizontal bracing is
added under the existing diaphragm, in which the existing framing with new diagonal members forms
the horizontal bracing system. The diaphragm shears are shared with the existing diaphragm in
proportion to the relative rigidity of the two systems (see Figure 6.8).
This method requires the accessibility of the lower side of the floor and may necessitate reinstalling of
pipes and ventilation ducts. The design of the bracing system should consider the logistics associated
with delivering and attaching the braces to their final locations. Further fire and corrosion protection are
necessary.
6.2.2.3. Steel collectors Addition of a new collector or strengthening of an existing collector is often needed when new steel
braced frames or concrete shear walls are added to an existing building. The new collector must extend
as far as necessary, often one or more bays from one or both ends of the new brace or wall, to draw the
required shear demand from the existing diaphragm. The new collector will be constructed of reinforced
concrete or steel, generally depending on whether the general building upgrade involves installation of
new concrete shear walls or steel braced frames. The new collector will most often be installed at the
underside of floor. At roofs, the collector may be placed either from below or above the roof.
In reinforced concrete buildings with some sort of concrete slab floor system, especially one with joists,
waffle ribs or beams crossing the path of the collector, the most common material choice for the new
collector is reinforced concrete. Often, this choice is made because concrete is aesthetically compatible
with the surrounding structure, especially in a condition exposed to view. Otherwise steel plate or
profile can be added to act as collector. At a steel plate collector, the elongation of the plate is not
compatible with the diaphragm slab. As the collector load accumulates towards the connection to the
new wall or brace, the elongation of the plate accumulates as well. The threaded rod anchors connecting
the plate collector to the diaphragm in the zone of greatest elongation can become overloaded to failure
by the plate bearing on the bolts.
A new collector often must extend one or more entire bays away from the new wall or brace in order to
draw the necessary load from the existing concrete diaphragm. Installation of the new collector at the
112
underside of the existing floor slab impacts any existing ceilings, partitions, ductwork, plumbing,
lighting, etc., located along its entire length. As a result, the new collectors will often have a greater
impact on the building’s other systems than the new walls or braces themselves. Furthermore,
consideration of these impacts will often affect placement of the new walls or braces. In many cases, the
new walls and their associated collectors are located along the exterior edge of the building specifically
to avoid or minimize these impacts on other building systems, especially in a case where building
occupancy is maintained during the construction.
Collectors have significant cost/disruption impact in a retrofit project primarily due to their length.
Thus, any available means of reducing collector length will probably be cost effective. A collector
installed at the exterior edge of a diaphragm will generally be less costly than one installed in the
interior and one installed above the diaphragm will be easier to install and, generally, less costly than
one installed from below. However, installation of any collector can be very disruptive to any building
occupants, due to the noise and vibration caused by drilling and coring through concrete, as well as the
likely need to relocate various utilities and service distribution systems.
A steel collector will have to be installed in manageable sections, generally about 10 to 20 feet in
length, and will be connected to the concrete diaphragm with drilled threaded rod anchors set in
adhesive or epoxy. In almost all cases, the steel plates will be installed at the top of the diaphragm as
shown in Figure 3. Although possible, it is extremely difficult to install heavy plate sections, connect
the bolts and make the necessary welded splices from below.
Figure 6.8. Steel plate collectors.
As discussed above, the primary concern with a steel plate collector is its lack of strain compatibility
with the concrete diaphragm, unless the collector is very short. The strain deformation of a steel
collector will vary from zero at its free end to a maximum at the connection to the wall or brace while
the concrete diaphragm will not experience similar deformations. In effect, the steel collector will
stretch like a very stiff rubber band relative to the concrete diaphragm. This relative deformation is
difficult to accommodate, especially in relatively long collectors. To do that, several conditions must be
considered. First, the various plate sections of the collector must be stepped in size so the strain is
distributed relatively equally along the length of the collector. Second the plates must be sized to limit
the maximum elongation to a reasonable amount of about one or two inches. Third, the threaded rod
anchors must be installed in slotted holes to allow the design elongation to occur without bearing on and
overloading the anchors. Fourth, to allow the slip to occur between the collector and diaphragm, load
transfer must be accomplished by friction using specially calibrated spring washers to generate the
appropriate clamping force in the anchors.
113
7. Retrofitting technique for foundation system Foundation rehabilitation schemes were evaluated in conjunction with retrofit solutions studied for
vertical resisting systems. Since the reinforced concrete building case study has been tested with all the
proposed retrofit techniques, comparative analysis have been performed on this case study in order to
make comparisons between retrofit solutions for foundation associated with different retrofit techniques
for vertical systems.
Retrofit of foundation is an essential step to assure that the complete rehabilitation achieves the selected
building performance level for the selected earthquake hazard level. In rehabilitation of foundation, new
rehabilitation elements are often used in conjunction with existing elements. The compatibility of new
and existing components and/or elements shall be checked at displacements consistent with the
performance level chosen.
The effects of rehabilitation on stiffness, strength, and deformability shall be taken into account in an
analytical model of the rehabilitated structure. Moreover, if the foundation system is poor or the retrofit
system requires expensive foundation rehabilitation, in many cases cost of interventions on foundation
can condition the judge on the appropriateness of the overall retrofit solution.
Steel solutions can be adopted to increase the stiffness of foundation or to transfer the loads to more
resistant layers of soil, through deep foundation elements. Micro-piles are used in foundation
rehabilitation and seismic retrofitting projects to enhance the foundation ultimate capacity and reduce
foundation deflection. This part of the project focuses on the effectiveness of using single micropile and
micropile groups in conjunction with different types of steel retrofit systems for the in-elevation
building. Two soil (Type B and C) are used to represent a common range of soil behaviors. Parametric
studies were performed for various independent variables including soil non-linearity, pile
configuration, and retrofit system. The FE element models were used to obtain prescriptive indications
to use in design practice.
7.1. Analysis of micro-piles for foundation retrofitting Micropiles are grouted and small diameter piles that are traditionally used in foundation retrofit.
Experimental evidence and a number of studies have indicated that micropiles behave well under
seismic loading and they can be conveniently adopted for retrofitting existing buildings. Micropiles
solutions were considered in order to increase the performance of foundation systems and allow the
complete retrofit solution to achieve the required performance levels.
However, their effectiveness is significantly different for different structural scheme of the “in-elevation
building”, because it depends on the way in which the structure transfers seismic loads to the
foundation. Several observations on micropile behavior were gleamed from the parametric study. The
use of interface elements that capture soil-pile friction and separation (gapping) is important to capture
adequately soil-structure interaction.
Gapping results in an increase in pile deflection. For a linear elastic soil, the increase in deflection due
to gapping is linearly related to the applied horizontal load. This implies that the gapping elements do
not introduce non-linearity in the pile-soil systems. The increase in deflection when gapping elements
are used compared to deflections in a system with perfect bonding between soil and pile is significant.
Most of the deformation occurs near the top of the micropile. Hence, it is important to incorporate
interface elements between the micropile and the soil at least within six diameter lengths from the
micropile head. Gapping also causes higher moments near the micropile head because a lesser amount
of load will be transferred to the neighboring soils. This, in turn, is due to the lower contact area
between the pile and the soil.
An increase in soil’s non-linearity causes an increase in deflection. Even though this conclusion is self-
evident, it points to the importance of using appropriate nonlinear models of soil behavior. The
mobilized pile moments in piles on inelastic soils are higher than those inserted in elastic soil. This
occurs because of the lesser degree of load transfer from the pile to the soil in the more non-linear
material.
The non-linear behavior of the soil has a significant influence on the response of the micropile to
seismic excitation. A soil classified as type C by EN1998:1 (average shear wave velocity in the upper
30 meters Vs30=300m/s) has been considered. Mechanical characteristics of the soil are derived from a
geotechnical investigation. In figure 7.1, the stratigraphic profile with the average velocity of primary
and secondary waves (Vp and Vs) is represented.
Mechanical parameters representing soil behavior are reported in the Table 7.1, where z is the depth, Vs
115
is the shear wave velocity, NSPT is the Standard Penetration Test Number, qb is the bond strength, G0 is
the elastic shear modulus, qc is the Continuous Penetration Test Strength, cu is the undrained strength,
ult is the shear strength.
z (m) Vs (m/s) NSPT qb (Mpa) G0 (Mpa) qc (Mpa) cu ult
0-5 229 7 0,07 100 4,81 0,17 0,50
5-10 314 11 0,07 187 6,10 0,21 1,89
10-15 268 28 0,12 136 5,42 0,18 1,64
15-23 309 32 0,12 181 6,03 0,20 1,80
23-30 383 49 0,15 279 7,08 0,23 2,09
Table 7.1. Mechanical parameters of Type C soil.
Both vertical and inclined piles were supposed to be used: inclination of micropiles provides larger
lateral stiffness and results in smaller displacements and accelerations at the micropile head as
compared to groups of vertical micropiles. Furthermore, inclination does not affect the strain levels in
the soil, implying that no additional stresses are being transmitted to the soil, and it decreases the
bending moment at the micropile head. This is due to the fact that the axial capacity of inclined
micropiles is also mobilized (in addition to their bending capacity).
Figure 7.1. Stratigraphic profile of Type C soil. Figure 7.2. FE model of micropiles.
The finite element method has been used as the basic framework for the analysis of the seismic
behavior of micropiles (see figure 7.2). A bounding surface plasticity model was used to represent the
nonlinear behavior of soils. The model accurately represents modulus reduction and the increase of
damping with increasing shear strain. Boundary conditions were represented by transmitting
boundaries. The finite element model was validated for various conditions including: pure site response
(e.g. the response of a soil column without the presence of piles), the response of single piles under
lateral load, and the response of micropile groups under static loading.
7.2. Soil-structure interaction assessment Construction of new braced frames, bracing systems and shear walls within an existing structure were
demonstrated to be effective measures for adding stiffness and strength to existing buildings. Shear
walls and braces are effective elements for increases in strength, but they may be significantly stiff and
they can induce high localized foundation loads. Micropiles can transfer these forces to more stiff and
strength deep soil layers. Several preliminary analysis have been carried out. Two types (a and b) of
micropiles are considered, whose characteristics are reported in Table 7.2.
By using the selected micropiles, rehabilitation of foundation of the reinforced concrete building case
study was performed. Four different retrofit solutions for the vertical resisting system has been
116
considered:
1. concentric braces;
2. eccentric braces;
3. Light Gauge Steel (LGS) shear walls
4. ductile shear walls.
Several different configurations of micropiles were considered. Each configuration was designed to
sustain a different level of forces, transferred at the base by the in-elevation building. The design forces
have been obtained by considering different sets of forces derived by the structural analysis of the
retrofitted buildings. All the reactions at the base of the retrofitted buildings were subdivided in 7 sets,
for each one of those a configuration of micropiles have been designed.
PROPERTIES Type a Type b
D=external diameter 140 mm (5 1/2 in) 127 mm (5 in)
t=thickness 12,66 mm (1/2 in) 6,33 mm (1/4 in)
A=cross sectional area 37,6 cm2 (5,83 in2) 23,9 cm2 (3,70 in2)
I=moment of inertia 803 cm4 (19,30 in4) 436 cm4 (10,47 in4)
d=internal diameter 114,4 mm (4,5 in) 114,4 mm (4,5 in)
E=elastic modulus of steel 200000 Mpa 200000 Mpa
GROUT PROPERTIES
Dg=external diameter 200 mm 200 mm
dg=internal diameter 114,4 mm 114,4 mm
Ag=internal cross sectional area 103 cm2 103 cm2
I=internal moment of inertia 841 cm4 841 cm4
E=elastic modulus of grout 23500 Mpa 23500 Mpa
DESIGN RESISTANCE
N=axial load 1070 kN 740 kN
NB=buckling load 2534 kN 1860 kN
M1=flexural moment at N=0kN 44 kNm 36 kNm
M2=flexural moment at N=Nd 27 kNm 20 kNm
V=shear force 122 kN 57 kN
Table 7.2. Characteristics of micropiles.
Each micropiles configuration differs from the others in terms of type, number and/or inclination of
micropiles, as shown in Table 7.3. Micropile type and number were directly linked to vertical forces
from the superstructure, while inclination was provided in order to sustain horizontal forces. Groups of
micropiles with an inclination angle (α)=10° with respect to the vertical direction have been considered.
Selected micropiles configurations appeared feasible to perform retrofit of foundation for all considered
case studies. Furthermore, since all configurations were realized by using only two different types of
micropiles, a direct comparison between retrofit solution for the foundation of all the case studies was
possible.
CONFIGURATION
OF MICROPILES
MICROPILE
TYPE
NUMBER OF
MICROPILES
INCLINATION
ANGLE X-DIR.
INCLINATION
ANGLE Y-DIR.
B00 Type b 2 0° 0°
B10 Type b 4 10° 0°
B01 Type b 4 0° 10°
B11 Type b 4 10° 10°
A10 Type a 4 10° 0°
A01 Type a 4 0° 10°
A11 Type a 4 10° 10°
Table 7.3. Configurations of micropiles.
117
The configurations of micropiles are represented in figure 7.3. Configurations B10 and B01 differ only
in terms of direction of the micropiles inclination, as well as configurations A10 and A01. For each
configuration the relation P-d (vertical force-vertical displacement) are represented. The P-d curves
have been limited to the design resistance evaluated in accordance with Eurocodes.
Figure 7.3. Configurations of micropiles and P-d curves.
By using forces acting on each configuration of micropiles, several spring elastic constants for the
evaluation of the effects of foundation flexibility were computed. Ten different spring constants have
been used, depending on the configuration of micropiles and the forces acting at foundation levels, as
briefly summarized in Table 7.4.
CONFIGURATION B00 B10 B01 B11 A10 A01 A11
SPRING
CONSTANT T11 T12 T21 T22 T13, T14 T31
T23, T32,
T33
Table 7.4. Spring labeling for configurations of micropiles.
Micropiles were added adjacent to existing foundation in order to adequate their compression/tension
capacity and anchored to plinths for load transfer. Retrofit solutions for the considered case studies are
represented in figure 7.4. Added elements are designed to satisfy performance requirements for both
vertical loads and seismic actions.
-0,02
-0,018
-0,016
-0,014
-0,012
-0,01
-0,008
-0,006
-0,004
-0,002
0
-2500-2000-1500-1000-5000
d (m
)
P (kN)
-0,025
-0,02
-0,015
-0,01
-0,005
0
-5000-4000-3000-2000-10000
d (m
)
P (kN)
-0,025
-0,02
-0,015
-0,01
-0,005
0
-5000-4000-3000-2000-10000
d (m
)
P (kN)
-0,03
-0,025
-0,02
-0,015
-0,01
-0,005
0
-7000-6000-5000-4000-3000-2000-10000
d (m
)
P (kN)
-0,03
-0,025
-0,02
-0,015
-0,01
-0,005
0
-7000-6000-5000-4000-3000-2000-10000
d (m
)
P (kN)
Configurations B10-B01
Configuration B11 Configurations A10-A11
Configuration B00
Configuration A11
118
Figure 7.4 Retrofit solutions for the foundation system.
7.3. Influence of foundation retrofitting Based on the results obtained, a comparison between the solutions found for the rehabilitation of
foundation has been done. All the retrofitted buildings achieves same performance requirements and are
subject to the same level of seismic forces and vertical loads. In Table 7.5, a synthesis of the retrofitted
foundation characteristics in terms of number and type of micropiles are reported for each retrofit
solution.
RETROFIT SOLUTION MicropilesType a (inclined) MIicropiles Type b (inclined)
Concentric braces 48 (40) 44 (16)
Eccentric brace 48 (40) 44 (24)
LGS shear walls 24 (24) 64 (24)
Ductile shear walls 60 (60) 60 (0)
Table 7.5. Characteristics of retrofit solutions for foundation.
As a synthetic result, it has been recognized that:
1. Ductile shear walls require more piles than the other retrofit solutions (120 piles)
2. Concentric braces and eccentric braces require the same number of piles (92 piles) but there are
more inclined piles for eccentric brace (64) respect concentric braces (56)
3. LGS require the minimum number of piles (88 piles) and of inclined piles too (48)
In terms of materials, the LGS shear walls require minimum amount of steel for retrofit of foundation,
119
while braces and ductile shear walls require a larger amount, respectively +17% and +52% respect to
the LGS shear wall retrofit system.
Shear walls retrofit solutions have been optimized in order to reduce forces acting on foundation
elements. LGS shear walls have been modified by introducing holes in the middle part of the walls,
making them less stiff; ductile shear walls have been modified by a redistribution of resisting elements.
The optimization leaded to significantly lower forces at the base of the building. For the LGS shear
walls, the average reduction of the vertical forces was ranging between -26% and -20% while for the
horizontal forces was between -41% and -38%, depending by the direction of the prevalent seismic
action. For the ductile shear walls, the average reduction of the vertical forces was ranging between -
36% and -35% while for the horizontal forces was between -35% and -20%, depending by the direction
of the prevalent seismic action.
The considered retrofit solutions for the vertical resisting system were found to be significantly
demanding for foundation elements considering also the poor quality of the ground (type C with low
bearing capacity). In such condition, the analysis performed showed that foundation retrofit cannot be
neglected when rehabilitation strategies are chosen and that effects at foundation levels can be
effectively used as decision criterion in the design process of optimized retrofit systems.
7.4. Connection system between new elements and existing foundation The retrofitting techniques at the foundation level must be designed taking into consideration a proper
flow of the forces from the structure to the ground, without having weak part o ‘bottle-neck’ areas in
which the demand imposed by flowing stresses could overpass the capacity of the system.
In particular, fastening zones between new micro-piles and existing foundations deserve a detailed
design and checking in order to guarantee a proper working condition during all seismic events: no
damage has to occur also under very rare earthquakes. So appropriate details must be considered and
action levels correspondent to the maximum demand expected on the entire structure.
Figure 7.5 (a)typological scheme of the intervention technique with micro-piles; (b) in-field work for
realizing connection system between micro-piles and existing foundation.
In figure 7.5 a schematic representation of the micro-piles and a photo showing a typical applicative
example are reported: it can be noted in such examples that the contact between new elements and
existing ones was critical part of the intervention and the transferring of force through the interface can
be realized using steel reinforcement details as well as friction properties between surfaces in contact.
This two mechanism are largely accepted for the design of prefabricated concrete elements assembled
using dry connecting systems, as reported in EN1992-1 where the maximum shear force that can be
transferred through such connection is equal to:
cossinmin,, ydnctdiRd ffcV (7.1)
where c and m are the friction coefficient and they depend on the roughness of the surfaces in contact;
120
fctd,min is the minimum tensile strength between the two materials in contact; n is the compressive force
eventually acting orthogonally to the surfaces in contact; is the reinforcement ratio (i.e. steel
reinforcement spread along the two surfaces in contact); is the inclination angle between
reinforcement and surfaces in contact. The contribution related to the c coefficient strongly depends on
the quality of the work carried out in-field: concrete shrinkage or surfaces not appropriate worked to be
rough enough could endanger this contribution; moreover, cyclic features of the seismic actions could
endanger the friction effectiveness. The second contribution is that related to and n: this part
represents the friction that could be exploited when a certain level of pre-stress (i.e. compression) is
acting perpendicularly to surfaces in contact. The effectiveness of such contribution can be relevant for
the resistance of the connecting system, but appropriate special details should be realized using, for
example, dywidag devices for squeezing together new elements and existing ones of the foundation.
The third contribution is related to the presence of the shear reinforcement that mechanically re-
establishes shear connection between elements.
Among analysed mechanism, the latter can be considered in all the situations, but its contributions alone
could bring to solutions very expensive in which hundreds of holes have to be realized between the
elements to be connected, see figure 7.5. The former mechanism strongly depends from execution
variables that can be controlled or estimated with a certain difficulties especially during the design
phase; for such reasons, within seismic applications, this first contribution should be neglected in all
calculations.
The intermediate contribution must be considered only when the forces transferring that has to be
realized is so demanding that the adoption of the steel reinforcement contribution makes the solution
not feasible.
121
8. Experimental testing
8.1. Experimental investigations on Steel Shear Walls for seismic retrofitting In this section experimental investigations on Steel Shear Walls (SSW) as vertical element for seismic
retrofitting and upgrading of existing reinforced concrete (RC) structures are presented. The test
program comprises 18 monotonic tension tests on connections between the shear panel and the
boundary elements of the SSW realized by welding and powder actuated fasteners. Furthermore, five
full scale cyclic tests on a pure RC-frame as reference, on pure SSW’s with different shear panels and
connection types as well as on RC-frames retrofitted by SSW’s were carried out. A new developed
connecting system between the SSW and the existing RC-structure was investigated directly in the full
scale tests to consider the realistic stiffness and strength of both members.
8.1.1. Tests on connections between shear panel and boundary elements SSW’s consist of two components: the shear panel as dissipative element and the boundary elements,
which should remain elastic. Usually the connection between these elements is established by bolts or
welds. Bolted connections however, has been found as unfavourable for construction purpose due to the
high requirements on the precision. Hence, for shear panels with a minimum thickness of 4 mm welded
connection were used and tested in the experimental program. As panels with a thickness below 4 mm
can not be welded with common welding technologies on construction site, other connection systems
have to be applied. For this purpose powder actuated fasteners provide advantages due to their high load
capacity, simple construction sequences and the erection is regardless of weather conditions. For both
connections types fin plates (t = 10 mm, S355) at the boundary elements were used as point of
attachment to provide a sufficient strength without stress concentration as well as to guarantee simple
and accurate assembling.
In order to determine a safe but economic arrangement of powder actuated fasteners, several tension
tests were performed with different spacing of the fasteners and two kind of steel grades for the shear
panels. The aim of these tests was to design the connection in such a way that extensive yielding of the
basic material of the shear panel is utilized and premature failure of the connection is prevented. In
general, connection with fasteners can fail due to hole bearing, net section failure or shear/tension
failure of the fasteners. While the first case provides some ductility, the latter failure modes are brittle
and should be prevented. Hole bearing and net section failure are directly related to the tensile strength
and thickness of the basic material and therefore the capacity of the connection is related to the material
properties of the shear panel. To overcome this problem the resistance of the connection was increased
by crimping the panel in the connection area to double the thickness of the panel and / or by using a
material where the yield strength is significant lower than the tensile strength. Such material properties
are provided by the steel grade DX56D, which is usually applied for cold forming (e.g. DX56D: fu/fy =
1.53 instead of DX51D: fu/fy = 1.11 and). Furthermore, this steel grade has a guaranteed maximum yield
stress, which leads to advantages for capacity design in seismic applications.
The test program comprises displacement-controlled monotonic tension tests on welded connections
with shear panels t = 4 mm in steel grade S235 as well as on connections with powder actuated
fasteners with shear panels t = 1 mm in steel grade DX51D and DX56D according to EN10346. The
connections were established with an angel of 43° to obtain similar stress conditions than in the test
(expected angle of the shear panel tension strips), Figure right. The width of the specimen was 89 mm
and therefore the section area of series 1 was A = 356 mm² and of series 3 and 4 A = 89 mm².
8.1.2. Tests on welded connections
Series 1 consist of three specimens with a 4 mm thick shear panel in S235 which was welded to a 10
mm thick plate in S355 by two fillet welds.
In all tests the observed failure mode was rupture of the basic material after considerable yielding. The
failure mode was not affected by the welds (Figure, top right). The load deformation curves of the three
specimens confirm that the full strength and ductility of the basic material was utilized (Figure, top
left). The average deformation capacity was 54.2 mm. Hence, this connection system is suitable for
SSW-shear panels and was used in the full scale tests 2 and 4.
123
Series Connection type Steel
grade
Spacing
[mm]
fy
[N/mm²]
fu
[N/mm²]
A
[%]
1 Welding S235 - 304 394 30,4
2.1 4 fasteners DX51D 33
367 / 402
*) 426 31,7
2.2 2 fasteners, panel
crimped
DX51D 65
2.3 4 fasteners, panel
crimped
DX51D 33
3.1 4 fasteners, panel
crimped
DX56D 33
157 / 177
*) 287 45,7
3.2 2 fasteners, panel
crimped
DX56D 65
3.3 3 fasteners, panel
crimped
DX56D 43
Table 8.1. Test program on connections and mechanical properties of the tested shear panels; *) yield
strength measured in longitudinal and orthogonal direction of rolling
8.1.3. Tests on connections with powder actuated fasteners, steel grade
DX51D In total the test program on powder actuated fasteners with shear panels in DX51D comprises three
different configurations, which include different spacing between the fasteners and partially crimping of
the panel (table 8.1).
The load deformation curves of these tests are shown in figure 8.1, middle left. In series 2.1 (four
fasteners without crimping) and series 2.2 (two fasteners with crimping) hole bearing was the governing
failure mode. The load capacity in both configurations is similar, which leads to the conclusion that
crimping of the panel double the hole bearing capacity. However, the full strength and ductility of the
basic material is not utilized. The specimens of series 2.3 showed different failure mechanisms: Cross-
section failure in test 2.3-2; combined tension and shear failure of the fasteners and then hole bearing
failure of the remaining fasteners in test 2.3-1; combined tension and shear failure of all fasteners in test
2.3-3. The different failure mechanisms are an indication that in series 2.3 the capacity of all failure
modes (net-section failure, hole bearing, shear/tension failure of the fastener) was close together.
Finally the configuration of series 2.3 was used in the full scale SSW test 3 (1 mm thick shear panel in
DX51D). However, the average deformation capacity of this connection (12.2 mm) was still not
satisfactory so that further investigations with steel grade DX56D were carried out.
8.1.4. Tests on connections with powder actuated fasteners, steel grade
DX56D The test series on powder actuated fasteners with shear panels made of DX56D comprises three
configurations with different spacing between the fasteners (see table 8.1). In all tests the panel in the
connection area was crimped.
The specimens in series 3.1 and 3.3 provide considerable deformation behaviour. They failed after
extensive yielding due to rupture of the basic material within the section of the last fastener(Figure,
bottom right). Both configurations showed a sufficient average deformation capacity of 26.9 mm and
31.7 mm respectively. The configuration of series 3.3 has advantages due to the reduced number of
fasteners. In series 3.2 shear failure of the fasteners occur after slight yielding of the basic material. The
deformation capacity was not satisfactory (12.8 mm). In the full scale test 5 the connection type of
series 3.1 was applied.
8.1.5. Test on Steel Shear Walls The test program on SSW’s comprises cyclic full scale tests on a pure RC-frame as reference, on pure
SSW’s and on SSW’s as retrofit measure of RC-frames (see table 8.2).
The RC-frames in test 1, 4 and 5 - made of a concrete strength of C16/20 (fc,cyl = 23 N/mm²) - had a
height of 3.4 m and a span of 4.8 m, while the cross-sections of columns and beam were 300 x 300 mm.
The longitudinal reinforcement in the beam was 4
124
A. Additionally stirrups 6 were placed with a spacing of 25 cm in the columns and of 15 cm in the
beam. The column feet were hinged connected to the strong floor.
The layout of the SSW in test 2 to 5 was identical excepting the shear panels figure 8.2. The height of
the SSW’s (h = 2.6 m) was fixed due to height of the RC-frame, while the span of L = 1.2 m was
chosen to obtain a sufficient resistance. However, this led to an inappropriate length-to-height-ratio,
which would cause high bending moments in the columns and an unfavourable angle of the tension
zone in the shear panel. Therefore an additional horizontal stiffener was applied to subdivide the SSW
in two shear panels with a length-to-height-ratio of about 1. The boundary elements were made of
HEB300 and the stiffener of HEB200, all in S355. The actual material properties of the shear panels in
S235, DX51D and DX56D are identical to the pre-test (seetable8.1).
Test set-up
Series 1
Series 2.1
Series 2.2
Series 2.3
Series 3.1
Series 3.2
Series 3.3
Figure8.1. Load deformation curves and failure modes of tension tests on connections: series 1 (top),
series 2 (middle) and series 3 (bottom).
The connection between SSW and RC-frame in test 4 and 5 was established by a transfer beam made of
two U300-profiles, which were placed on both sides of the RC-beam. The U300-profiles were
connected at the RC-beam by rods next to the corner, which were inserted through vertical holes in the
RC-beam and grouted afterwards. The U-profiles transferred not only the horizontal but also vertical
forces between SSW and RC-frame. This led to a significant reduction of vertical support forces at the
base-points of the SSW without adding unfavourable shear forces into the RC-beam. The same transfer
beam was also used as a hinged steel frame in test 2 and 3 to apply the horizontal load in the same
manner than in the other tests.
0
20
40
60
80
100
120
140
0 10 20 30 40 50 60jack displacement [mm]
forc
e [
kN
]
1
0
5
10
15
20
25
30
35
0 10 20 30 40jack displacement [mm]
forc
e [
kN
]
2.2
2.1
2.3
0
5
10
15
20
25
30
35
0 10 20 30 40jack displacement [mm]
forc
e [
kN
]
3.1
3.2
3.3
125
Test RC-
frame
SSW Shear panel
1 yes - -
2 - yes t = 4 mm, DX56D, welded
3 - yes t = 1 mm, DX51D,
fasteners
4 yes yes t = 4 mm, DX56D, welded
5 yes yes t = 1 mm, DX56D,
fasteners
Table 8.2. Test program on full scale Steel Shear Wall
Figure 8.2. General layout of Steel Shear Walls as retrofit measure of a RC-frame (test 4 and 5)
8.1.5.1. Loading procedure and measurements The load was applied as compression force at the corners of the outer frame by an actuator anchored to
the reaction wall (figure 8.3, top right). The loading procedure complied with the ECCS-guideline
“Recommended Testing Procedure for Assessing the Behaviour of Structural Steel Elements under
Cyclic Loads” with increasing displacement amplitudes after each three cycles. The reference
displacement y used in this procedure was determined analytically at the elastic limit.
Besides force and displacement of the actuator, the horizontal displacement of the RC-beam and the
SSW at the upper boundary element and the stiffener as well as the column foot rotations was
measured. Furthermore, strain gauges were applied in the corners of the shear panels and at critical
areas of the boundary elements and strain gauge rosette were used in the centre of the shear panels to
determine the angle of the tension zone.
126
Figure 8.3. Test set up of test 5 and load deformation curves of test 1 to 5.
8.1.5.2. Test 1: pure RC-frame In the first test the behaviour of the pure reinforced concrete frame under cyclic loading was analysed as
reference for test 4 and 5. The specimen was loaded displacement controlled up to a maximum
displacement of 180 mm. During the test the frame reached a maximum load of 13 kN. The collapse of
the specimen was caused by high bending moments at the beam-column joints at a displacement of 140
mm. Even if the stiffness and load capacity was quite low, the deformation capacity was significantly
higher than expected and known from literature, (e.g. maximum deformation capacity according to
FEMA356 crit = 31 mm). Possibly, the very low concrete compressive strength has led to an increase
of rotation capacity even if the stirrup spacing was insufficient according to current standards.
8.1.5.3. Test 2: Steel Shear Wall with welded shear panel in S235 In the second test a pure SSW was tested with 4 mm thick shear panels in S235 welded to the boundary
frame. The first visible buckling could be detected at a displacement of about 30 mm. At this point the
SSW system already carried a load of almost 700 kN. The maximum load capacity of the specimen was
825 kN at a displacement of 76 mm. At 80 mm displacement the first crack next to the lower horizontal
welds of the lower shear panel occurred, figure 8.4. In the course of the test further cracks occurred next
to the horizontal welds, which grew with each cycle and led to a first significant load drop of about 100
kN at a displacement of 106 mm. In the following the cracks also extended to the vertical welds. The
test procedure was stopped at a displacement of 139 mm after complete rupture of the shear panels at
the horizontal welds. The SSW showed a good load bearing capacity also in the elastic range and
offered an excellent ductile behaviour.
actuator
SSW
RC-frame
Test 1
Test 2
Test 3
Test 4
Test 5
127
Figure 8.4. First cracks next to the welds (left) and buckling behaviour at 80 mm (middle) as well as
at the end of the test (right) (Test 2)
8.1.5.4. Test 3: Steel Shear Wall with shear panel in DX51D fixed by
fasteners In test 3 a shear panel in DX51D with a thickness of 1 mm was used, which was fixed by fasteners. The
nominal spacing of the fasteners was 33 mm and the edge distance 20 mm, while the shear panel was
crimped in the connection area. Compared to test 2, buckling occurred very early already after a few
millimetres of displacement. At a displacement of 36 mm hole bearing failure of the horizontal
connection started at the lower panel. The load of 206 kN at this displacement was not exceed during
the remaining loading procedure. At this time also local buckling in the crimped area was visible. In the
second cycle at a displacement amplitude of 56 mm cracks grew through the horizontal connections and
the load decreased rapidly within the following cycles. Finally, the horizontal connections collapsed,
while also considerable cracks in the vertical connections were visible. Due to the premature failure of
the connection, the system lost its capacity earlier than the SSW in the test 2, which led to a reduced
ductility.
Figure 8.5. Cracks through the net section area of the section (left) and buckling behaviour at 36
mm displacement (Test 3)
8.1.5.5. Test 4: RC-frame retrofitted by Steel Shear Wall with welded shear
panel in S235 In test 4 retrofitting of a RC-frame by a SSW identical to the system used in test 2 was investigated. In
general the load deformation behaviour was very similar to test 2, as the resistance of the RC-frame was
small in comparison with the SSW. The first buckling in the shear panels occurred at a displacement of
28 mm, while the load was 700 kN. Again, the cracks in the concrete were concentrated at the corners
of the frame. The maximum load of 857 KN was reached at 84 mm displacement. Afterwards the load
decreased with each cycle, as cracks grew next to the horizontal welds between the shear panels and the
frame of the SSW. The test was stopped at a displacement of 140 mm, where the specimen carried less
than 200 kN. The failure mode of the SSW was similar to test 2 due to rupture of the shear panel next to
the horizontal welds between panels and frame. No reduction of stiffness and capacity of the RC-frame
could be measured, as the behaviour of the SSW was dominant. The connection system between SSW
and RC-frame transferred the forces sufficiently. The slippage between U-profiles and concrete beam
was negligible. The system of test 4 showed an almost similar load capacity and the same good ductile
128
behaviour than test 2.
8.1.5.6. Test 5: RC-frame retrofitted by Steel Shear Wall with shear panel in
DX51D fixed by fasteners In test 5 again a RC-frame retrofitted by a SSW was tested; however, the shear panel was made of 1
mm thick sheets in DX56D and connected by fasteners in a similar way than in test 3.
Right after a few millimetres of displacement the shear panels started to buckle. After 80 mm of
displacement the first fasteners failed at the corners of the panels. At this time a maximum load of 178
kN was applied to the SSW system. The connection of the panels performed significantly better than in
test 2 due to the lower yield stress of the DX56D sheet. Failure of the connection occurred at similar
displacements than for the welded connections. The load deformation behaviour showed an excellent
ductile behaviour of the system. Even after many cycles in the plastic range the system behaved very
stable. Hence, the SSW with shear panels with low yield strength offers a considerable higher ductility
than test 2. The common bearing behaviour of RC-frame and SSW was again sufficient.
8.1.5.7. Evaluation of test results according to the ECCS-procedure To characterize steel elements under cyclic loads the ECCS recommendation “Recommended Testing
Procedure for Assessing the Behaviour of Structural Steel Elements under Cyclic Loads” provides
several parameters to characterize the seismic behaviour of the tested SSW systems. As no monotonic
loaded pre-tests were performed on the SSW’s the yield force and the corresponding displacement in
the positive and negative range was directly determined based on the recorded cyclic load deformation
curve.
In figure 8.6 the relative resistance functions of the tested SSW’s are plotted against the partial ductility.
The relative resistance is defined as the minimum peak load of three cycles with the same displacement
amplitude divided by the yield force. The partial ductility is determined with the corresponding
displacement divided by the displacement at the elastic limit. The result is a dimensionless skeleton
curve which visualizes the ductility of each SSW. As expected, test 2 and 4 lead to similar results as the
capacity of the RC-frame is negligible in comparison to the SSW. The discrepancy between these
curves can be explained by the uncertainty of the yield displacement definition and its big influence on
the determination of the partial ductility. The ductility ratio of the SSW with welded shear panel is
between 3.5 and 5.5. In contrast, the ductility ratio of the SSW in test 3 is about 1 as premature failure
of the connections led to an early reduction of the resistance. However, very good behaviour is also
obtained for the SSW system used in test 5. It shows the most stable behaviour in the plastic range and
leads to a ductility ratio of 8. Furthermore, the significant hardening in the plastic range leads to some
additional safety margin.
Another important parameter is the resistance drop ratio, which is defined as the decrease of load
capacity during the cycles at the same displacement. The curves of test 2 and 4 confirm the
aforementioned ductility ratio, but show also the sudden resistance drop at the end of the testing
procedures, figure 8.7. The SSW in test 5 provides a significant more robust behaviour due to the slow
and continuous resistance drop fall. The curve of test 3 shows again the resistance drop at an early
stage, but shows also some residual strength up to a ductility ratio of 5.
Figure 8.6. Relative resistance function of test 2 to
5.
Figure 8.7. Resistance drop ratio function of test
2 to 5.
-1.5
-1.0
-0.5
0.0
0.5
1.0
1.5
-10 -8 -6 -4 -2 0 2 4 6 8 10
partial ductility [-]
rela
tiv
e r
es
ista
nc
e [
-]
2
3
4
5
0.0
0.2
0.4
0.6
0.8
1.0
1.2
-10 -8 -6 -4 -2 0 2 4 6 8 10partial ductility [-]
resis
tan
ce d
rop
ra
tio
[-]
2
3
4
5
129
8.1.5.8. Tests on connection system between Steel Shear Wall and existing
structure As mentioned previously, a transfer beam consisting of two U300-profiles was used in test 4 and 5 to
connect the SSW to the RC-frame. The U-profiles were attached at the left and right side of the RC-
beam, while steel plates connect the U-profiles at their flanges. The connection between steel plates and
RC-beam was established by post-installed anchors, which were inserted vertical through the beam,
figure 8.8. After the whole system was built-on the anchors were grouted.
The transfer beam has several advantages:
1. Reduction of vertical reaction forces in the foundation of the SSW
2. Additional shear forces are prevented in the RC-beam
3. Only axial forces introduced in RC-beam and RC-columns
The assembling procedure of the insert through anchoring is as follow:
1. Core drilling in RC-frame
2. Erection of steel shear wall and transfer-beam
3. Insertion of anchors
4. Grouting of rods
The design of the insert through anchoring is carried out. The measured relative displacement between
transfer beam and RC-frame was negligible (< 0.05 mm). The connection carried the transfer forces
sufficiently without significant slippage.
Figure 8.8. Connection between SSW and RC-frame: Transfer beam and insert through anchoring
(left), hinged connection between transfer beam and SSW (right)
8.1.6. Tests on connection system between new roofing / floor systems and
existing structures Unfavourable diaphragm action of existing floors and roofs subjected to seismic loads can be upgraded
by various techniques. These retrofitting measures are connected to the walls of the existing structure
and act mainly in tension. For RC-structures many certified connection systems for seismic loads are on
the market (e.g. undercut anchor systems). However, the design of connections in masonry is still
afflicted with uncertainties. Hence, a test program is performed to determine the stiffness and strength
of such connections under defined conditions.
8.1.6.1. Test program and test set up Two displacement controlled tension tests on anchors insert in masonry walls were carried. The
dimensions of the brick wall were 1500 x 1500 mm, with a thickness of d = 175 mm in test 1 and d =
240 mm in test 2, figure 8.9. The lime-sand bricks fulfilled the requirements of class 12 (1.2 N/mm²)
and a mortar class II (friction 0.04 N/mm²) in according to DIN 1053-100. The steel elements consisted
of a steel connector plate (500 x 500 mm, t = 20 mm) and a tie rod (Ø 20 mm), all in steel grade S235.
The masonry was supported by a steel frame on four sides and the tension force was applied on the tie
rod.
8.1.6.2. Test results In test 1 already after a few millimetre displacement cracks occurred between mortar and bricks, as the
bond was very poor. The cracks propagated in diagonal direction through the whole masonry wall and
130
the maximum load of 2.6 kN was reached after 10 mm. Afterwards the loads stayed rather stable until
the end of the test, which could be explained by friction forces between mortar and bricks.
In test 2 some horizontal cracks between mortar and bricks occurred very early succeeded by diagonal
cracks similar to that one in test 1. However, after passing a load plateau between 2 and 7 mm the force
increased again until the maximum load of 6.7 kN at a displacement of 17 mm was reached and the load
dropped off. This can be explained by the development of a compression vault within the masonry wall.
In both tests the load capacity was rather low. However, the governing failure mechanism was not
directly the connecting system (e.g. due to punching), but the failure mechanism included the whole
wall. Hence, the anchor was sufficient to transfer the load into the wall, even if the load capacity of the
wall itself is low.
Figure 8.9. Test set-up for connection in masonry
wall
Figure 8.10. Load deformation curves of
connections in masonry wall with two different
thicknesses d
8.2. Experimental Qualification of BRB systems for seismic retrofitting of
R.C. frames The seismic performance of an existing RC building was analyzed by using nonlinear static and
dynamic analysis. The structure showed very poor ductility and failed in a brittle manner. The structure
was retrofitted by means of Buckling Restrained Braces (BRB) and Concentrically Braced System
(CBS). The application of the BRB retrofitting technique showed an important improvement, especially
in strength and stiffness, but also in ductility. Based on the good results obtained a testing program was
developed in order to prove the efficiency of the retrofitting system based on BRB and CBS. The
retrofitting systems were applied to a RC portal frame, selected from the RC building.
The experimental program aims at evaluating the performances of the retrofitted structure. The
performances of the BRB and CBS system are evaluated in terms of acceptance criteria. The connection
of the retrofitting systems to the existing concrete frame structure is very important, both in terms of
performance and workability.
The RC frame extracted from the RC building is located at the second floor on Y direction. The main
reason for selecting the frame from this floor comes from the limitation of the testing capacity in the
Laboratory. Concrete elements of this floor are reduced, compared to the elements of the lower floors.
All details regarding the number of rebars, the distribution of rebars in element cross section, the
distance between stirrups (15 cm for columns and 25 cm for beams) and diameters were similar to those
from the Benchmark structure. The cover concrete was considered 2.5 cm.
F
500
50
0
d1500
15
00
0
2
4
6
8
0 10 20 30
forc
e [
kN
]
jack displacement [mm]
d = 240 mm
d = 175 mm
131
a) b)
Figure 8.11. a) RC frame location - 3D view; b) RC elements cross sections (columns and beam)
As the frame selected for the experimental program is an interior frame, the longitudinal reinforcements
from the columns and beam need to be anchored appropriately. In order to assure a sufficient anchorage
length, the rebars were bent so as to assure a sufficient anchorage length. In order to limit the influence
on the strength capacity of beams and columns, the bent was made inside the beam–column joint.
a) b)
Figure 8.12. RC frame and node details: a) rebars bent in the joints; b) formwork of the concrete frame
In order to keep the same construction details, plane rebars were used for all reinforcements. The results
of the coupon test on the steel from the BRB core plate are presented in the table from Figure 8.14.
Also, Figure 8.14, shows details of the test specimens and presents the stress-strain curves for BRB
steel core plates.
a) b) c)
Figure 8.13. a) Theoretical vs. quality certificate vs. experimental rebars samples material
characteristics; Characteristics of the concrete used for: b) RC frame; c) BRB infill material
Materials used for RC Frame Theoretical Quality Certificate Experimental
Standard
Stirrups Φ6 OB37 OB37 Specimen Test
Minimum Yield strength Re [N/mm2] 235 289 - 303 NA
Tensile strength Rm [N/mm2] 360 402 - 424 NA
Minimum Elongation % 25 38.0 - 41.5 NA
Materials used for RC Frame Theoretical Quality Certificate Experimental
Standard
Beam rebars Φ14 OB37 OB37 Specimen Test
Minimum Yield strength Re [N/mm2] 235 312 497
Tensile strength Rm [N/mm2] 360 448 623
Minimum Elongation % 25 36 31
Materials used for RC Frame Theoretical Quality Certificate Experimental
Standard
Column rebars Φ18 OB37 OB37 Specimen Test
Minimum Yield strength Re [N/mm2] 235 287 402
Tensile strength Rm [N/mm2] 360 402 537
Minimum Elongation % 25 38 25
STAS 438/1-89 & ST 009 - 2005
STAS 438/1-89
STAS 438/1-89 & ST 009 - 2005
Concrete material for RC frame (1m3):
(C20/25 => Rc = 20.5 N/mm2)
- aggregates: 1708 Kg
type I: (0-4) mm – 632 Kg
type I: (4-8) mm – 427 Kg
type I: (8-16) mm – 649 Kg
- cement: II BM(S-V)32.5R - 400Kg
- additive: BV3M (2l)
- water: 195l
=> Rc = 35.5 N/mm2 (28 days)
Concrete material for BRB infill (1m3):
(C25/30 => Rc = 24.3 N/mm2)
- aggregates: 1660 Kg
type I: (0-4) mm – 614 Kg
type I: (4-8) mm – 415 Kg
type I: (8-16) mm – 631 Kg
- cement: II BM(S-V)32.5R - 430Kg
- additive: BV3M (1%-from cement)
- water: 195l + 10l
=> Rc = 35.1 N/mm2 (22 days)
132
Figure 8.14. BRB steel plate specimens, material characteristics of the BRB steel core plates and stress-
strain curves for BRB steel core material
The BRB elements were manufactured and tested in the Laboratory of Steel Structures from the
“Politehnica” University of Timisoara. The following operations were performed: mechanical cut,
welding of the web stiffeners, positioning of the polystyrene, wrapping of the unbonding material (PVC
transparent foil, 1mm thick), insertion and calibration of the wrapped steel core into restraining steel
tube and the filling up of the infill material (concrete). In Figure 8.15, the same parameters are
presented for CBS a circular hollow tube 101.6x3.6.
Figure 8.15. CBS steel plate specimens, material characteristics of the BRB steel core plates and stress-
strain curves for BRB steel core material
8.2.1. Testing set-up The scheme with the testing rig and the loading system Figure 8.16.a, while in figure 8.17.b and figure
8.17.c, the RC frame and RC frame+BRB installed in testing rig is presented.
(a)
(b)
(c)
Figure 8.16. Testing rig and the loading system: a) scheme of the testing rig; b) RC portal frame and
BRB system (MRF+BRB); b) RC portal frame and CBS system (MRF+CBS)
133
Pinned connections have been used between the BRB elements and the beam and at the base of the
columns. In order to prevent the slip of the connection between the BRB and the RC beam, high
strength preloaded ties have been used. The effectiveness of the connecting device has been preliminary
checked by FEM simulation. The maximum force applied to all bolts (Ft x nbolts) by bolt pretension (Ms
= 200Nm), creates a pressure (σpl) which is smaller than the compressive strength of the RC beam.
Consequently, the friction force (Ff) between the steel plate and the concrete element should be larger
than the cumulated horizontal BRBH force. CBS – RC Frame connections system were the same as for
BRB system tests.
(a)
(b)
(c)
(d)
Figure 8.17. Connection details of: a) BRB and RC column; b) BRB - RC beam; c) CBS and RC
column; d) CBS - RC beam
The numerical simulation aimed to calibrating the level of pre-stressing forces in the ties in order to
avoid the slippage of the connection. Local pressure on the concrete was also checked, in order to keep
the connection “elastic”. In order to monitor the connection between BRB/CBS and the RC columns,
four measurement devices were applied on the bottom of each column and two monitoring devices were
installed on the RC beam in order to monitor the slippage of the connection between the retrofiting
system and the RC element. Also, displacement transducers were assembled in order to measure the top
displacement and axial displacements of each BRB/CBS element.
8.2.2. Experimental Results
8.2.2.1. Monotonic tests Monotonic tests were also conducted on the frame in order to evaluate the yield point. The results from
monotonic tests are also used as reference values when comparing to the cyclic tests. The quasi-static
cyclic testing was carried out according to a loading protocol based on the ECCS Recomandations.
Figure 8.19 shows the force–displacement curves for the initial RC frame MRF and for the retrofitted
frames (MRF+BRB and MRF+CBS). The effectiveness of the seismic strengthening of the RCF frame
by means of a BRB/CBS system is confirmed by the increase of the stiffness and strength.
Figure 8.18. Monotonic tests: a) MRF; b) MRF+BRB; c) MRF+CBS
134
Figure 8.19. Monotonic tests results
8.2.2.2. Cyclic tests The modified ECCS loading protocol was applied in cyclic tests. This modified procedure is
characterized by a single loading at Dy/4, 2Dy/4, 3Dy/4 and Dy, followed by three repetitions of the
cycles increased by 0.5 Dy (1.5Dy, 2Dy). The strain rate in the cyclic tests was 5mm/min, so that the
application of the load was considered quasi-static.
Figure 8.20 and Figure 8.21 show the initial RC frame under cyclic loading test. The distribution of the
cracks from bending and shear are presented in Figure 9 b) and Figure 10 a). Bending cracks occurred
first and were followed by shear cracks. The development of shear cracks is mainly due to the
inadequate distribution of stirrups. Figure 8.20.b) shows the failure of the beam-to-column joint.
(a)
(b)
Figure 8.20. a) RC frame under cyclic load; b) development of bending cracks
(a)
(b)
Figure 8.21. RC frame under cyclic load: a) development of shear cracks; b) failure of the node
Figure 8.22 show the retrofitted RC frame (MRF + BRB) under cyclic loading test. Bending cracks
occurred first and were followed by shear cracks. The development of shear cracks is mainly due to the
inadequate distribution of stirrups. It may be observed that no cracks occurred at BRB – RC beam
connection.
Monotonic experimental tests
0
50
100
150
200
250
0 50 100 150 200 250
Displacement [mm]
Fo
rce
[K
N]
MRF MRF+BRB MRF+CBS
ACBS = 11cm2
ABRB = 3 cm2
135
(a)
(b)
Figure 8.22. a) MRF + BRB under cyclic load, b) bending moment cracks, c) shear cracks at ultimate
stage
Figure 8.23 show the retrofitted RC frame (MRF +CBS) under cyclic loading test. Bending cracks
occurred first followed by shear cracks. Unlike MRF+BRB, in this case cracks occurred at BRB – RC
beam connection due to buckling of the braces.
(a)
(b)
Figure 8.23. a) MRF + CBS under cyclic load, b) bending moment and shear cracks
When the left side BRB failed in tension, the horizontal displacements recorded at the connection
between BRB and the RC beam amounted to 5 mm, only (Figure 8.24. a)). While, in the case of RC
frame retrofitted by CBS many cycles and larger slippage of the beam connection may be noticed
(Figure 8.24.b).
(a)
(b)
Figure 8.24. Hysteretic curve of the connection between: a) the BRB – RC beam; b) CBS – RC beam
BRB (left side)
-200
-150
-100
-50
0
50
100
150
200
-6 -4 -2 0 2 4 6
Beam Connection Displacement [mm]
MR
F F
orc
e [
KN
]
CBS-RC beam connection (Cyclic Test)
-300
-200
-100
0
100
200
300
400
-15 -10 -5 0 5 10 15 20 25
Displacement [mm]
Fo
rce [
KN
]
136
Figure 8.25. The initial RC frame vs. the retrofitted frames
Figure8.25 shows the force – displacement curves for RC frame before and after retrofitting. It may be
noticed the contribution of the retrofitting system in terms of strength, stiffness and ductility. The
behavior of the frame after retrofitting shows similar performances in tension and compression and a
large strain hardening.
Figure8.26, show the force–displacement curves for the left and the right braces. The two hysteretic
curves show similar behavior in tension and compression, a stable plastic behavior and a very large
ductility.
Figure 8.26. a) Left BRB during cyclic test; b) Right BRB during cyclic test
Figure8.27 show the steel core plates after the test (left brace BRB-C-L and right brace BRB-C-R). The
failure of the BRB took place before the failure of the concrete elements.
Figure 8.27. BRB steel core plates during cyclic test
BRB (left side)
-200
-150
-100
-50
0
50
100
150
200
-50 -40 -30 -20 -10 0 10 20 30 40 50
Steel Plate Displacement [mm]
MR
F F
orc
e [
KN
]
BRB (rigth side)
-200
-150
-100
-50
0
50
100
150
200
-50 -40 -30 -20 -10 0 10 20 30 40 50
Steel Plate Displacement [mm]
MR
F F
orc
e [
KN
]
137
8.3. Experimental testing on novel dissipative bracing element The experimental programme carried out for the qualification of intervention techniques about the steel
bracing members, in particular eccentrical braces, was modified and enriched focusing the attention on
the development of a novel dissipative device morphologically similar to a common brace or BRB but
characterized by the following mechanical properties:
Replacing of steel fuses after seismic events;
Re-centering features for having zero residual drift at the end on ground shaking;
Flexible calibration of mechanical properties by means of defining appropriate fuses and re-
centering devices.
This system was named as Flag Shaped Hysteretic Device – FSHD –, currently under patenting process;
the system is completely made of steel and made up of the following components:
an external case;
an internal sliding frame;
a piston used for the introduction of the external load;
2 anchor plates;
a dissipative elements system;
2 prestressing cables.
The pre-stressing cables and the dissipative elements can be suitably defined in order to reach precise
values of yielding stress, energy dissipation, elongation or stiffness. In particular, section of fuses and
the steel qualities can be suitably defined. In particular, different type of steel qualities have been
selected and previously tested in order to have appropriate fuses type, see table 8.3.
Table 8.3. Steel qualities selected for realizing steel fuses preliminary tested.
(a)
(b)
(c)
Figure 8.28. (a) dissipative fuses; (b) testing set-up; (c) buckling restraining system for testing.
Figure 8.29. Cyclic testing on different steel qualities at different maximum strain
C Si Mn P S Cr Nb V Al Ti CEQ
PH10 0,004 0,02 0,18 0,015 0,01 0,04 0,07 0,034
PH20 0,004 0,15 0,2 0,015 0,01 0,035 0,03 0,025 0,044
BH3R 0,035 0,015 0,012 0,005 0,05 0,055 0,078
CH3N 0,04 0,025 0,29 0,015 0,015 0,025 0,01 0,088
B040 0,06 0,02 0,25 0,02 0,055 0,1
RS54 0,08 0,02 0,6 0,02 0,01 0,035 0,05 0,025 0,18
-80.0
-60.0
-40.0
-20.0
0.0
20.0
40.0
60.0
80.0
-0.20% 0.30% 0.80% 1.30% 1.80% 2.30%
-100
-80
-60
-40
-20
0
20
40
60
80
100
-1.00% 0.00% 1.00% 2.00% 3.00% 4.00% 5.00% 6.00% 7.00% 8.00%
138
The pre-stressing cables have a section and a length suitably defined in order to reach desired level of
yielding and maximum elongation (i.e. failure of dissipative device); the cable type is presented in the
figure 8.30 and it is an open spiral strands equipped with adjustable cylindrical socket with threaded rod
provided by Redaelli Tecna Spa.
Figure 8.30. Prestressing cable
The steel fuses were suitably worked in order to be anchored to the internal case and anchor plate inside
the FSHD system, see figure 8.31. They are obtained by dog bone shaped sheet and jointed by friction
bolts to the anchor plate and to the internal frame. The dissipative elements are equipped with a system
that avoid the lateral buckling during the compression phase.
a) b)
Figure 8.31. a) Dissipative element b) buckling restraining system
The other parts of the FSHD are the rigid elements at which pre-stressing cable, steel fuses and existing
structure must be connected.
External case
The external case is made up mainly of 2 sheets 10 mm thick linked as shown in figure 8.32. On one
end the case has a perforated element that allow the connection, by means of a pin, to the external
structure. Within the case four sheets are welded. They are used as leading system for the sliding frame
and as contrast system for the anchor plates. The case is equipped with side panels that shall avoid
buckling phenomena due to the external compression.
Figure 8.32. Global view and sections of external case
500
40
10
170
2307
1895322
Side
panel
B
B
A
A
Sec. A-ASec. B-BLeading and
contrast system
451
258
139
Internal sliding frame
The internal frame is realized with a couple of square hollow element 70x8.3 and 924mm long. Both
element, at both ends, are welded with rectangular hollow elements 160x80x10 and 190mm long.
Figure 8.33. Global view and sections of internal sliding frame
Anchor plates
2 plates, with a thickness respectively of 50mm and 70mm. As shown in figure 8.34, both plates have 4
rectangular openings and 2 welded sheets necessary for the insertion and the joint of the dissipative
elements. The 70mm thick plate has also a circular opening necessary to the insertion of the piston.
Figure 8.34. Connecting plates
Piston
As shown in figure 8.35, it is obtained by a circular hollow element Ф88.9x3.2. It is jointed at one end
to the internal frame by bolts and it has on the other end a perforated plates necessary to join the piston,
by means of a pin, to the external structure.
Figure 8.35. Piston
8.3.1. Test setup Low cycle fatigue tests on the self-centring dissipative deivce were conducted in the "Laboratorio
Ufficiale per le Esperienze dei Materiali da Costruzione" of the Civil Engineering Department at the
University of Pisa. The general test setup is shown in figure 8.36.
As load system has been used a 40 tonns hydraulic actuator, equipped with a load cell and a
displacement transducer. The hydraulic actuator, placed horizontally at an height of 1395mm, has been
connected at one end to the reaction wall and on the other end to a steel structure that assure the vertical
support but allow the horizontal movement of the actuatork. To the same structure the dissipating
device has been linked by a pin joint. The other end of the dissipator has been linked to a concrete wall.
1324
428
A
A
B
B
Sec. A-ASec. B-B
268
14470123144 50 123
260 1080
1540
100
200
140
Fig. 8.36. General test setup
8.3.2. Gauge system In order to measure displacement, strain and load, 8 LVDT (Linear Variable Differential Transformer)
sensors, 14 strain gauges and the hydraulic actuator internal load cell. All these sensors were connected
to a National Instrument Data Acquisition System. Sensors position are shown in figure 8.37. LVDT
Figure 8.37. Sensor position
8.3.3. Testing procedure Short testing procedure suggested by ECCS was used. In this procedure monotonic displacement
increase tests are not foreseen and only the low cycle fatigue test is carried out using a step of
displacement sufficiently small to ensure that at least four levels of displacement are reached before the
yielding displacement.
For the execution of the lab test an initial displacement step of 0.1mm has been used until the
displacement level reaches 0.5mm. Reached this value, the displacement step becomes equal to 0.5mm
and for every displacement level, 3 cycles are performed as schematically shown in fig.5.12. The testing
displacement rate has been fixed equal to 3mm/min.
5613
1752500
1395
2250+150=2400
400 kN IDRAULIC
ACTUATORFSHD DISSIPATOR
REACTION WALL
CONCRETE WALL
538
LVDT Displacement
sensor
Strain gauge
12
89
7
4 3
6
5
8.88.88.88.8 8.8 8.8 8.8 8.8
8.8 8.8 8.8 8.88.88.88.88.8
8.8
8.88.88.88.8 8.8 8.8 8.8 8.8
8.88.88.88.8 8.8 8.8 8.8 8.8
8.88.88.88.8 8.8 8.8 8.8 8.8
8.8 8.8 8.88.88.88.88.8
8.8 8.8 8.8 8.88.88.88.88.8
1 2
8 9
11
1413
10
12
8.8 8.8 8.8 8.88.88.88.88.8
141
Figure 8.38. Displacement history used for the short testing procedure
8.3.4. Results Three main tests were carried out and many pilot tests were carried out (and here not reported for sake
of shortness) also for solving some initial problems due to elimination of internal friction and not proper
working of the prototype and due to the acquisition systems which resolution was lowered in order to
assure a proper working. The initial tests were carried out in order to improve the shape of the flag
hysteresis that in the first trials was irregular due to a not perfect closure of the system and to contact
lack; in the figure 8.39 there is reported a graphs of first test where previous problems happened; in
particular, the curve was not symmetric due to the contact lack in one direction producing the absence
of load bearing.
Figure 8.39. First experimental tests: no satisfactory result due to different behaviour in tension and in
compression
It can be taken from figure 8.9 that, in every cycle, the residual displacement level is lower than 0.5mm
and so the dissipating device has an effective self-centring capacity. It also can be noted that the device
shows a stable hysteresis loops for every displacement level reached during the test, assuring a constant
level of energy dissipation.
The stability of hysteresis loops also during the unloading phase were assured by the presence of the
dissipative element buckling restraining system. In fact during this phase the dissipative elements are
subjected to a compression action that yield the elements. Thanks to the buckling restraining system it
has been possible to plasticize the dissipative element in compression without the presence of a global
lateral buckling, as shown in figure.
-8
-6
-4
-2
0
2
4
6
8
0 500 1000 1500 2000 2500
Dis
pla
cem
en
t [m
m]
Time [s]
Displacement History
-250.00
-200.00
-150.00
-100.00
-50.00
0.00
50.00
100.00
150.00
-10.00 -5.00 0.00 5.00 10.00 15.00
Forc
e [k
N]
Displacement [mm]
Force - Displacement
142
The different behaviour in tension and in compression can be attributed to the excessive transversal
deformation, happened during the test, of one of the welded sheet within the external case and the
subsequent loss of an anchor plate contrast as shown in figure 4.27. This contrast loss caused a different
stiffness of the dissipating device in tension and in compression, but did not compromise the self-
centring capacity of the dissipating device.
Figure 8.40. Loss of contact between the anchor
plate and the welded sheet
Figure 8.41. C-formed element used to assure
the contrast
Currently the problem has been solved with a C-shaped element jointed to the above mentioned welded
sheet that provide a larger contrast surface, as shown in figure 8.41. Other experimental tests were
carried out modifying internal mechanical properties of FSHD components in order to define dissipative
devices suitable for the application to the case study “Bagnone building” where retrofitting technique
was studied.
The modification of steel fuse geometry and the section and pre-stressing rate of the post-tensioned
cable allow to define different FSHD with different yielding level, dissipated energy (i.e. area of cycle),
maximum elongation and hardening ratio, see figure 4.29. In particular, case (a) in figure 4.29 was
defined for the first story of bagnone building where higher yielding level and high energy dissipation
were required. In order to obtain high dissipation level the pre-stressing rate of the cable was set equal
to 50% of its yielding. On the contrary, in the FSHD systems presented in the graph (b), higher
prestressing level was adopted, about 60%, with low resistance fuses, suitable for high storeys of
Bagnone building,
(a)
(b)
Figure 8.42. (a) pre.stress 50% - steel fuses fy=350N/mm2 and section equal to 450 mm
2; (b) pre.stress
60% - steel fuses fy=200N/mm2 and section equal to 300 mm
2
-2000
-1500
-1000
-500
0
500
1000
1500
2000
-60 -40 -20 0 20 40 60
Axi
al F
orc
e [k
N]
Displacement [mm]
-1000
-800
-600
-400
-200
0
200
400
600
800
1000
-60 -40 -20 0 20 40 60
Axi
al F
orc
e [k
N]
Displacement [mm]
143
9. Application to case studies and design guidelines In the present part the analysis of three case studies is presented: a masonry building; a church (stone
and masonry building of high historical value); a reinforced concrete building. For each of them the
same logical process has been followed on the basis of the work conceptually dome in the case of the
benchmark buildings. In particular, each structure has been analysed in order to assess the structural
deficiencies using the selected PBEE framework; then a retrofitting technique has been selected on the
basis of previous studies and analyses carried out within the project; finally, the performance of the
retrofitted structures have been assessed. Each case is presented using the same approach: general
description; vulnerabilities assessment; selection of the intervention technique; final assessment of the
retrofitted structure. The seismic hazard and the reference seismic actions considered in the examples
here reported are summarized in the table 9.1.
Table 9.1. Earthquake levels.
9.1 Patras House
9.1.1 General Description of the building The existing building that has been selected as a Case Study is located Patras, North Peloponnesus. It is
a typical structure of the 1930s with general dimensions 12.25 m x 15.65 m with two levels that has
been used as residence in the rural area of Patras. The structure has suffered many severe earthquakes
since this part of Greece is considered as a High Seismicity area. Its condition prior the strengthening
intervention that took place in 1997 was bad. The engineer responsible for the repair and strengthening
of the building had to remove all the coating in order to reveal any possible damages to the wall body
underneath. Despite the earthquake events this structure suffered, it had limited severe damages. The
exterior walls consist of rumble (field) stones combined with lime mortar of poor quality and they have
a thickness that varies from 65 cm at the ground level to 55 cm at the upper level. The wall thickness of
the main interior walls is 50 cm, while some partition walls have thickness of 20 cm.
The floor system is made of timber and it consists of timber joists that support wider timber plates
placed above the joists in the normal direction. The roof is also made of typical timber rafters placed in
regular intervals. The rafters support timber purlins and above the purlins tiles are used to cover the
whole roof. In the lower part, the roof is covered with ceiling. As a foundation system, a continuation of
the masonry wall below the ground level for about 1.50 meters is used. A general view of the building
and the plan drawing are presented in Figure 9.1.
(a) (b)
Figure 9.1. Plan drawing (a) and general view (b) of the structure
EARTHQUAKE LEVEL RETURN PERIOD
FREQUENT EARTHQUAKE (operational limit state) TR = 30 years
OCCASIONAL EARTHQUAKE (occupancy limit state) TR = 50 years
RARE EARTHQUAKE (life safety limit state) TR = 475 years
VERY RARE EARTHQUAKE (collapse prevention limit state) TR = 975 years
145
9.1.2 Assessment of the structural vulnerabilities The unreinforced masonry structure consisted of rumble (field) stoned with great irregularity, poorly
bonded together by low quality lime mortar, previously assessed by a technician that rehabilitated it
using reinforced concrete based techniques. The estimation, previously made, of the compressive
strength of the masonry elements resulted to the value of fwc=35 kg/cm2=3.5 MPa. The tensile strength
of the masonry elements was considered as a fraction of the compressive strength depending on the
direction of the tensile action: in the normal direction to the mortar joints was assumed equal to 0.35
MPa while in the sideway direction it was assumed equal to 0.23 MPa. The deformation characteristics
were defined as: Young modulus, Ew=1200 MPa, Shear modulus, G=500 MPa, and Poisson’s ratio,
ν=0.2
The earthquake actions were calculated according to NEAK (The National Earthquake Regulation of
Greece). The area of Patras is in Seismic Zone III and according to the regulation it has peak ground
acceleration equal to 0.24 g. More details concerning the complete analysis and design considerations
are available in the complete report of the Case Study analysis.
The evaluation of the structural performance of the building was carried out using two finite element
models: first model was an elastic model developed in SAP2000 structural analysis software; the
second and more elaborate model was created in ABAQUS software. In these models the actual
geometry of the building has been simulated in detail for capturing all relevant structural vulnerabilities
or weaknesses.
9.1.2.1. The developed numerical model The original elastic model of the Case Study was developed in SAP2000 structural analysis software.
With the use of this linear elastic model, a first estimation of the most stressed parts of the structure was
made. Despite the simplicity of the analysis conducted using this software, the results led to a better
understanding of the total structure behavior. After conducting the preliminary analysis using the
SAP2000 model, a similar nonlinear model was created in ABAQUS using concrete plasticity-cracking
models; the floor and the roof were initially not assumed as diaphragms, because the existing floor and
roof system was judged inadequate to provide such behavior to the structure. A schematic
representation of the developed model is shown in the figure 9.2.
Figure 9.2. The developed nonlinear finite element model in ABAQUS software
In the ABAQUS software, the material compressive and tensile behaviors were modeled separately.
The Concrete Damaged Plasticity model was used and the corresponding properties are presented
below. The material properties for the structural modeling of the masonry walls adopted by the designer
of the initial retrofitting technique were judged as very optimistic. After a review in the relative
literature and masonry building design codes, it was decided to use the material property values
proposed by Tomazevic. In detail, the compressive strength was assumed equal to 0.9 MPa, the tensile
strength equal to 0.21 MPa and the Young’s modulus equal to 1000 MPa. Finally, the Poisson ratio was
146
assumed as equal to 0.2 and the material density equal to 21 KN/m3. The dependence of the
compressive and tensile strength from the inelastic strains and displacements respectively as it has been
introduced in the numerical model is presented in the following graphs.
The live load applied was equal to q=2 KN/m2
uniformly distributed on the floor and a roof live load
equal to q=0.75 KN/m2
was applied too; the structural assessment of the building was carried out using
EN1998 procedure: non-linear static push-over, using a load combination for the vertical loads equal to
Gravity Load + 0.3 Live Load and finally applying the N2 method for comparing structural capacity
and seismic demand.
In order to define the response spectra to be adopted for finally assessing the structural performance,
shapes and amplification factors due to local effects from EN1998 are used. In particular, Response
Spectrum Type 1 with 5% damping was used and the peak ground acceleration ag for the Life Safety
Performance Level was taken equal to 0.23 g and for the Collapse prevention level was taken equal to
0.39 g; the other parameters defining spectrum shape and protection level have been assumed equal to
those applied to the benchmark case studies: Importance class II → γI = 1.0; Ground Type B: S = 1.2,
TB = 0.15 s, TC = 0.5 s, TD = 2.0 s.
(a)
(b)
Figure 9.3. The material behavior in compression (a) and tension (b)
9.1.2.2. Performance of the Un-retrofitted Masonry Structure Pushover analyses were conducted in the two main directions of the structure (x and z). Due to the
geometry of the structure and the resisting system differences at each direction, the structure exhibits
weaker resistance in the x-direction. It is worth recalling that the main assumption made in this analysis
is that the wall elements are un-cracked and have the nominal thickness described in the available
drawings. This implies that in the actual case, all the visible damages have to be repaired prior any
installation of the selected strengthening technique, in order to be consistent with results here presented.
The application of the N2-method clearly showed that the un-retrofitted structure was incapable of
achieving the strength requirements imposed by EC8.
As it is clearly presented in figure 9.4, the original structure failed to satisfy the two demand levels in
the x-direction, while in the z-direction, the performance curve marginally reached the demand curve
for PGA 0.23g, confirming the global structural inadequacy of the building. Therefore, a properly
designed strengthening technique has to be applied in both directions in order to satisfy the demand
levels.
The analysis showed that there is extensive cracking of the internal walls near the floor and roof lever
and detaching between the internal and external walls. Moreover, the external walls were cracked near
the wall openings (doors-windows), confirming the initial deficiency recognitions made with the use of
the elastic model in SAP2000 software and the compatibility of the results with the observed ones
presented in figure 9.5 and observed by the engineer that retrofitted the building.
147
(a)
(b)
Figure 9.4 Pushover curves (a) and Demand – Capacity curves (b) for the un-retrofitted structure
(a) (b)
Figure 9.5 Un-retrofitted structure (a) FE model, (b) cracks on the real structure
9.1.3. Intervention techniques selected for the case study The intervention technique considered for such building directly derives from those analyzed on the
benchmark building in §5; in particular the following steel techniques were planned to be applied: steel
ring beam at the roof level coupled with bracing in-plane elements for coupling all walls and creating a
strong and diaphragmatic effect. The steel profile used for the ring beam was an HEA100 fully
connected to the walls and diagonal steel ties of 12 mm used as braces. Moreover, in order to limit the
extensive damage due to the out-of-plane bending which results to plastification and failure of the
internal walls observed during the assessment, 100 mm U-profiles and steel bracings used at roof level
have been added at each floor in order to improve the diaphragmatic action of the floor. The adopted
solutions are presented in figures 9.6 and 9.7 while in the figures 9.8 the technical details about the
connection systems between steel parts and masonry building are presented.
Figure 9.7 Steel Ring beam and diagonal braces
P us hov er c urv es
0
200
400
600
800
1000
1200
1400
1600
1800
2000
0 20 40 60 80 100
T op dis plac ement [mm]
Ba
se
sh
ea
r [k
N]
masonry Zdirection
masonry Xdirection
E C 8 - Demand C urv es
0
2
4
6
8
10
12
14
0 0.02 0.04 0.06 0.08 0.1
S d [m]
Sa
[m
/se
c2]
ag= 0.39
ag= 0.23
Mas onry z
Mas onry xT*=0.19 s ec
T*=0.13 s ec
148
(a)
(b)
Figure 9.8 (a) and (b) Distribution of plastic deformations on the retrofitted structure.
(a)
(b)
(c)
Figure 9.9 Proposed connections for the adopted retrofitting techniques; (a) Diagonal brace corner
connection, (b) Top steel ring-beam connection, (c) Perimeter beam connection at the floor level
9.1.4. Assessment of the retrofitted structure The assessment of the retrofitted structure was presented in the figure 9.10; the insertion of steel
elements presented in the previous paragraph clearly showed that the structure is now capable of
satisfying performance required at Life Safety limit state and at Collapse prevention limit state.
Moreover, looking at the figures 9.8 the distribution of plastic deformations on the masonry structure
shows that only minor plastification exists at the point where the diagonal braced are connected to the
masonry walls while the rest of the walls do not present relevant stress concentrations or damages. It
can be concluded that the application of the steel-based retrofitting techniques described in the present
analysis, improve significantly the performance of the building in a cost-effective way. The feasibility
of application of the proposed techniques is high and their cost is comparable to the corresponding cost
of concrete-based intervention techniques, usually applied into the practice.
It can be also interesting comparing this proposed solution with the solution applied into the practice.
The building was retrofitted adopting a shotcrete (Gunite) coating technique for all vertical elements
because it is the most common technique used in Greece for such interventions: concrete coating had a
thickness of 5 cm and steel mesh reinforcement properly anchored to the wall by the use of steel
anchors placed in regular spacing. On the contrary the proposed technique considers only the local
repair of the wall in correspondence of existing cracks.
The existing wood made floors and roof were in good conditions when inspected by the technician and
only the deteriorated wood parts were replaced with new ones. The same approach has been considered
in the proposed approach.
Plates 1400x300x10 mm
Plates 750x300x10 mm
Ö12/300 mm
Diagonal brace Ö12 mm
Roof wooden rafters
HEA 100
2Ö6/500 mm
UPN100
Ö10/400 mm
Floor wooden joist
149
(c)
(d)
Figure 9.10 Comparison between the un-retrofitted and retrofitted structure performance. Pushover
curves and Demand – Capacity curves in (a,b) x-direction and in (c,d) z-direction
9.2 “Immaculate conception” church
9.2.1 General description of the building The intervention on historical buildings is more demanding than in cases of contemporary buildings due
to the character of European heritage and specific properties of structures and materials. In seismic
rehabilitation of historical buildings, one of the most problematic issue is the compatibility between
protection systems behavior and heritage buildings behavior, as well as the long-term compatibility
between traditional and new materials. Uncertainties due to lack of experiences with the long-term
effect of new technologies suggest that conservative approaches should be better applied in retrofit of
historical buildings. Choice of performance requirements and methods for the safety evaluation should
be chosen considering the characteristics of the particular building to be retrofitted and qualitative
verifications for the identification and elimination of major structural defects should not be discouraged
by the quantitative analytical approach proper to the modern technical Standards, as suggested by
EN1998-3.
The “Immaculate Conception” Church in Maderno (Italy) has been chosen as case study in order to
show how steel retrofit techniques can improve the seismic behavior of a simple historical building,
against both local and global collapse mechanisms. The selected case study presents characteristics
common to a large number of historical buildings spread out in Europe. The exemplificative framework
that leads to the definition of the retrofit systems includes mechanical characterization of the building,
evaluation of performance requirements, vulnerability analysis of the building, choice of the solutions
in order to reduce the building vulnerability, and safety evaluation after interventions.
The “Immaculate Conception” Church (1580) is part of the monumental complex of S. Andrew Church.
The plant is roughly rectangular, with maximum external dimensions of 7.1x13.5m, maximum external
height of 7.7 m and internal height of 6.8 m. The church has a hall composed by a nave and an apse
divided into two bays (figure 9.11). The building has a side wall in common with an adjacent building
and the terminal wall of the apse in common with the sacristy behind which can be accessed by two
small doors. The hall is covered with vaults, while the apse is a barrel vault with lunettes. There are
tension cables in correspondence of the division of the hall in two spans, under the triumphal arch, and
behind the façade. The roof is made of wood and keeps the original static scheme consisting of main
beams disposed parallel to the gutter line and of inclined secondary beams.
P us hover c urves
0
500
1000
1500
2000
2500
0 20 40 60 80 100
T op dis plac ement [mm]
Ba
se
sh
ea
r [k
N]
R oof B racesZ direction
Un-retrofittedZ direction
E C 8 - Demand C urves
0
2
4
6
8
10
12
14
0 0.02 0.04 0.06 0.08 0.1S d [m]
Sa
[m
/se
c2]
ag= 0.39
ag= 0.23
R oof R ing B eam Z
Un-retrofitted Z
150
Figure 9.11 Front view and floor plan of the “Immaculate Conception” church.
9.2.2 Assessment of the structural vulnerabilities In order to assess the resistance of the structure, the input data have been collected from a variety of
sources, including:
- available documentation specific to the building in question,
- field investigations and,
- in-situ and laboratory measurements and tests.
The following inspections have been carried out:
- N°1 sonic inspection (SO);
- N°2 visual inspection on masonry (IM);
- N°1 visual inspection on junctions (VA);
- N°1 visual inspection on the anchors (IC);
- N°2 test for the tension in cables (TC)
- N° 2 visual inspection of vaults (IV)
- N° 2 laboratory tests on masonry elements (PM).
The following mechanical characteristics of materials have been identified by visual inspection, tests on
mortar, compression tests on masonry elements:
- Mortar resistance: 1,2 MPa
- Average thickness of the mortar joints: 12mm
- Brick compression strength: 134±38 MPa
- Brick elastic modulus: 6200 MPa
- Estimated mortar Poisson modulus: 0,35
- Estimated brick Poisson modulus: 0,125
- Specific weight masonry of walls: 19 kN/m2
- Specific weight masonry of vaults: 18 kN/m2
An “Extended knowledge level” according to EN1998-3 has been reached. In order to determine the
properties of existing materials to be used in the calculation of the capacity the mean values obtained
from in-situ tests and from the additional sources of information, have been divided by the confidence
factor, CF=1.20.
The fundamental requirements refer to the state of damage expected in the structure for different levels
of earthquake actions. The performance requirements are defined by choosing
1. Levels of the seismic action
2. Accepted levels of damage
3. Safety coefficient for the verifications (closeness to the accepted level of damage)
According to the Italian National Standards, four earthquake levels and the relative Limit States are
defined. Return period for the earthquake levels are reported in Table 9.1, while in Table 9.2 the
maximum ground acceleration expected at the site are shown. The Limit States considered in the safety
assessment are:
151
- LSO = Limit State for Operational Performance Level
- LSI = Limit State for Immediate Occupancy Performance Level
- LSL = Limit State for Life Safety Performance Level
- CLS = Limit State for Collapse Prevention Performance Level
Values of q-factor associated with the accepted levels of damage for each Limit States are reported in
Table 9.3. Values of safety coefficient S for each Limit State are reported in Table 9.4. Each safety
coefficient should be interpreted as a nominal value representing the accepted distance by the
occurrence of the mechanisms associated with the Limit State.
Table 9.2. Maximum ground acceleration for the earthquake levels.
Table 9.3. Values of q-factor associated with the accepted levels of damage.
Table 9.4. Values of the safety coefficient S for each Limit State.
In order to carry out an analysis of the building and to develop an effective safety assessment, it is
essential to focus on the fundamental characteristics of the response of masonry structures to earthquake
actions. The damage mechanisms due to earthquakes can be attributed to two main categories,
depending on the response of the whole building, called first mode and second mode mechanisms. First
mode mechanisms concern with the collapse of masonry walls out of their plane, due to bending and
rocking behavior. Second mode mechanisms concern with the response of the walls in their plane, with
damage typically due to shear and bending stresses. The activation of these failure modes is highly
dependent on technological and typological characteristics of the walls. Weaknesses in the connections
between orthogonal walls and between walls and horizontal elements make the structure not able to
develop a global response during the earthquake: the individual walls have, therefore, an independent
behavior. In this case, collapse of walls is dominated by mechanisms developed outside the plane. The
presence of good connections between the walls, for example through the inclusion of tension cables,
leads to greater use of the resources of strength and stiffness in the plane of the walls. The probability of
the occurrence of out of plane mechanisms can be further reduced through the link provided by the
horizontal elements.
In case of churches, the observation of post-earthquake damages has shown that these artifacts present a
behavior that can be attributed to the analysis of architectural portions, called “macroelements”, which
show a substantially autonomous behavior in case of earthquake. For this reason, it is not very
significant to proceed through the development of analysis based on complex models and is generally
preferable to work through local verifications concerning the various macroelements which provide
information that can be attributed to first or second mode mechanisms. In order to analyze the structural
behavior taking into account the collapse mechanisms, plastic limit analysis method has been used. The
theorems of plastic limit analysis require satisfaction of certain conditions:
LSO LSI LSL CLS
ag
(m/s2)0.36 0.61 1.52 2.24
Levels of the seismic actionLEGENDA
LSO = Limit State for Operational Performance Level
LSI = Limit State for Immediate Occupancy Perf. Level
LSL = Limit State for Life Safety Performance Level
CLS = Limit State for Collapse Prevention Performance Level
LSO LSI LSL CLS
q 1.00 1.00 2.00 2.00
Accepted Level of DamageLEGENDA
LSO = Limit State for Operational Performance Level
LSI = Limit State for Immediate Occupancy Perf. Level
LSL = Limit State for Life Safety Performance Level
CLS = Limit State for Collapse Prevention Performance Level
LSO LSI LSL CLS
S 2.00 1.00 1.50 1.00
LEGENDA
LSO = Limit State for Operational Performance Level
LSI = Limit State for Immediate Occupancy Perf. Level
LSL = Limit State for Life Safety Performance Level
CLS = Limit State for Collapse Prevention Performance Level
Safety coefficient for the verifications (closeness to the accepted level of damage)
152
1. equilibrium condition: the computed internal actions must represent a state of equilibrium between
the internal and external loads (the corollary of the equilibrium conditions are compatibility conditions,
which should instead be satisfied if an energy method is being used).
2. mechanism condition: sufficient releases must be made to transform the structure into a mechanism.
3. yield condition: the stresses in the material must be everywhere less than or equal to the material
strength (e.g. shear, crushing and tensile strength limits must all be respected).
In order to evaluate the safety of the church, 10 mechanisms are considered:
- Mechanisms 1-7 are relative to vertical structures
- Mechanisms 8-10 are relative to roofing systems
In Table 9.5, results of analyses are summarized in terms of values of collapse-accelerations aC for each
mechanism.
Table 9.5. Collapse-accelerations aC for each mechanism.
The aC values should be compared with the performance requirements in terms of acceleration on the
building. With this aim, the ag values should be amplified for the amplification induced by vibrations of
the building: with this aim the amplification factor F=(1+1.5 Z/H) is used, being Z the vertical position
of the resultant horizontal force and H the height of the building. In Table 9.6, performance
requirements are summarized in terms of values of accelerations a= F ag / (q S) for each mechanism.
The safety assessment before retrofit is performed by the evaluation of the ratio aC/a (≥1 means safe) for
each Limit State. Results are reported in Table 7.
Table 9.6. values of accelerations a= F ag / (q S) for each mechanism.
M#1 M#2 M#3 M#4 M#5 M#6 M#7 M#8 M#9 M#10
aC
(m/s2)0.98 1.04 1.47 2.45 3.15 1.88 0.69 0.78 1.16 2.12
M#1 M#2 M#3 M#4 M#5 M#6 M#7 M#8 M#9 M#10
aC/aILS 0.36 0.40 0.37 0.37 0.44 0.37 0.37 0.34 0.38 0.42
aC/aOLS 0.61 1.37 1.27 1.25 1.48 1.26 1.27 1.17 1.29 1.42
aC/aLLS 1.52 1.14 1.05 1.04 1.23 1.05 1.05 0.97 1.07 1.18
aC/aCLS 2.24 1.26 1.16 1.15 1.36 1.16 1.16 1.07 1.18 1.31
153
Table 9.7. Ratio aC/a for each Limit State – before retrofit.
Collapse mechanisms that presents values of the ratio aC/a < 1 are mechanisms #1 (first mode
mechanism involving rigid rotation of a portion of the façade wall), #2 (first mode mechanism
involving rigid rotation of the façade wall), #7 (second mode mechanism of the arch), #8 (second mode
mechanisms of the vaults).
9.2.3. Intervention techniques selected for the case study Seismic strengthening of existing buildings can be achieved through the anchored ties (tension cables),
reinforced mortar joints, braced frames, bond beams, moment-resisting frames, shear walls, and
horizontal diaphragms. Traditional methods of strengthening, e.g. anchored ties, can be used
successfully, if properly designed to conform to the historic character of the building. In addition, there
are new technologies and better schemes for traditional connection devices as well as a greater
acceptance of alternative approaches to meeting seismic requirements, that can be used by ensuring that
historic buildings will not be damaged by them. For the considered case study two type of interventions
are considered: the first one is addressed to improve the connections between the wall, in order to
prevent mostly mechanisms #1, #2, #7; with the second one, a significant improvement of the
diaphragmatic effect given by vaults and the roofing system is achieved in order to prevent mostly
mechanism #8.
In order to allow the structure to manifest a satisfactory global behavior, it is necessary to improve the
connections between masonry walls, and between walls and floors and walls and roofs. This goal may
be achieved inserting tendons at the top of the building, under the vaults. An effective connection
between floors and walls is useful since it allows a better load redistribution and applies a restraining
action towards the walls overturning (figure 9.12).
Figure 9.12. Interventions to improve wall-to-wall connections.
M#1 M#2 M#3 M#4 M#5 M#6 M#7 M#8 M#9 M#10
aILS (m/s2) 2.43 2.84 3.98 5.61 8.47 5.02 2.00 2.05 2.76 4.89
aOLS (m/s2) 0.72 0.84 1.18 1.66 2.50 1.48 0.59 0.61 0.81 1.44
aSLS(m/s2) 0.86 1.01 1.42 1.99 3.01 1.78 0.71 0.73 0.98 1.74
aCLS(m/s2) 0.78 0.91 1.28 1.80 2.72 1.61 0.64 0.66 0.89 1.57
154
Since the timber roofing structure is weakly connected to walls and has a lack of resistance, a technique
that will improve the diaphragm-behavior and that can prevent sliding mechanism and collapse of the
floor has been considered. Steel diagonal ties can be installed between adjacent walls and, considered
the considerable distance between cross walls, a complete metal truss can be installed immediately
under the timber structure. Using steel anchor bolts the truss is connected to the walls (figure 9.13).
Figure 9.13. Interventions to improve roofing diaphragm-effect.
9.2.4 Assessment of the retrofitted structure The seismic safety assessment as a result of the realization of the proposed retrofit measures was
conducted through the use of local models consistent with the models used in the analysis of the
building behavior before the rehabilitation. The safety evaluation after retrofit is performed by the ratio
aC/a (≥1 means safe) for each Limit State. Results are reported in Table 9.8.
Table 9.8. Ratio aC/a for each Limit State – after retrofit.
From the data shown in Table 9.3, the effectiveness of interventions in eliminating the vulnerabilities
related to collapse mechanisms with the lowest safety factors in the present state is evident.
It also appears important to highlight that the interventions planned to reduce the vulnerabilities related
to the out of plane mechanisms of the façade and the in plane mechanisms of the arch and the vaults
lead to significant increase of the collapse accelerations for all the mechanisms, as well as of the related
safety factors.
M#1 M#2 M#3 M#4 M#5 M#6 M#7 M#8 M#9 M#10
aILS (m/s2) 4.00 5.34 6.93 6.90 9.75 7.28 3.44 3.82 5.80 11.73
aOLS (m/s2) 1.18 1.58 2.05 2.04 2.88 2.15 1.02 1.13 1.71 3.46
aSLS(m/s2) 1.42 1.90 2.46 2.45 3.46 2.59 1.22 1.36 2.06 4.17
aCLS(m/s2) 1.29 1.72 2.23 2.22 3.13 2.34 1.11 1.23 1.86 3.77
155
9.3. Bagnone building
9.3.1 General description of the building The school building is made up of three parts (blocks A, B, C) realized in the 50-60’s and divided by
structural joints; the case study object of the following analyses and studies is building A, represented in
figure 9.14. The “Bagnone building” was realized at the beginning of the 1960’s, following the
prescriptions imposed by Royal Decree 2229/1939 (Regio Decreto 16/11/1939 n. 2229, 1939).
According to this standard, specific rules for taking into account the effect of seismic action were
considered in some Italian districts, including Lunigiana and in particular Bagnone.
(a)
(b)
Figure 9.14. Plan view of case study: (a) location of studied building A; (b) structural scheme of the
building.
All columns, whose dimensions are equal to 30x45 cm, present a longitudinal steel reinforcement
composed by three bars of diameter equal to 14 mm disposed along the 45 cm length side. There are
eleven beam sections, different for shape (rectangular and L beams) and dimensions: the height of
beams varies from 24 cm (for internal beams, equal to floor thickness) to 50 cm (for external beams).
The longitudinal reinforcement of beams is made up of bars of diameters 12, 14 and 16 mm, while for
transverse reinforcing bars diameters equal to 6 and 8 mm are used. The spacing of stirrups is equal to
20 cm in all columns and vary from 15 cm to 25 cm in beams. The foundation plan of the building is
formed by a grid of inverted-T beams; only the vertical rib of the inverted-T of foundation presents
three longitudinal steel reinforcing bars whose diameter is equal to 14 mm. The floor system is a typical
Sapal floor, widely used in Italy during the 1950s-1960s and made up of brick joists with 4 longitudinal
bottom reinforcing bars ( 5 mm) contained into the brick and 2 additional longitudinal reinforcing
bars ( 12 mm) in the concrete ribs. A concrete slab of thickness equal to 40 mm without any steel
mesh completes the floor system.
With regards to not structural elements, three main categories of infill panels were individuated: double
internal or external infill of hollow bricks with internal air cavity (12+6+12 cm), simple internal infill of
solid bricks (12 cm) and external infill with multiple layers (solid bricks, internal filling with poor
concrete, external stone covering: 12+33+15 cm), respectively named in the text “infill 1”, “infill 2”
and “infill 3”. The general disposition of the internal infills is not regular
9.3.2 Assessment of the structural vulnerabilities Punctual (down-hole) and linear tests (evaluation of speed refraction for P and S waves) were executed
to establish the soil type in proximity of Bagnone building; the Vs,30 evaluated was equal to 885 m/s
and consequently the soil belonged to category A (rigid soil characterized by a speed of shear waves
higher than 800 m/s), according to what established by the actual Italian standards for constructions.
As regards mechanical properties of materials, experimental tests were executed on concrete elements
and steel reinforcement bars. Twenty-two different structural elements were tested with destroying and
not destroying tests for concrete: three columns for each storey, one column in the terrace and four
different beams including also foundation. The structural elements to test were selected according to the
prescriptions imposed by Region of Tuscany. The poor quality of concrete was highlighted by the mean
values of compression strength obtained, that were, for 2nd, 3rd and 4th storey, respectively equal to 11,
pr act icabl e t er r ace
pr act icabl e t er r ace
156
10 and 9 N/mm2 and lower than the limit imposed by Royal Decree 2229/1939 (Regio Decreto
16/11/1939 n. 2229, 1939)) and equal to 15 N/mm2.
Results of experimental tests on r.c. elements and mean values assumed for the numerical analyses and
assessments are summarized in Table 9.10, being RmT the value of compressive strength for tested
elements, Rm(i) and Em(i) the mean values of compressive strength and elastic modulus (obtained
using actual standards) of concrete elements for each floor. As regards Element ID, the first letter
indicates the type of element (P=pillar, B=beams, BF=foundation beam), the second group of letters
indicate the floor position (UF=underground floor, GF=ground floor, F1=first floor).
Two tensile tests were executed on two steel reinforcing bars of diameter 8 and 10 mm extracted from
the terrace’s columns. The results of the tests showed a yielding strength variable from 350 N/mm2 to
375 N/mm2: with reference to Royal Decree 2229/1939 (Norme per l’esecuzione delle opere in
conglomerato cementizio semplice od armato), these values suggested the use of hard steel in Bagnone
building, characterized by yielding strength equal or higher than 350 N/mm2.
Infill Typology Thickness Description fmk Em
- [cm] - [N/mm2] [N/mm2]
Infill 1 12+6+12 double hollow bricks 6 6000
Infill 2 12 simple bricks 8.6 8600
Infill 3 12+33+15 hollow bricks with cover stone 6.24 13870
Table 9.9. Description of infill panels’ characteristics.
Element ID
Not destroying test Destroying test Values for models and analysis
RmT RmT Rm(i) Em(i)
(N/mm2) (N/mm2) (N/mm2) (N/mm2)
P/UF/07 22 16
16 27594 P/UF/26 14 14
P/UF/41 - 17
P/GF/27 - 10
15 27267 P/GF/43 2 17
P/GF/48 13 15
Table 9.10. Experimental tests results and mechanical properties assumed in the analyses (only some
examples; more tests were executed on-field).
In order to better characterize the structural model of the building, before the execution of the seismic
assessment, an additional experimental programme was carried out. In particular, the global dynamic
behaviour of the building was analysed by means of EMA techniques recording the structural
accelerations under the so-called Ambient Vibrations and under impulsive forces produced by a sledge
hammer. A total of 76 measuring points (4 horizontal sensors for each level, 3 vertical sensors for each
level, 4 horizontal sensors at 2° and 3° level at structural joint with building B, 8 horizontal sensors at
2° and 3° level at corner stairs columns and 10 vertical sensors at 3° level for floor vibrations, figure
9.15) were covered using 16 accelerometers (10 PCB 3701 capacitive sensors and 6 PCB 393C
piezoelectric sensors) in different setup and a LMS SCADAS III recording device. An example of
recorded acceleration time histories and spectra, related to ambient vibrations, are reported in the figure
9.16.
The Modal Identification process was performed by means of Operational Modal Analysis techniques
such as Operational PolyMAX. It allowed to identify 5 global mode shapes listed in Table 9.11 and
illustrated in figure 9.17. It’s possible to observe that the modal deflections represent mixed flexural and
torsional displacements, probably due to the asymmetry of the building structure. The mode shapes
were also compared by the Modal Assurance Criterion (MAC), index related to mode correlation (MAC
= 100% well correlated modes, while low MAC values indicate poor correlation between modes).
Figure 9.17 shows the MAC matrix of the identified modes by which it’s possible to observe that the
identified modes present different geometrical deformation so resulting not correlated.
157
a) b)
Figure 9.15 Example of sensor locations: a) third floor; b) fourth floor.
a) b)
Figure 9.16. Example of recorded ambient vibrations: a) time histories; b) auto and cross spectra.
Mode f [Hz] T [s] [%] Description
1 2.971 0.337 1.05 Bending Y / Torsion
2 3.985 0.251 0.89 Bending X
3 5.94 0.168 0.96 Torsion
4 8.769 0.114 0.73 Bending Y / Torsion
5 11.848 0.084 0.92 Bending X / Torsion
Table 9.11. Modal properties of identified mode shapes.
After the programme carried out for completely characterizing the mechanical behaviour of the
Bagnone building, three different numerical finite element models were elaborated for analytically
reproducing the dynamic response of the case study, individuated, as previously described, through the
execution of an EMA. The models, representative of an undamaged condition, differed for the
modelling of masonry infill panels: a first preliminary model neglected the stiffening contribution of not
structural elements, introducing only their corresponding mass (Figure 9.18.a), a second model was
characterized by equivalent diagonal struts modelled (Figure 9.18.b) and a third model presented
masonry walls modelled using shell elements of thickness and mechanical properties equal to the infill
(Figure 9.18.c). The FE model (frame with equivalent truss elements) showed still some differences
with experimental modal analysis results. Thus the model was upgraded by Finite Element Model
Updating techniques optimizing the dynamic properties of the model to match at best the
experimentally identified modal features: the elastic modulus of the concrete Ec, the masonry infill wall
elastic moduli Einfill1, Einfill2 and Einfill3, boundary elastic restraint stiffness Kx and Ky (simulating
the interaction with the adjacent building B, figure 9.14.a).
In the table 9.13 are summarized the results of Model Updating showing a substantial reduction of
frequency error at the end of the process. As can be observed the updated finite element model is able to
represent the real experimentally identified dynamic behaviour in a better way than the initial model.
0.00 1700.00s
-0.07
0.07
Real
( m/s
2)
0.00
1.00
Am
plit
ude
F Time P0301:-Y
F Time P0301:+X
F Time P0303:+X
F Time P0302:-Y
F Time P0303:-Y
F Time P0302:+X
0.00 20.00Linear
Hz
0.00
460e-12
Am
plit
ude
( m/s
2)2
2.94 3.99 11.895.95 8.78
AutoPow er P0303:+X
CrossPow er P0301:+X/P0303:+X
CrossPow er P0301:-Y/P0303:+X
CrossPow er P0302:+X/P0303:+X
CrossPow er P0302:-Y/P0303:+X
CrossPow er P0303:-Y/P0303:+X
0.00 20.00Linear
Hz
0.00 20.00Hz
-180.00
180.00
Phase
°
2.94 3.99 11.895.95 8.78
158
Mode 1
Mode 2
Mode 3
Mode 4
Mode 5
MAC matrix
Figure 9.17 The first five identified mode shapes and corresponding MAC matrix.
a)
b)
c)
Figure 9.18. a) Bare frame model; b) equivalent strut model; c) shell element model.
Mode Bare frame Equivalent strut Shell element
T [s] T [s] T [s]
1 1.253 0.593 0.386
2 0.910 0.440 0.264
3 0.843 0.414 0.249
Table 9.12. First three periods for bare frame, equivalent strut frame and shell elements frame.
Mode Experimental Numerical Error
Initial Updated
[Hz] [Hz] [Hz] [%] [%]
1 2.941 2.231 2.783 24.13 5.36
2 3.947 3.521 4.093 10.79 3.69
3 5.893 5.157 5.857 12.48 0.61
4 8.718 6.899 9.656 20.86 10.76
5 11.834 7.756 10.522 34.46 11.09
Table 9.13. Comparison between experimental and numerical eigen-frequecies and related errors.
The nonlinear model of reinforced concrete Bagnone building was developed by SeismoStruct software
(Seismosoft, 2010), using force-based fiber beam-column elements and special elements for masonry
infill walls. The Mander and Menegotto-Pinto material models, available in the programme library,
were used respectively for the nonlinear modelling of concrete and steel rebars. The shear behaviour of
fiber based beam-column elements was assumed as linear elastic according to the software capabilities
159
but the resistance of the elements was monitored as damage parameter.
The mechanical properties adopted to define such nonlinear material models were derived from
experimental mechanical test performed on concrete core samples and steel rebars (see Table 9.14).
The masonry infill walls were described using the model developed by Crisafulli (1997, 2000)
implemented into SeismoStruct software by Smirou et al. (2006), using the parameters of Table 12.
The floor system was modelled with a stiff plane bracing system with elastic behaviour and truss
elements in order to let reinforced concrete beam ends free to rotate. From geological tests (seismic
refraction) the foundation soil was identified as category A (rock) according to Italian seismic standard
NTC (2008). In the figure 9.19 the complete model is shown.
Concrete Compressive
strength
fc [N/m2]
Tensile
strength
fct [N/m2]
Strain at peak
stress
em [m/m]
Collapse
strain
eu [m/m]
Foundation 20750000 0.0 0.0022 inf
Floor 0 13280000 0.0 0.0022 inf
Floor 1 12450000 0.0 0.0022 inf
Floor 2 9130000 0.0 0.0022 inf
Floor 3 8300000 0.0 0.0022 inf
Floor 4 7470000 0.0 0.0022 inf
Floor 5 11620000 0.0 0.0022 inf
Steel
Young modulus [N/m2] 206000000000
Yield strength [N/m2] 350000000
Strain hardening parameter 0.005
Table 9.14. Mechanical parameters of concrete and steel material assumed in the model.
Nonlinear Static Procedure, using the previous described model, was performed for the seismic
assessment of Bagnone building. In the figure 9.20 are illustrated the results of the pushover analysis
performed in the X direction consisting respectively in the capacity curve, the equivalent bilinear SDOF
curve (both in the force-displacement and acceleration-displacement plane) and the displacements and
interstorey drift profiles at different limit states. Similar results coming from Y direction pushover
analysis are shown in the figure 9.21. Apparently, the structure seems to satisfy the demand imposed by
the design spectra for IO, LS and CP limit states, but it was only apparent.
It can be observed that during the X-direction pushover the first shear failures of columns takes place
for a very little top displacement, while for beams it happens at a top displacements equal to 1.2 cm.
Almost all the beams and columns reached the shear failure and the development of the plastic hinges
takes place generally before in beams than in columns, see figure 9.22.
Also in the Y-direction pushover, the first shear failure of column elements takes place at nearly zero
top displacements and in beam members it is reached for 0.4 cm top displacement. From the figure 9.23
it can be observed that several other beams and columns manifest shear failure before the first plastic
hinge formation. The large amount of elements that fails before the expected target displacements for
all three considered limit states suggested that the local retrofitting is not a feasible solution and then
intervention techniques for improving the global structural performance has to be firstly considered.
160
Infill1: Double-layer
hollow bricks
masonry
Infill2: Single layer
solid brick masonry
Infill3: Multiple
layer masonry
Young modulus Em [N/m2] 6000000000 8600000000 13900000000
Compressive strength fm
[N/m2]
6000000 8600000 13900000
Tensile strength ft [N/m2] 0 0 0
Strain at max stress em 0.0012 0.0012 0.0012
Ultimate strain eu 0.024 0.024 0.024
Shear bond stress [N/m2] 300000 300000 300000
Friction coefficient 0.7 0.7 0.7
Maximum shear resistance [N] 600000 600000 600000
Thickness [m] 0.24 0.12 0.6
Strut area [m2] 0.0867 0.048 0.24
Table 9.15. Mechanical parameters of infill walls assumed in the model.
a) b)
Figure 9.19 Nonlinear model of Bagnone building: a) extruded 3D view; b) typical Y direction frame.
a) b)
c) d)
Figure 9.20 Nonlinear static procedure applied to Bagnone building (X direction): a) capacity and
equivalent bilinear curves; b) ADRS representation; c) displacement and d) interstorey drift profiles at
CP, LS, DL and IO limit states.
0
500
1000
1500
2000
2500
3000
3500
4000
4500
5000
5500
6000
0.000 0.005 0.010 0.015 0.020 0.025 0.030
Fb
* [k
N]
d* [m]
Capacity curve
Bilinear
0.00
1.00
2.00
3.00
4.00
5.00
6.00
7.00
0.00 0.05 0.10 0.15
acc
ele
rati
on
[m
/s2]
displacement [m]
LSspectrum
sdof
T*
CPspectrum
DLspectrum
dt*LS
dt*CP
demandmuLS
demandCP
IOspectrum
PUSHOVER DIR.X
0
5
10
15
20
0.000 0.005 0.010 0.015 0.020 0.025 0.030 0.035 0.040
H [
m]
displ. [m]
displ-IO displ-LS displ-CP floor1 floor2 floor3 displ-DL floor4 floor5
0
1
2
3
4
5
0.00% 0.10% 0.20% 0.30% 0.40% 0.50%
Flo
or
Interstorey Drift
IO
LS
CP
floor1
floor2
floor3
DL
floor4
floor5
161
a) b)
c) d)
Figure 9.21 Nonlinear static procedure applied to Bagnone building (Y direction): a) capacity and
equivalent bilinear curves; b) ADRS representation; c) displacement and d) interstorey drift profiles at
CP, LS, DL and IO limit states.
(a)
(b)
Figure 9.221 Capacity curve and first failures (shear, yielding and ultimate chord rotation) for column
and beam elements: a) X direction; b) Y direction.
162
10. Design guidelines
10.1. Steel buckling restrained braces Buckling restrained braces (BRB) are characterised by their ability to prevent local and overall buckling
of the brace in compression. Inelastic cyclic response of standard braces is characterised by buckling
under compression forces which leads to strength and stiffness degradation, and highly non-symmetric
response. In contrast, buckling restrained braces have a stable hysteretic response, providing a stable
and effective seismic resistant element. Most of the BRBs developed to date are proprietary, but their
principle of operation is similar. A typical BRB consist of a steel core encased in a steel tube filled with
mortar or concrete. A layer of unbonding material or a small air gap is provided between the steel core
and the mortar in order to minimise the transfer of axial forces from the steel core to the mortar and
steel tube during elongation and contraction of the steel core, and also allows for its expansion when in
compression.
Figure 10.1. The conceptual scheme of a BRB, and characteristic force-displacement relationship
10.1.1. BRB system model Such a BRB element was developed and tested at CEMSIG laboratory (UPT). The geometry and the
conceptual scheme are presented in Figure 10.2.
Figure 10.2. Geometry and components of the tested BRB, developed at CEMSIG laboratory (UPT)
Due to its high seismic vulnerability, the Steel Retro reference benchmark building was retrofitted by
means of an inverted V BRB braced system. The BRB’s, pinned at the ends, are installed in the external
frames of the RC building, as it can be seen in Figure 10.3.
10.1.2. Specific provisions in design codes From late 1999 to 2001 an AISC and SEAOC joint task group developed a document called
Recommended Provisions for Buckling-Restrained Braced Frames. The Recommended Provisions were
subsequently updated in July 2003. Since this development, buckling-restrained braces have been
included in Section 8 of the NEHRP Recommended Provisions for Seismic Regulations for New
Buildings and Other Structures, and in Section 16 of the 2005 AISC Seismic Provisions for Structural
Steel Buildings. These documents provide guidelines for the design of buckling-restrained brace
elements, connections, and make recommendations for brace testing, when it is required.
BRB steel tube
Polystyrene
Polyethylene film
BRB steel core
Concrete
163
Figure 10.3. a) STEELRETRO reference benchmark RC building model and BRB system distribution;
b) Elastic and design response spectrum
Although American standard AISC 2005 contain provisions about BRB’s, this norm also considers that
„a small amount of test data on BRBF system is available to structural engineers, it is also
recommended that engineers refer to the following documents to gain further understanding of this
system i.e. Uang and Nakashima (2003), Watanabe and others (1988), Clark and others (1999),
Tremblay and others (1999) and Kalyanaraman [12] – AISC 2005.
The AISC provisions contain: requirements about BRB design/modeling (force-displacement diagram
strength adjustment parameters) and basic requirements about experimental tests to certify BRB’s
(possible subassemblies, loading protocol).
Regarding the European guidelines or provisions about BRB’s, there are no such dates. The same
situation is in the Romanian seismic standard P100-1/2006. However, in September 2009 EN 15129
“Anti-seismic devices” was approved by CEN dealing with the general design of the dissipative devices
used in a structure. Thus, there are specified some functionality requirements, general rules of design,
material characteristics, manufacturing and testing requirements, but also conformity evaluation,
installation and maintenance conditions.
In order to have a control on BRB’s modeling and analysis, the following parameters should be
established.
In an elastic analysis, a BRB can be modelled using an elastic truss element (when a pinned connection
is used, or when stiffness of a rigid connection is neglected in analysis) or a frame element.
In this particular case, the BRB design started with a steel core cross section of minimum 3 cm2 (1 cm
thickness and 3 cm wide) and it was made according to European EN 1993-1-1 [15] design rules taking
into account the provisions from American codes (AISC2005 /NEHRP200). The design axial strength
of a BRB can be written as (in Eurocode 3 notation, adapted from AISC 2005a):
0
ysc sc
ysc
M
f AP
(10.1)
where: yscf - specified minimum yield stress of the steel core, or actual yield stress of the steel core as
determined from a coupon test, N/mm2; scA - net area of steel core, mm
2; 0M - partial safety factor (
0 1.1M ).
The relationships between the brace overall strain (εwp) and the inter-story drift θ can be approximated
as:
wpθ sin2
ε2
(10.2)
In order to assure a homogeneous dissipative behavior of the diagonals, it should be checked that the
maximum overstrength (Ωi) does not differ from the minimum value Ω by more than 25%. The
following BRB core plate cross section were obtained:
in X direction: ground floor = 2cm x 4 cm; 1st level = 1cm x 4cm; 2nd level = 1cm x 3cm.
in Ydirection: ground floor = 2cm x 3cm; 1st level = 1cm x 5cm; 2nd level = 1cm x 3cm.
Taking into account the variation of cross-section of the BRB described above, variation of core cross-
sectional area should be accounted for in analysis. The BRB cross section is represented in the model as
constant along the length. Therefore, a reduction of the axial stiffness K [KN/m] is applied. However,
0
1
2
3
4
5
6
7
0 0.5 1 1.5 2 2.5 3 3.5 4T[s]
Se
(T),
Sd
(T)
q=4 (BRB)
TB TC
TD
q=1.5 (RCF)
TB TC
TD
164
some authors suggested approximating brace stiffness to the one of the yielding segment alone, as most
of the elastic deformations and all of the plastic ones are concentrated here (Clark et al., 1999 [10]).
Seismic reduction factor (q) for spectral analysis:
In order to perform an elastic global analysis the seismic load was defined by EN-1998-1 elastic
spectrum, with the peak ground acceleration (PGA) of 0.23g, I=1.0, TB=0.15s, TC=0.5 s, TD=2.0 s, S =
1.2. For the original reinforced concrete structure, a seismic behaviour factor q = 1.5 was used. For the
reinforced concrete structure retrofitted with BRB system, the seismic behaviour factor q amounted to 4
(see Figure 10.3.b). Based on standard analogies the seismic reduction factor (q) was taken to be equal
with 4. The American standards AISC 2005 and NEHRP 2003 recommend a force reduction factor R=8
(where R is the equivalent of the q factor in Eurocode 8) for Buckling Restrained Braced Frames
BRBF, Moment Resisting Frames MRF and Eccentrically Braced Frames EBF. As in Eurocode 8 there
is no reference for BRB systems, a q factor equal to 6 was initially adopted for BRB framing, similar to
that of MRF and EBF systems. However, the q factor defined according to previous codes is valid for
the design of new steel buildings. Romanian Seismic Evaluation standard recommends for existing RC
buildings a q factor equal to 2.5 and a q factor equal to 4 for existing EBF. Therefore, it was considered
more appropriate to take an average value of the q factor, 2.5<q<4. Thus, considering that BRBS has an
adequate contribution to the system, a q factor of 4 was considered.
BRB main modeling parameters (ductility (µ), strain hardening adjustment factor (ω) and compression
adjustment factor (β))
As it concern the modeling, the design and the acceptance criteria of a BRBS for new/existing
buildings, it should be mentioned that there is no “public” standard, in order to assure their
functionality; this is made only based on experimental tests and the “good” experience of people
involved in this domain.
When modelling a BRB for a nonlinear static analysis, two factors are to be accounted for in addition to
the initial stiffness. The first one is the compression-strength adjustment factor, , reflecting higher
strength in compression in comparison with the strength in tension. The second one is the tension
strength adjustment factor, , accounting for strain hardening (AISC 2005b). Both factors are intended
for computation of maximum forces in tension Tmax and in compression Pmax that can be developed by
the BRB, for design of connections and beams and columns. Yield strength in tension Ty is determined
as (using Eurocode notations):
y ov ysc scT f A (10.3)
where: Ty – yield strength in tension of the BRB; ov - material overstrength factor, to account for the
possibility that the actual yield strength of steel is higher than the nominal yield strength.
Up to date, design provisions for buckling restrained braces require that brace design be based on
qualifying tests (AISC 2005a, NEHRP 2003). Therefore, yscf is determined directly from tensile tests,
and material overstrength factor ov need not be considered. A simple bilinear model based on the
above consideration is shown in figure 10.4. This force-displacement relationship can be incorporated
in a nonlinear truss element in order to obtain a complete model of a BRB for a pushover analysis.
Figure 10.4. a) Diagram of brace deformation versus inter-storey drift angle relationship; b) Bilinear
modelling of BRB (AISC 2005b)
For this particular case and a BRB cross section made of S235 steel, the geometry of the core was
defined so that all braces have the same active length of 1.7 m. Thus, for this active length and the end
165
restraints, the yield displacement amounts to Δy = 1.9 mm. The estimation of the ultimate displacement
Δu was based on the results of the experimental tests carried on BRB elements. Based on these results,
ductility ratios Δu/Δy were estimated for tension and compression amounted to 22, as the average of the
values obtained from AISC cyclic loading protocol. In order to obtain the adjustment of the design
strengths (maximum compression strength Cmax and maximum tension strength Tmax), the following
formulas were applied:
max y yT = R f A ; max y yC = R f A (10.4)
where, fy is the yield strength, Ry is the ratio of the expected yield stress to the specified minimum yield
stress fy (may be considered equal to 1).
The values of the compression adjustment factor β=1.2 and a strain hardening adjustment factor ω=1.9
was obtained from the experimental tests, using the following formulas:
max
max
C =
T ;
max
fysc
T =
f A
(10.5)
where: fysc is the measured yield strength of the steel core.
BRBS acceptance criteria (needed in order to establish a PBSD (Performance Based Seismic Analysis) for
retrofitting a RC MRF GLD building:
In the authors' view, general acceptance criteria for BRBs are difficult to be established based on the
existing data from literature because BRB’s are rather manufactured than built. That is, they are
typically made by a specialty manufacturer, rather than by a contractor or steel fabricator (although
such a method of producing BRB’s is possible). Design of BRBs is required to be validated by tests,
and therefore performance criteria can be established on a case-by-case basis. In fact, the purpose of acceptance criteria for an element (BRB in our case), is to establish some “points” on force-
deformation relation (table 10.1 and figure 10.5) where the element is considered to be in IO, LS or CP stage.
Thus the acceptance criteria are based on the American FEMA356/ASCE41. To have some starting indicative
values, another option is to use the values for braces in tension, recommended by FEMA (Table).
Table10.1. Steel Braces in Tension Acceptance Criteria for Nonlinear Procedures (FEMA356)
Figure 10.5. Generalized Force-Deformation Relation for Steel Elements or Components (FEMA356)
In the case of the design of BRB’s for seismic upgrading of RC structures, the performance criteria of
this device depend on the RC lateral displacement response. RC frames generally yield for an
interstorey-drift of about 1%, while the performance criterion for Collapse Prevention corresponds to
2.5% for a seismic event with a 10% probability of occurence in 50 years (10/50). Then, assuming a
brace ductility capacity in the range of maxy=4÷8, BRB’s should be designed as to yield for an
interstorey-drift of 0.25% (obtained by dividing an interstory drift of 1% per the ductility capacity ) in
a 10/50 seismic event. In this way, the maximum displacement demand corresponds to the first RC
damaging. While, in case of a 2/50 seismic event (i.e. with a 2% probability of occurence in 50 years),
it seems conservative not to exceed twice the ductility capacity considered for a life safety design.
In table 10.2 a, some indicative values of core plastic deformation ratio max/y that may be appropriate
to a performance based design are reported. The symbols IO, LS and CP are in the place of Immediate
Occupancy, Life Safety and Collapse Prevention, respectively.
166
Table 10.2. Acceptance Criteria for BRB’s.
In this case, the modelling parameters (β, ω, µ) were obtained from the experimental tests on BRB
specimens developed at CEMSIG laboratory (UPT) (see Table 10.3). The BRB system acceptance
criteria were based on FEMA356/ASCE41 - for steel braces in tension, adapting the ductility of around
22Dt, (which is twice the value given by FEMA 356, i.e. 11Dt).
Based on Bilinear modelling of BRB (AISC 2005b), the inelastic behaviour of BRB system was
modelled considering the concentrated tri-linear plasticity curve with strain hardening and strength
degradation of 0.8 from maximum capacity, according to FEMA356 (see figure 10.6).
Figure 10.6. BRB tri-linear model: a. on X direction; b. in Y direction
The BRB tri-linear model used in the present analysis is characterized by the following parameters
(tablr 10.3):
Table 10.3. BRB modeling parameters for the final benchmark analysis
10.1.3. Connections Detailing and design of connections between BRBs and the existing structure is highly dependent on the
particular type of structure to be strengthened (steel, r.c. or masonry).
Brace connections are to be designed with sufficient overstrength with respect to the brace, in order to
keep it free of damage. AISC 2005a requires the brace connection (in new steel BRB frames) to be
designed for a force equal to 1.1 times the adjusted brace strength in compression Pmax .
10.2. Design guideline for Steel Shear Wall as seismic retrofit measure In this section design and construction rules for Steel Shear Walls for seismic retrofitting and upgrading
are summarized.
10.2.1. General description of the retrofitting technique Steel Shear Walls (SSW) consist of a thin shear panel surrounded by a frame of beams and columns,
Figure left. These boundary elements can be connected either hinged or rigid to each other. It can be
shown that for SSW’s with a rather small span L, only hinged connected elements leads to a sufficient
BRB (fy=235 N/mm2) force - displacement - on X direction
-400
-200
0
200
400
-0.08 -0.06 -0.04 -0.02 0 0.02 0.04 0.06 0.08
Displacement [m]
Fo
rce
[K
N]
BRB ground floor [2x4] cm2 BRB 1'st level [1x4] cm2 BRB 2'nd level [1x3] cm2
Compression
Tension
BRB (fy=235 N/mm2) force - displacement - on Y direction
-300
-200
-100
0
100
200
300
-0.08 -0.06 -0.04 -0.02 0 0.02 0.04 0.06 0.08
Displacement [m]
Fo
rce [
KN
]
BRB ground floor [2x3] cm2 BRB 1'st level [1x5] cm2 BRB 2'nd level [1x3] cm2
Compression
Tension
Final Benchmark analysis
Modeling Curve type triliniar (FEMA/ASCE model)
Material steel S235
Aria-core c.s. Ac [cm2] 1x3 (tested cross section)
Core length Lc [m] 1.7
Yielding displacement Δy [mm] 1.9
Ductility displacement µ 22 (cyclic AISC)
IO 0.5Δt
LS 14Δt
CP 18Δt
BRB effective stiffness Ke considered
Compression adjustment
factor β
1.2 (minimum from cyclic
ECCS+AISC)
Acceptance criteria
(modified FEMA356/ASCE41
acceptance criteria for
braces in tension)
BRB properties
Strain hardening adjustment
factor ω
1.9 (minimum from cyclic
ECCS+AISC)
167
design. Furthermore, it can be advantageous to subdivide the SSW in several areas by stiffeners to
obtain favourable L / h-ratio of about 1 and to reduce the bending forces in the boundary elements (see
Figure). The infill plate is the energy dissipating element, which starts to buckle and to yield during the
seismic action. Thereby, the plate develops a tension field. The boundary frame absorbs the forces of
the plate and should be designed to stay elastic during cyclic loading. In frames with rigid connected
elements the plastic hinges should be formed at the end of the beams.
10.2.2. Pre-Design, modelling and assessment rules for Steel Shear Walls
10.2.2.1. Pre-Design The most established model to analyse SSW’s is the strip model based on Thorburn et al. (1983). It
idealizes the shear panel by pinned tension stripes as shown in Figure (right).
The design of the steel SSW is an iterative process, as the angle of the tension strips need to be
recalculated and the model to be revised. To simplify this procedure it is adequate to estimate the angle
of inclination to = 40°. This leads to an accurate ultimate capacity and a slightly conservative elastic
stiffness. The maximum base shear force of a SSW with hinged connected boundary elements can be
determined by :
)2sin(2
1 LtfV wy
(10.6)
where fy = yield stress of shear panel, tw = thickness of shear panel, L = distance between vertical
boundary element centrelines and = angle of the tension field measured relative to the vertical.
Figure 10.7. Elements of the Steel Shear Wall and idealized strip model by Thorburn et al. (1983)
The stiffness of the SSW can be calculated by:
)2(sin4
1 2
h
LtEK w (10.7)
where E = Elastic modulus of shear panel, h = distance between horizontal boundary element
centrelines and other terms are as previously defined.
Knowing the required base shear force, equation (10.7) can be used to determine the shear panel
thickness (or the distance between the vertical boundary element centerlines). So if the frame geometry
is given, equation (10.6) results to:
tw 2 V
fy L sin(2) (10.8)
If the stiffness is the governing parameter, the shear panel thickness (or the distance between the
vertical boundary element centerlines) can be determined by equation (10.7):
)2²(sin
4
LE
hKtw
(10.9)
The aspect ratio has to be in the range of 0.8 < L / h ≤ 2.5. Furthermore the limit on the slenderness of
Frame PlateMoment
+
Strip model
L
h
h
168
the shear panel should be limited to:
(10.10)
It has to be mentioned that some SSW´s performed sufficiently without fulfilling expression (10.10), so
it can be seen as a conservative recommendation. To assure that the shear panel forms a fairly uniform
tension field the vertical and horizontal boundary elements require a sufficient flexural stiffness.
Therefore the “column flexibility parameter“ h for the vertical boundary elements and the “End (top
and bottom) panel flexibility parameter“ L for the horizontal boundary elements of the SSW have to be
in given limits. These parameters establish a relationship between the boundary elements flexural
stiffness and the deviation of the shear panel tension field from the uniform case.
h 0,7 htw
2 L Ic4
(10.11)
L 0,7h4
IcL4
Ib
tw
4 L4
(10.12)
where Ic = moment of inertia of the vertical boundary element, Ib = moment of inertia of the horizontal
boundary element and other terms are as previously defined. h is limited to be smaller that 2,5 and L
smaller than 2,5 for the top horizontal boundary element and smaller than 2,0 for the bottom horizontal
boundary element.
With the limit on h the minimum moment of inertia of the column results indirectly to:
(10.13)
Furthermore, the web thickness of the boundary elements should be higher than the thickness of the
shear panel.
10.2.2.2. Modelling After the pre-design based on the required base shear force or stiffness, the SSW can be modelled by
non-linear beam elements with the strip model (e.g. for push-over analysis). When using the strip
model, a sufficient number of strips for an appropriate modelling of the plates is 10.
The angle of inclination of the tension field can be established by the following equation:
tan2
1tw L
2 Ac
1 tw h1
Ab
h3
360 Ic L
(10.14)
The strip model can also be used to verify the capacity of the boundary elements, where capacity design
rules have to be applied by considering the expected overstrength of the shear panel.
The available ductility of SSW’s is mainly dominated by the material properties of the shear panel and
its connection to the boundary elements. For ordinary steel grades a member ductility of = 4 can be
considered, if sufficiently designed welded connections or connections by fasteners are used. The
application of low yield point steel can increase the member ductility up to 8.
10.2.2.3. Connection between shear panel and boundary elements The connection between shear panel and boundary elements can be established by welds, bolts or
powder actuated fasteners. The connection is to be designed for the yield strength of the shear panel
considering the angle of the tension field, while capacity design rules have to be applied. A satisfactory
overstrength can be assumed for welded connections designed according to EN1993-1-8, but also for
connection with fasteners designed according to EN1993-1-5, if the shear panel is crimped in the
connection area and the yield ratio fu / fy is sufficient high (e.g. 1.5). If connections with bolts or powder
actuated fasteners can not provide a sufficient capacity to capture the overstrength of the shear panel,
ductile failure modes (e.g. hole bearing) instead of brittle failure modes (e.g. shear failure) has to be
guaranteed. Furthermore, in such cases the reduced strength capacity of the shear panel has to be
considered.
min(L,h)
tw 25
E
fy
Ic 0,00307 tw h
4
L
169
10.2.2.3.1. Connection of Steel Shear Wall to existing structure The connection to the existing structure has to transfer the horizontal as well as vertical forces and it has
to enable deformations of the shear wall. However, also the existing structure has to be able to carry the
additional forces introduced by the retrofit measure. Hence, it has to be decided where and how
horizontal and/or vertical forces are transferred. Additional load transfer beams has been found as
favourable as they enable to direct the forces to parts of the existing structure with a sufficient capacity.
Insert through anchoring designed has been validated as favourable rigid connecting system for RC-
structures due to the high capacity and the possibility to balance tolerances.
mkusdRk fAkV /,, (10.15)
where k = 0.8 for group behaviour, = 0.4 for concrete strength ≤ C20/25, As = section area of anchor,
fu = tensile strength of anchor.
The assembling procedure of the SSW connected by a transfer beam and insert through anchoring to the
existing RC-structure can be summarized as follow:
1. Core drilling in RC-frame
2. Erection of steel shear wall and transfer-beam
3. Insertion of anchors
4. Grouting of rods
Figure 10.8. Connection to existing structure: (1) only horizontal forces, (2) horizontal and vertical
forces, (3) with additional transfer beam
170
11. Results, general conclusions and perspectives The research project dealt with the complex problem of defining appropriate intervention techniques for
existing buildings, a not simple task given that in the design practice all retrofitting interventions can be
considered as unique because of particular boundary/environmental conditions that the building has.
Nevertheless, the research consortium tried to face the problem suitably combining different tools and
methods in order to have a systematic approach and at the same time an experimental programme was
also carried out for developing and testing retrofitting techniques to be proposed as valuable solutions to
the practitioners.
In particular, during the research project the following steps (assumed as ‘methodology’) have been
followed in order to systematically treat the seismic retrofitting of existing constructions:
1. defining a framework for surveying existing constructions and recognizing potential
vulnerabilities;
2. choosing a PBEE methodology, composing together design strategy, hazard model, modelling
techniques, simulation method, acceptance criteria (i.e. FEMA 356 and EN1998), technical
aspects and economic model for cost estimation;
3. defining a matrix approach that have been used a first pre-selecting method for analysing most
common techniques (also not steel based) and individuating those that were technically not
convenient (i.e. accessibility, difficulty level for applicability, manpower skill for in-field
works, demolition, previous technical evidences…);
4. defining two benchmark structures on which different steel solutions, pre-selected or derived
from the application of the matrix approach at point 3, have been applied (using chosen PBEE)
and the results of such applications have been so able to be compared;
5. analysis of the structural response at the foundation level, evaluating the required bearing
capacity of the foundations and designing of the intervention techniques;
6. considering the upgraded foundation system applied to the structures, definition of a simplified
soil-structure interaction model and re-analysis of the complete retrofitted structures in order to
secure the reached safety level, previously determined, and eventually optimize the structural
elements in the upper structure.
In general, these steps should be considered as mandatory for every designer engaged in the seismic
retrofitting of the existing constructions, considering that this sequence of steps has been applied to
different structural systems in the research project, confirming the applicability of the methodology.
In particular, the knowledge phase of the structure – step 1 – it is always a fundamental process that is
usually executed in a different way according to personal skills or to different structural types. The step
1 of the methodology adopted in the research could support the designer in this phase, because it faces
the approach to the structural system irrespectively of the types or of the configuration, in a quite
systematic way. At the end of this logic process, the potential vulnerabilities and the structural parts on
which focusing the investigations can be highlighted and the structural assessment can be executed,
using calculus method that designer considers much more appropriate inside the vulnerability
framework herein adopted.
Another important step is the selection of retrofitting techniques to be analysed and the designers should
look at those techniques that, first of all, are characterized by technical feasibility if examined in the
perspectives of the preliminary information obtained from the preliminary vulnerability assessment of
the existing construction to be retrofitted. Also in this case, practitioners are often used facing the
problem without a general approach or with a partial analysis; the step 2 of the methodology here
proposed tried to answer to his point in a simplified way, applicable in the practice, but maintaining a
systematic approach. The designer can use the matrix approach, considering the (qualitative) variables
that for him have more importance to compare and preselect the techniques before the application of
PBEE that requires a high computational effort.
The steps 3, 4, 5 and 6 are those related to the application of the PBEE and, above all, to the execution
of numerical analyses for sizing the retrofitting techniques, quantifying their effectiveness and
completing the design process. Of course, the step 1 and step 2 are fundamental in the methodology
because their information drive the development of the next phase of the design process.
The application of the methodology to several techniques has allowed, in the first steps, to pre-select
those more interesting and afterwards has allowed the final assessment of seismic performance of those
more performing: Steel bracing configurations; parallel steel frames; BRB bracing configurations; shear
171
steel walls; light gauge steel walls; steel strips. Moreover, it has been also executed an economic
comparison between different techniques in order to appreciate the impact of costs of the different
solutions.
The complete application of the methodology to those different techniques as allowed also the accurate
analysis of three steel based intervention techniques and the designing of three base cases, sized on the
same benchmark structure – r.c. – that have been subjected to experimental testing. The test
programme, in particular, has been focused on the retrofitting of r.c. concrete structures but the results
and the techniques could be directly extended and applied to masonry structures also.
The three techniques experimentally tested have been:
Buckling Restrained Bracing system; - BRB
Shear Steel wall (with innovative connection system); - SSW
Flag Shaped Hysteretic Dissipative Bracing system with re-centering capabilities. - FSHD
All these three techniques have been selected from the previous numerical simulations because they can
effectively answer to the problems related to the retrofitting of existing constructions, in which strength,
stiffness and ductility deficiencies could be detected contemporary or separately, obliging the designers
for looking at different techniques for addressing such deficiencies singularly, coupled or altogether. In
particular, the development of such techniques and their application to the benchmark structures
allowed verifying their flexibility in grading mechanical properties (i.e. strength, stiffness and ductility),
confirmed also by experimental testing programme carried out in three different laboratories.
Moreover, it also important to stress that one of the major problems of seismic retrofitting is the
localization of stresses/forces that pass from existing structure to the new ones (retrofitting system) and
this phenomena is as much pronounced as less stiffness and strength cannot be controlled into the
retrofitting systems. This aspect has been taken into account; in fact, BRB system and FSHD system do
not localize high level of forces due to their intrinsic possibility of modifying their yielding threshold
and their initial stiffness, through a refined sizing of their internal components. The SSW system in
general are considered as retrofitting techniques characterized by high stiffness (only), high resistance
and by imposing an high resistance demand on surrounding columns, obliging so the designers to costly
and complex local retrofitting actions. These shortcomings from SSW system have been brilliantly
solved defining a novel mechanically composed system in which steel panels can be taken from a wide
variety of qualities (i.e. automotive <1mm to structural >3mm), graduating so the strength and the
stiffness. Moreover, the system is connected to the structure using a beam system connected to the floor
and able to do not create over-turning moments; in such a way, the surrounding columns and the beams
are not overloaded by the retrofitting scheme.
These three techniques represent solutions with a high technological and conceptual contents and their
flexibility proposes those as appropriate for the application of PBEE to the seismic retrofitting of
existing constructions (i.e. grading structural response of retrofitted structures with the different
earthquake intensities and correlating them with expected building performance). Moreover, design
guidelines have been developed for BRB system and for SSW system, while the guidelines for FSHD
system are still under development due to the patenting process at which this system has been subjected.
At the end of the research project, some real case studies have been analysed in order to individuate
their vulnerabilities and proposing retrofitting techniques between those analysed during the research.
The STEELRETRO project presented as main general outcome the development of steel based
techniques endowed with high technological content; in particular, two of those are novel techniques
and one of those is subjected to a patenting process.
Moreover, the development of these techniques has required the definition of a ‘real’ and ‘technically
sound’ working environment in order to develop, size and assess these techniques using
applicable/feasible methods and to compare their performance with real or representative demands.
For such a reason, inside the STEELRETRO project a methodology for approaching to the problem of
the seismic retrofitting has been set up, combining together several tools for treating/managing the
various aspect that a seismic retrofitting always involve. In particular, the methodology has been
defined following the logical process that a good practitioner should follow.
172
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List of figures
Figure I. General flow-chart of the research……………………………………………………………...7
Figure II. General framework in which the vulnerabilities identification were inserted…………………8
Figure III. (a) decisional matrix for the judgment of a single solutions; (b) summarizing tables of
intervention techniques for floor systems…………………………………………………….9
Figure IV. (a) r.c. benchmark building; (b) FEM model of r.c. benchmark for structural assessment…10
Figure V. (a) masonry benchmark building; (b) ABAQUS FEM model for structural assessment…….10
Figure VI. (a) hot-rolled steel plates; (b) BRB system; (c) light gauge steel walls; (d) elastic bracings;
(e) eccentric bracing systems; (f) bracing system with additional dissipative devices……..11
Figure VII. (a) parallel steel frame; (b) braced steel frame; (c) insertion of steel strips inside masonry;
(d) modification of roof diaphragmatic action: very stiff the roof and deformable the
floors………………………………………………………………………………………...12
Figure VIII. Study on techniques for improving existing foundations: (a) micro-pile model; (b)
geotechnical information of soil characterization…………………………………………...13
Figure IX. (a) Floor deformation equipped with different techniques; (b) roof in-plane deformation…13
Figure X. (a) full-scale testing on BRB +R.C. Frame systems; (b) initial qualification of material
properties……………………………………………………………………………………14
Figure XI. (a) steel shear wall coupled with r.c. frame; (b) preliminary tensile tests on steel qualities; (c)
tested coupon………………………………………………………………………………..14
Figure XII. (a) testing on steel quality; (b) FSHD system; (c) buckling restraining system for steel
fuses…………………………………………………………………………………………14
Figure XIII. (a) first tests of FSHD system – not satisfactory behaviour/modification of the system; (b)
and (c) two examples from second series of tests carried out modifying internal properties of
the system (note: to shorten the test procedure only 1 cycle was executed for each
displacement level)………………………………………………………………………….15
Figure 1.1. Type of steel ribbed bars analyzed during the data collection: (a)Thor steel; (b) RUMI steel;
(c) star shaped steel; (d) ribbed bar………………………………………………………….25
Figure 1.2. Analysis of test certificate produced in 1962 by official laboratory in Pisa: (a) grouping by
bar type; (b) grouping by steel qualities…………………………………………………….25
Figure 1.3. Statistical analysis on 1962 production, Aq42 steel: (a) yielding stress, (b) elongation at
fracture………………………………………………………………………………………26
Figure 1.4. Demolished buildings: (a) pillars of Villafranca building; (b) Workers Union building…...26
Figure 1.5. Tensile testing of steel bars sampled from demolished buildings: (a) RUMI steel – end of
‘60s; (b) rounded bars – ‘20s………………………………………………………………..26
Figure 1.6. (a) correlation between tensile strength and Mn content; (b) linear regression between
mechanical properties (measured) and a possible chemical-data based model……………..27
Figure 1.7. (a) compressive tests on small cylinder; (b) statistical evaluation of the results……………27
Figure 2.1. Performance Based Engineering framework and Performance Based Assessment sub-
framework…………………………………………………………………………………...29
Figure 2.2 Mean Return Periods (TR, MRI) and expected maximum ground acceleration ag…………..32
Figure 2.3 Generalized Component Force-Deformation Relations for Depicting Modeling and
Acceptance Criteria………………………………………………………………………….35
Figure 2.4 Complete procedure for applying the non-linear static analysis method and interpreting the
results in terms of capacity and demand…………………………………………………….36
Figure 2.5 Seismic safety evaluation of buildings using nonlinear analysis………………..37
Figure 2.6. Analysis of the concept of strengthening solutions…………………………………………38
Figure 3.1 Enhance the deformation capacity of the building…………………………………………..41
Figure 3.2 Data concerning with intervention techniques using typological analysis…………………..41
Figure 3.2. (a) Installation of Near Surface Mounting GFRP bars; (b) Rectangular FRP grids; (c)
Application examples of CAM; (d) New r.c. slab on existing floor deck; (e) Steel braces for
stiffening of floor systems; (f) In-field execution of ring-beam technique; (g) Typical
application of reinforced concrete jacketing to r.c. columns; (h) Reinforced concrete
jacketing of beam……………………………………………………………………………44
Figure 3.3. (a) Realization of new reinforced concrete shear wall; (b) Buckling Restrained Brace; (c)
application of steel bracings system; (d) Dissipative steel eccentric bracing; (e) insertion of
external micro-piles with the addition of reinforced concrete cap; (f) Micropile Enhancement
177
to Existing Strip Footing…………………………………………………………………….45
Figure 4.1. Reinforced concrete benchmark building: (a) first floor plan, (b) second floor plan……….51
Figure 4.2. Reinforced concrete benchmark building: (a) third floor plan view, (b) foundations………52
Figure 4.3. Typical main frame of the structural scheme in the reinforced concrete benchmark……...52
Figure 4.4. Typical secondary frame of the structural scheme in the reinforced concrete benchmark…52
Figure 4.5. Masonry benchmark building – plan views: (a) first floor; (b) second floor……………….53
Figure 4.6. Section views: (a) C-C section; (b) B-B section……………………………………………54
Figure 4.7. (a) A-A section view of the building; (b) particular of floor systems at the last floor under
the roofing system…………………………………………………………………………...54
Figure 4.8. (a) Confined (i.e. inside the reinforcing cage) and (b) un-confined (i.e. outside of reinforcing
cage) concrete material properties………………..................................................................55
Figure 4.9. Reinforced concrete material nonlinear model based on Kent and Park; (b). modified Park
nonlinear model of steel reinforcement……………………………………………………..56
Figure 4.10. Deformation controlled action model with nonlinear load-deformation parameters and
acceptance criteria (FEMA356)……………………………………………………………..56
Figure 4.11. Effective stiffness of RC-elements according to the FEMA356…………………………..57
Figure 4.12 Moment-rotation curve for section 1 by section analysis and FEMA 356 with
nonconforming (NC) and conforming (C) transverse reinforcement……………………….57
Figure 4.13 Stress-strain models adopted in OPENSEES: (a) reinforcing steel; (b) concrete (slightly
confined)…………………………………………………………………………………….57
Figure 4.14 Cross section fiber subdivision: (a) subdivision in different zones; (b) definition of the
concrete fibres; (c) position of steel reinforcement…………………………………………58
Figure 4.15 Equivalent truss system for floor modelling……………………………………………….58
Figure 4.16. Calibration of the constitutive model for masonry in the ABAQUS software…………….59
Figure 4.17 FEM model of the benchmark building realized using ABAQUS software……………….59
Figure 4.18 (a) 3D model . deformed shape; deformation in the last captured step: (b) X, (c) Y
direction………………………………………………………………………..……………60
Figure 4.20. Static pushover curves of the 3D frame in the X and Y direction with identification of
several failure modes……………………………………………………………………......60
Figure 4.21. Application of ADRS method for seismic performance assessment in X, Y direction…...61
Figure 4.22. (Y-Y) direction stresses in the masonry from vertical loads………………………………62
Figure 4.23. Vertical load vs. vertical displacement…………………………………………………….62
Figure 4.24. Plastic-strain/cracking pattern at failure for (a) X direction and (b) Z direction
pushover……………………………………………………………………………………..63
Figure 4.25. Deformations in the points of figure 4.24. vs. the base shear in (a) X direction and (b) Z
direction loading…………………………………………………………………………….63
Figure 5.1. Optimal bracing configuration for the “regular building” (type 1)…………………………68
Figure 5.2. Optimal bracing configuration for the “dumpbell shaped building” (type 2a)……………..69
Figure 5.3. Optimal bracing configuration for the “L-shaped building” (type 2b)……………………...69
Figure 5.4. Optimal bracing configuration for the “asymmetric re-entrant profile building” (type
3a)…………………………………………………………………………………………...70
Figure 5.5. Optimal bracing configuration for the “symmetric re-entrant profile building” (type 3b)…70
Figure 5.5. Optimal bracing configuration for the “symmetric re-entrant profile building” (type 3b)…71
Figure 5.7. Different bracing configurations in terms of path of forces………………………………...71
Figure 5.8. (a) STEELRETRO reference benchmark RC building model and BRB system distribution
(b) Elastic and design response spectrum…………………………………………………...72
Figure 5.9 Geometry and components of the tested BRB (CEMSIG)…………………………………..72
Figure 5.10. BRB tri-linear model: a. on X direction; b. in Y direction………………………………...73
Figure 5.10.a. Performance of the Benchmark building retrofitted using different techniques (global
approach – BRB – and local strengthening – FRP)…………………………………………74
Figure 5.11. Possible strengthening strategies by shear walls for the RC-benchmark building………...75
Figure 5.12. Type of analysed shear walls: steel shear wall with rigid connections (a), with hinged
connections (b), with flanges (c), composite shear wall (d)………………………………...76
Figure 5.13. Possible strengthening with shear walls, ground view…………………………………….76
Figure 5.14. Possible strengthening with shear walls, Section axis A and E…………………………...77
Figure 5.15. Possible strengthening with shear walls, Section axis 1 and 6…………………………….77
Figure 5.16. Structural model for shear wall……………………………………………………………77
178
Figure 5.17. Load-displacement characteristic of shear wall…………………………………………...77
Figure 5.18. Base shear force-displacement curves in X-direction (4 span), strategy A………………..78
Figure 5.19. Storey drift over the height of the structure in X-direction (4 span), strategy A………….78
Figure 5.20. Base shear force-displacement curves in Y-direction (5 span), strategy A………………..78
Figure 5.21. Storey drift over the height of the structure in Y-direction (5 span), strategy A…………78
Figure 5.22. Demand spectra vs. capacity diagram in X-direction (4 span), strategy A………………..78
Figure 5.23. Demand spectra vs. capacity diagram in Y-direction (5 span), strategy A………………..78
Figure 5.24. Base shear force-displacement curves in X-direction (4 span), strategy B………………79
Figure 5.25. Storey drift over the height of the structure in X-direction (4 span), strategy B…………..79
Figure 5.26. Base shear force-displacement curves in Y-direction (5 span), strategy B………………79
Figure 5.27. Storey drift over the height of the structure in Y-direction (5 span), strategy B…………..79
Figure 5.28. Demand spectra vs. capacity diagram in X-direction (4 span) for retrofitting strategy B...79
Figure 5.29. Demand spectra vs. capacity diagram in Y-direction (5 span) for retrofitting strategy
B……………………………………………………………………………………………..79
Figure 5.30. Partial-width shear walls: a) configuration C; b) configuration D………………………...80
Figure 5.31. Nonlinear Static Analysis of C retrofitting configuration: a) and c) ADRS representation
(pushover X and Y); b) and d) interstorey drift profiles (pushover X and Y)………………81
Figure5.32. Nonlinear Static Analysis of D retrofitting configuration: a) and c) ADRS representation
(pushover X and Y); b) and d) interstorey drift profiles (pushover X and Y)………………81
Figure 5.33. Suggested use of the LGS steel shear walls.........................................................................82
Figure 5.34. Possible strengthening with LGS shear walls (a) W1, (b) W2.............................................82
Figure 5.35. Deformed shape before failure from pushover in (a) X and (b) Y directions......................83
Figure 5.36. Demand and capacity diagram of the equivalent SDOF system (Annex B, EN 1998)........83
Figure 5.37. Modeling the LGS shear walls as inclined strips (W2-Strips)…………………………….83
Figure 5.38. Pushover curves of the W2 and W2-Strips configurations………………………………..83
Figure 5.40. Demand and capacity diagram of the equivalent SDOF system (Annex B, EN 1998): (a, c)
X and Y direction of the W2 model, (b, d) X and Y direction of the W2-strip model...........84
Figure 5.41. Capacity & demand of structure with LGS wall & roof......................................................85
Figure 5.42. Adopted concentric bracing scheme and cyclic behaviour………………………………..86
Figure 5.43. Eccentric bracing systems: a) adopted scheme, b) finite element model, c) shear and d)
bending behaviour of the link……………………………………………………………….86
Figure 5.44. Concentric bracing schemes: a) X direction; b) Y direction……………………………...87
Figure 5.45. X direction retrofitting solution: a) ADRS format representation, b) collapse mechanism
and ductility assessment……………………………………………………………………..87
Figure 5.46. Y direction retrofitting solution: a) ADRS format representation, b) collapse mechanism
and ductility assessment……………………………………………………………………..87
Figure 5.47. X1 eccentric bracing scheme and link properties………………………………………….88
Figure 5.48. Eccentric bracing schemes analyzed in the Y direction with adopted link properties…….88
Figure 5.49. X retrofitting solution: a) capacity curve in ADRS format representation, d) collapse
mechanism and ductility assessment………………………………………………………..88
Figure 5.50. Y retrofitting solution: a) capacity curve in ADRS format representation, d) collapse
mechanism and ductility assessment…………………………………………………….….89
Figure 5.51 Abaqus model of masonry benchmark building: (a) 3D model: (b) deformed shape at
collapse; (c) constitutive law in compressione; (d) constitutive law in tension……………89
Figure 5.52 Pushover deformations with 24mm, fy=350N/mm2 tying at the top of the walls……….90
Figure 5.53. Pushover curves of structure tied at top with 24mm, fy=350N/mm2 ties. (a) X (b) Z
direction……………………………………………………………………………………..91
Figure 5.54 Views of the deformed shape and distribution of tension cracking for (a) X and (b) Z
direction pushover……………………………………………………………...……………91
Figure 5.55 PSASD plot vs. pushover curve transformed in SDOF format (a) X & (b) Z direction…...92
Figure 5.56 (a) Technical solution for horizontal LGS strips and (b) expected working principle……..92
Figure 5.57. Deformation shapes and distribution of tension cracks for LGS model. (a) X direction and
(b) Z direction pushover…………………………………………………………………….93
Figure 5.58. Comparison of pushover curves without and with LGS strengthening of selected external
walls (i.e. diaphragm provided only at roof level)…………………………………………..93
179
Figure 5.59. Comparison of pushover curves without and with LGS strengthening of selected external
walls (i.e. diaphragm provided only at each slab)…………………………………………..93
Figure 5.60. Deformed shape and tensile cracking pattern for (a) X and (b) Z direction pushover…….94
Figure 5.61. Scheme of retrofitting technique: coupling of masonry building using steel elements…...94
Figure 5.62 Retrofitting technique using coupled steel Moment resisting frames. (a) masonry (b) steel
(c) masonry and steel………………………………………………………………………..95
Figure 5.63 Demand-Capacity diagram according the EN1998-1-1 spectrum.........................................95
Figure 5.64. Application of vertical bracings: a) 3d view; b) lateral view of the bracings......................95
Figure 5.65. X retrofitting solution: a) capacity curve in ADRS format representation, d) collapse
mechanism and ductility assessment…………..……………………………………………96
Figure 5.66. Y retrofitting solution: a) capacity curve in ADRS format representation, d) collapse
mechanism and ductility assessment……………………….……………………………….96
Figure 5.67.a Performance obtained using BRB technique in an optimized application……………….97
Figure 5.67.b Performance obtained using CB technique – limited ductility / more strength – in an
optimized application………………………………………………………………………..98
Figure 5.67.c Performance obtained using EBF technique –ductility / strength – in an optimized
application…………………………………………………………………………………...98
Figure 5.67.d Performance obtained using LGS technique –ductility / strength – in an optimized
application…………………………………………………………………………………...99
Figure 5.67.e Performance obtained using Shear Wall technique –ductility / strength – in an optimized
application…………………………………………………………………………………...99
Figure 5.68 Total cost of the intervention for sm of useful floor area and costs of the four selected
economic parameters………………………………………………………………………101
Figure 5.69 Influence of each voice on the total costs of the intervention techniques………………...102
Figure 5.70 Connection technique between braces and existing elements using pre-tensioned elements
and limiting the holes drilling inside main structural elements……………………………103
Figure 6.1. (a) 3D model of the masonry benchmark; (b) model of the floor system…………………105
Figure 6.2. Check Point at Floor – Roof……………………………………………………………….106
Figure 6.3. Deflection reduction of Floor, (a), and Roof (b) systems comparison…………….………107
Figure 6.4. Horizontal displacement reduction – Floor systems………………………………………107
Figure 6.5. Details of connecting systems for application of intervention techniques………………...109
Figure 6.6. External post tensioning…………………………………………………………………...112
Figure6.7. Additional steel bracings……………………………………….…………………………..112
Figure 6.8. Steel plate collectors……………………………………………………………………….113
Figure 7.1. Stratigraphic profile of Type C soil………………………………………………………..116
Figure 7.2. FE model of micropiles……………………………………………………………………116
Figure 7.3. Configurations of micropiles and P-d curves……………………………………………...118
Figure 7.4 Retrofit solutions for the foundation system……………………………………………….119
Figure 7.5 (a)typological scheme of the intervention technique with micro-piles; (b) in-field work for
realizing connection system between micro-piles and existing foundation………………120
Figure8.1. Load deformation curves and failure modes of tension tests on connections: series 1 (top),
series 2 (middle) and series 3 (bottom)…………………………………………………….125
Figure 8.2. General layout of Steel Shear Walls as retrofit measure of a RC-frame (test 4 and 5)……126
Figure 8.3. Test set up of test 5 and load deformation curves of test 1 to 5…...………………………127
Figure 8.4. First cracks next to the welds (left) and buckling behaviour at 80 mm (middle) as well as at
the end of the test (right) (Test 2)…………………………………………………...……..128
Figure 8.5. Cracks through the net section area of the section (left) and buckling behaviour at 36 mm
displacement (Test 3)………………………………………………………………………128
Figure 8.6. Relative resistance function of test 2 to 5………………………………………………….129
Figure 8.7. Resistance drop ratio function of test 2 to 5……………………………………………….129
Figure 8.8. Connection between SSW and RC-frame: Transfer beam and insert through anchoring (left),
hinged connection between transfer beam and SSW (right)……………………………….130
Figure 8.9. Test set-up for connection in masonry wall……………………………………………….131
Figure 8.10. Load deformation curves of connections in masonry wall with two different thicknesses
d……………………………………………………………………………………………132
Figure 8.11. a) RC frame location - 3D view; b) RC elements cross sections (columns and beam)…..132
Figure 8.12. RC frame and node details: a) rebars bent in the joints; b) formwork of the concrete
180
frame……………………………………………………………………………………….132
Figure 8.13. a) Theoretical vs. quality certificate vs. experimental rebars samples material
characteristics; Characteristics of the concrete used for: b) RC frame; c) BRB infill
material…………………………………………………………………………………….132
Figure 8.14. BRB steel plate specimens, material characteristics of the BRB steel core plates and stress-
strain curves for BRB steel core material………………………………………………….133
Figure 8.15. CBS steel plate specimens, material characteristics of the BRB steel core plates and stress-
strain curves for BRB steel core material…………………………………………………133
Figure 8.16. Testing rig and the loading system: a) scheme of the testing rig; b) RC portal frame and
BRB system (MRF+BRB); b) RC portal frame and CBS system (MRF+CBS)…………..133
Figure 8.17. Connection details of: a) BRB and RC column; b) BRB - RC beam; c) CBS and RC
column; d) CBS - RC beam………………………………………………………………..134
Figure 8.18. Monotonic tests: a) MRF; b) MRF+BRB; c) MRF+CBS………………………………..134
Figure 8.19. Monotonic tests results…………………………………………………………………...135
Figure 8.20. a) RC frame under cyclic load; b) development of bending cracks……………………...135
Figure 8.21. RC frame under cyclic load: a) development of shear cracks; b) failure of the node……135
Figure 8.22. a) MRF + BRB under cyclic load, b) bending moment cracks, c) shear cracks at ultimate
stage………………………………………………………………………………………..136
Figure 8.23. a) MRF + CBS under cyclic load, b) bending moment and shear cracks……………….136
Figure 8.24. Hysteretic curve of the connection between: a) the BRB – RC beam; b) CBS – RC
beam………………………………………………..………………………………………136
Figure 8.25. The initial RC frame vs. the retrofitted frames……………………………………….….137
Figure 8.26. a) Left BRB during cyclic test; b) Right BRB during cyclic test………………………..137
Figure 8.27. BRB steel core plates during cyclic test………………………………………………….137
Figure 8.28. (a) dissipative fuses; (b) testing set-up; (c) buckling restraining system for testing……..138
Figure 8.29. Cyclic testing on different steel qualities at different maximum strain………………......138
Figure 8.30. Prestressing cable………………………………………………………………………...139
Figure 8.31. a) Dissipative element b) buckling restraining system……………………………….......139
Figure 8.32. Global view and sections of external case………………………………………………..139
Figure 8.33. Global view and sections of internal sliding frame………………………………………140
Figure 8.34. Connecting plates………………………………………………………………………...140
Figure 8.35. Piston……………………………………………………………………………………..140
Figure 8.36. General test setup………………………………………………………………………...141
Figure 8.37. Sensor position…………………………………………………………………………...141
Figure 8.38. Displacement history used for the short testing procedure………………………………142
Figure 8.39. First experimental tests: no satisfactory result due to different behaviour in tension and in
compression………………………………………………………………………………..142
Figure 8.40. Loss of contact between the anchor plate and the welded sheet…………………………143
Figure 8.41. C-formed element used to assure the contrast…………………………………………....143
Figure 8.42. (a) pre.stress 50% - steel fuses fy=350N/mm2 and section equal to 450 mm
2; (b) pre-stress
60% - steel fuses fy=200N/mm2 and section equal to 300 mm
2………………………...…143
Figure 9.1. Plan drawing (a) and general view (b) of the structure…………………………………....145
Figure 9.2. The developed nonlinear finite element model in ABAQUS software…………………....146
Figure 9.3. The material behavior in compression (a) and tension (b)………………………………...147
Figure 9.4 Pushover curves (a) and Demand – Capacity curves (b) for the un-retrofitted structure…..148
Figure 9.5 Un-retrofitted structure (a) FE model, (b) cracks on the real structure…………………….148
Figure 9.7 Steel Ring beam and diagonal braces………………………………………………………148
Figure 9.8 (a) and (b) Distribution of plastic deformations on the retrofitted structure……………….149
Figure 9.9 Proposed connections for the adopted retrofitting techniques; (a) Diagonal brace corner
connection, (b) Top steel ring-beam connection, (c) Perimeter beam connection at the floor
level………………………………………………………………………………………...149
Figure 9.10 Comparison between the un-retrofitted and retrofitted structure performance. Pushover
curves and Demand – Capacity curves in (a,b) x-direction and in (c,d) z-direction………150
Figure 9.11 Front view and floor plan of the “Immaculate Conception” church……………………...151
Figure 9.12. Interventions to improve wall-to-wall connections………………………………………154
Figure 9.13. Interventions to improve roofing diaphragm-effect……………………………………...155
Figure 9.14. Plan view of case study: (a) location of studied building A; (b) structural scheme of the
181
building…156
Figure 9.15 Example of sensor locations: a) third floor; b) fourth floor………………………………158
Figure 9.16. Example of recorded ambient vibrations: a) time histories; b) auto and cross spectra…..158
Figure 9.17 The first five identified mode shapes and corresponding MAC matrix…………………..159
Figure 9.18. a) Bare frame model; b) equivalent strut model; c) shell element model………………..159
Figure 9.19 Nonlinear model of Bagnone building: a) extruded 3D view; b) typical Y direction
frame……………………………………………………………………………………….161
Figure 9.20 Nonlinear static procedure applied to Bagnone building (X direction): a) capacity and
equivalent bilinear curves; b) ADRS representation; c) displacement and d) interstorey drift
profiles at CP, LS, DL and IO limit states………………………………………………....161
Figure 9.21 Nonlinear static procedure applied to Bagnone building (Y direction): a) capacity and
equivalent bilinear curves; b) ADRS representation; c) displacement and d) interstorey drift
profiles at CP, LS, DL and IO limit states…………………………………………………162
Figure 9.222 Capacity curve and first failures (shear, yielding and ultimate chord rotation) for column
and beam elements: a) X direction; b) Y direction………………………………………...162
Figure 10.1. The conceptual scheme of a BRB, and characteristic force-displacement relationship….163
Figure 10.2. Geometry and components of the tested BRB, developed at CEMSIG laboratory
(UPT)………………………………………………………………………………………163
Figure 10.3. a) STEELRETRO reference benchmark RC building model and BRB system distribution;
b) Elastic and design response spectrum…………………………………………………..164
Figure 10.4. a) Diagram of brace deformation versus inter-storey drift angle relationship; b) Bilinear
modelling of BRB (AISC 2005b)………………………………………………………….165
Figure 10.5. Generalized Force-Deformation Relation for Steel Elements or Components
(FEMA356)………………………………………………………………………………...166
Figure 10.6. BRB tri-linear model: a. on X direction; b. in Y direction……………………………….167
Figure 10.7. Elements of the Steel Shear Wall and idealized strip model by Thorburn et al. (1983)…168
Figure 10.8. Connection to existing structure: (1) only horizontal forces, (2) horizontal and vertical
forces, (3) with additional transfer beam…………………………………………………..170
182
List of Tables
Table 1.1. Requirement for steel reinforcement adoption in structural design – 1957-1972…………...24
Table 1.2. Chemical, metallographic and mechanical properties compared……………………………26
Table 2.1. Performance matrix for the definition of global building performance……………………...31
Table 2.2 Earthquake hazard level; PE - Probability to exceed; MRI - Medium recurrence interval…..33
Table 2.3 Comparison of the design strategies proposed by different standard………………………...33
Table 2.4 Building performance objectives for use in STEELRETRO project…………………………39
Table 3.1 Decisional Matrix condensing all relevant aspects for a preliminary judgment of the structural
intervention technique. Legend for scoring L = low, M = medium, H = high; Mark – L (5-6),
M (7-8), H (9-10)……………………………………………………………………………42
Table 3.2. Typological form to be adopted with the decisional matrix in the preliminary selection of
intervention technique – form filled for ring beam technique for roof in masonry
building……………………………………………………………………………………...42
Table 3.3 (a) typological analysis on micro-piles intervention on foundations; (b) typological analysis
on horizontal bracings for floor/roof stiffening……………………………………………..43
Table 3.4. Masonry wall typologies and main limitations of rehabilitation method; Yes - Possible to use
the method for both restoration and strengthening; Int - Only on the interior surface of the
wall; *- If the wall had plastering which can be remade than S or “-”; A – Applicable; NA –
Not Applicable; SC – Special Care; G – Good; IM – Intermediate; P – Poor; M – Major; S –
Small; - – None………………………………………………………………………..46
Table 3.5 Flooring systems in masonry building and main limitations of rehabilitation method………47
Table 3.6 Roofing systems in masonry building and main limitations of rehabilitation method……….47
Table 3.7 Foundation systems in masonry building and main limitations of rehabilitation method…...47
Table 3.8 Roofing systems in masonry building: suitability of rehabilitation methods………………...48
Table 3.9 Roofing systems in masonry building: Improvements due to rehabilitation methods……….48
Table 3.10 Flooring and roofing systems in r.c. buildings: Applicability of analyzed techniques to floor
types…………………………………………………………………………………………48
Table 3.11 Flooring and roofing systems in r.c. buildings: Non Structural Properties of analyzed
techniques…………………………………………………………………………………...49
Table 3.12. Suitability for foundation typologies in r.c. and main limitations of rehabilitation method;
Yes - Possible to use the method for strengthening; A – Applicable; NA – Not Applicable;
SC – Special Care; M – Major; S – Small; - – None………………………………………..49
Table 3.13. Suitability for foundation typologies in r.c. and failure mechanism improved by the
rehabilitation method………………………………………………………………………..49
Table 3.14 Suitability for foundation typologies in r.c. and failure mechanism improved by the
rehabilitation method………………………………………………………………………..50
Table 4.1. Mechanical properties of masonry materials in benchmark building………………………..59
Table 4.2. Maximum displacement, required and available ductility determined from different
software……………………………………………………………………………………...61
Table 4.3. Recognition of main structural vulnerabilities in the r.c. benchmark………………………..61
Table 5.1. Mechanical characteristics of the elementary frame………………………………………....67
Table 5.2. Mechanical characteristics of the elementary frame…………………………………………68
Table 5.3: BRB modelling parameters for the final benchmark analysis……………………………….73
Table 5.4. Parameters of steel shear walls for strengthening strategy A and B…………………………77
Table 5.4 Mechanical parameters of shear walls in configuration C……………………………………80
Table 5.5 Mechanical parameters of shear walls in configuration D…………………………………...80
Table 5.6 LGS shear walls in X and Y directions....................................................................................82
Table 5.7 Distribution of the horizontal loads in the 3D structure……………………………………...85
Table 5.8 Summary of the properties of the equivalent SDOF (Annex B, 1998) in all strengthening
cases…………………………………………………………………………………………86
Table 5.9 Costs for each voice obtained from the Italian prices of Commerce Chambers………….....100
Table 5.10 Total cost and cost breakdown for all the optimized solutions…………………………….100
Table 5.11 Relative influence of each single voice on the total……………………………………….101
Table 7.1 Mechanical parameters of Type C soil…………………………………….………………..116
Table 7.2 Characteristics of micropiles……………………………………………….……………….117
Table 7.3 Configurations of micropiles………………………………………………….…………….117
183
Table 7.4 Spring labeling for configurations of micropiles…………………………………………...118
Table 7.5 Characteristics of retrofit solutions for foundation…………………………………………119
Table 8.1 Test program on connections and mechanical properties of the tested shear panels; *) yield
strength measured in longitudinal and orthogonal direction of rolling……………………124
Table 8.2 Test program on full scale Steel Shear Wall………………………………………………..126
Table 8.3 Steel qualities selected for realizing steel fuses preliminary tested………………………...138
Table 9.1 Earthquake levels…………………………………………………………………………...145
Table 9.2 Maximum ground acceleration for the earthquake levels…………………………………..152
Table 9.3 Values of q-factor associated with the accepted levels of damage…………………………152
Table 9.4 Values of the safety coefficient S for each Limit State…………………………………….152
Table 9.5 Collapse-accelerations aC for each mechanism…………………………………………….153
Table 9.6 values of accelerations a= F ag / (q S) for each mechanism………………………………..153
Table 9.7 Ratio aC/a for each Limit State – before retrofit……………………………………………154
Table 9.8 Ratio aC/a for each Limit State – after retrofit……………………………………………...155
Table 9.9 Description of infill panels’ characteristics………………………………………………...157
Table 9.10 Experimental tests results and mechanical properties assumed in the analyses (only some
examples; more tests were executed on-field)……………………………………………..157
Table 9.11 Modal properties of identified mode shapes……………………………………………....158
Table 9.12 First three periods for bare frame, equivalent strut frame and shell elements frame……..159
Table 9.13 Comparison between experimental and numerical eigen-frequecies and related errors….159
Table 9.14 Mechanical parameters of concrete and steel material assumed in the model……………160
Table 9.15 Mechanical parameters of infill walls assumed in the model……………………………..161
Table 10.1 Steel Braces in Tension Acceptance Criteria for Nonlinear Procedures (FEMA356)…….166
Table 10.2 Acceptance Criteria for BRB’s…………………………………………………………....167
Table 10.3 BRB modeling parameters for the final benchmark analysis……………………………...167
184
List of Acronyms
a acceleration
AD Acceleration displacement
ADRS Acceleration displacement response spectrum
AISC American institute of steel construction
BRB Buckling restrained
CBF Concentric braced frame
CBS Concentrically braced system
CP Collapse prevention
CSM Capacity spectrum method
D Dimension
DL Damage limitation
EBF Eccentric braced frame
EC Euro code
EMA Environmental monitoring assessment
EQ Earthquake
FE Finite element
FEM Finite element method
FEMA Federal emergency management agency
FRP Fiber reinforced product
FSHD Flag shaped hysteretic device
GR Greece
IO Immediate occupancy
LGS Light gauge steel
LGSW Light gauge shear wall
LS Life safety
LVDT Linear variable differential transformer
MAC Modal assurance criterion
MRF Moment resisting frame
MRI Medium recurrence interval
NC Near collapse
NEAK National Earthquake Greek Regulation
NEHRP National Earthquake Hazard Reduction Program
NSP Non structural properties
PBA Performance based assessment
PBD Performance based design
PBD Performance based design
PBE Performance based engineering
PBEE Perfomance based earthquake engineering
PE Probability of exceedance
PGA Peak ground acceleration
PVC Polyvinyl chloride
R.C. Reinforced concrete
RO Romania
SC Structural classification
SD Significant damage
SDOF Single degree of freedom
SEAOC Structural Engineering Association Of California
SSW Steel shear wall
185
TC Technical classification
TR Return period
WP Work package
186
European Commission EUR 25894 — Steel solutions for seismic retrofit and upgrade of existing constructions (Steelretro) Luxembourg: Publications Office of the European Union 2013 — 186 pp. — 21 × 29.7 cm ISBN 978-92-79-29046-6doi:10.2777/7937
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doi:10.2777/7937
KI-NA-25894-EN
-N
The majority of existing buildings are in need of seismic retrofit. The main reasons are: the original design was not optimised with respect to the required safety level, poor construction quality, modifications or enlargements of buildings during their life and increase in the requirements of the seismic design. Even if steel solutions can often be more efficient and economic, their possibilities are practically unknown and their application has been limited to a few particular cases. The aim of the research proposal focused to set up steel solutions for the seismic retrofit of existing buildings, furnishing design and construction methodologies, tools for dimensioning of elements and connections.
Studies and reports
